ML19290C976
| ML19290C976 | |
| Person / Time | |
|---|---|
| Site: | Trojan File:Portland General Electric icon.png |
| Issue date: | 02/13/1980 |
| From: | Broehl D PORTLAND GENERAL ELECTRIC CO. |
| To: | Schwencer A Office of Nuclear Reactor Regulation |
| References | |
| TAC-07551, TAC-11299, TAC-7551, NUDOCS 8002150276 | |
| Download: ML19290C976 (94) | |
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February 13, 1980 Trojan Nuclear Plant Docket 50-344 License NPF-1 Director of Nuclear Reactor Regulation ATTN:
Mr.
A.
Schwencer, Chief Operating Reactors Branch 41 Division of Operating Reactors U.
S.
Nuclear Regulatory Commission Washington, D.
C.
20555
Dear Sir:
Attached are 40 copies of the following items which respond to requests from your staff in telephone conversations during the week of January 28, 1980:
(1)
" Effects of Gross Bending on Shear Wall In-Plane Capacities" (2)
" Vertical Shear Transfer, and Single Curvature Horizontal Shear Capacities and Displacements" (3)
"The Load and Resistance Paths Between the Concrete Core and the Block of Composite Walls" (4)
" Dead Load Reductions Due to Shrinkage" (5)
" Capacity-to-Force Ratios for New Structural Elements" (6)
" Sliding Resistance of Wall Panels" (7)
" Temporary Loadings on Walls During Modification Program" (8)
" Resistance to Overturning and Sliding of the Modified Complex" (9)
" Potential Frequency Shift Due to Gross Bending and Vertical Shear Transfer Mechanisms" (10)
Supplemental response to NRC Ouestion 6 dated September 14, 1979 which supersedes Licensee's previous response dated December 22, 1979 to that question.
(11)
Revised figures for Licensee's response dated June 29, 1979 to NRC Ouestion 3 dated May 18, 1979.
(12)
Corrected Response Spectra to replace some pages of Attach-ment 21-1, provided with Licensee's response dated December 21, 1979 to NRC Question 2L 7 7ted October 2, 1979.
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Mr.
A.
Schwencer Washington, D.
C.
Page 2 of 2 We wish to make clear that in some areas these responses do not represent what Licensee believes to be the most appropriate way to evaluate the modified Complex.
We continue to believe that PGE-1020 provides a conservative approach to analysis of the structural behavior of the modified Complex.
This infor-mation is provided in order to be responsive to verbal requests from your staff.
Sincerely, Y
Donald J.
Br 11 DJB:jgl Attachments c:
Mr.
R.
11. Engelken, Director U.
S.
Nuclear Regulatory Cammission Region V Mr. Lynn Fr itk, Director State of Oregon Departme'it of Energy
UNITED STATES OF AMERICA NUCLEAR REGULATORY COMMISSION BEFORE THE ATOMIC SAFETY AND LICENSING BOARD In the matter of
)
)
PORTLAND GENERAL ELECTRIC COMPANY,
)
Docket No. 50-344 et al.
)
(Control Building Proceeding)
)
(Trojan Nuclear Plant)
)
CERTIFICATE OF SERVICE I hereby certify that on February 13, 1980, Licensee's letter to the Director of Nuclear Reactor Regulation dated February 13, 1980 with information in response to NRC Staff requests has been served upon the persons listed below by depositing copies thereof in the United States mail with proper postace affixed for first class mail.
Marshall E.
Miller, Esq., Chairman Joseph R.
Gray, Esq.
Atomic Safety and Licensing Board Counsel for NRC Staff U.S. Nuclear Regulatory Commission U.S.
Nuclear Regulatory Washington, D.C.
20555 Commission Washington, D.C.
20555 Dr. Kenneth A. McCollum, Dean Division of Engineering, Maurice Axelrad, Esq.
Architecture and Technology Lowenstein, Newman, Reis, Oklahoma State University Axelrad & Toll Stillwater, Oklahoma 74074 1025 Connecticut Ave.,
N.W.
Suite 1214 Dr. Hugh C.
Paxton Wa shing ton,
D.C.
20036 1229 - 41st Street Los Alamos, New Mexico 97544 Frank W. Ostrander, Jr.,
Esq.
Assistant Attorney General Atomic Sa fety and Licensing Board State of Oregon Panel Department of Justice U.S. Nuclear Regulatory commission 500 Pacific Building Washington, D.C.
20555 520 S.
W. Yamhill Portland, Oregon 97204 Atomic Safety and Licensing Appeal Panel William Kinsey, Esq.
U.S. Nuclear Regulatory Commission Bonneville Power Admin.
Washington, D.C.
20555 P.O.
Box 3621 Portland, Oregon 97208 Docketing and Service Section (3)
Of fice of the Secretary U.S. Nuclear Regulatory Commission Washington, D.C.
20555 CE-13
CERTIFICATE OF SERVICE Ms. Nina Bell Mr. Eugene Rosolie 728 S.E. 26th Avenue Coalition for Safe Power Portland, Oregon 97214 215 S.E.
9th Avenue Portland, Oregon 97214 Mr. John A.
Kullberg Route 1, Box 2500 Columbia County Courthouse Sauvie Island, Oregon 97231 Law Library Circuit Court Room Mr. David B.
McCoy St. Helens, Oregon 97051 348 Hussey Lane Grants Pass, Oregon 97526 Dr. Harold I.
Laursen 1520 N.W.
13th Ms.
C. Gail Parson Corvallis, Oregon 97330 P.O.
Box 2992 Kodiak, Alaska 99615 a
Ronald W.
J son Assistant Gene Counsel Portland General Electric Company Dated:
February 13, 1980 CE-13
r
-- Ell PortlandGeneralElectricCompany February 13, 1980 Donaa; Eces en tact v.ce Pe w:
Trojan Nuclear Plant Docket 50-344 License NPF-1 Director of Nuclear Reactor Regulation ATTN:
Mr.
A.
Schwencer, Chief Operating Reactors Branch #1 Division of Operating Reactors U.
S.
Nuclear Regulatory Commission Washington, D.
C.
20555
Dear Sir:
Attached are 40 copies of the following items which respond to requests from your staff in telephone conversatidns during the week of January 28, 1980:
(1)
" Effects of Gross Bending on Shear Wall In-Plane Capacities" (2)
" Vertical Shear Transfer, and Single Curvature Horizontal Shear Capacities and Displacements" (3)
"The Load and Resistance Paths Between the Concrete Core and the Block of Composite Walls" (4)
" Dead Load Reductions Due to Shrinkage" (5)
" Capacity-to-Force Ratios for New Structural Elements" (6)
" Sliding Resistance of Wall Panels" (7)
" Temporary Loadings on Walls During Modification Program" (8)
" Resistance to Overturning and Sliding of the Modified Complex" (9)
" Potential Frequency Shift Due to Gross Bending and Vertical Shear Transfer Mechanisms" (10)
Supplemental response to NRC Question 6 dated September 14, 1979 which supersedes Licensee's previous response dated December 22, 1979 to that question.
(11)
Revised figures for Licensee's response dated June 29, 1979 to NRC Question 3 dated May 18, 1979.
(12)
Corrected Responsa Spectra to replace some pages of Attach-ment 21-1, provided with Licensee's response dated Decembe r 21, 1979 to NRC Question 21 dated October 2, 1979.
c 2 Src, kee :cuna Oer, rr.:
Podiand General ElectricCongiaoy Mr. A.
Schwencer Washington, D.
C.
Page 2 of 2 We wish to make clear that in some areas these responses do not represent what Licensee believes to be the most appropriate way to evaluate the modified Complex.
We continue to believe that PGE-1020 provides a conservative approaca to analyr.is of the structural behavior of the modified Complex.
This infor-mation is provided in order to be responsive to verbal requests from your staff.
Sincerely, Donald J.
Br 1
DJB:jgl Attachments c:
Mr.
R.
H.
Engelken, Director U.
S. Nuclear Regulatory Commission Region V Mr. Lynn Frank, Director State of Oregon Department of Energy
A.
Schwencer February 13, 1980 Page 1 of 2 EFFECTS OF GROSS BENDING ON SHEAR WALL IN-PLANE CAPACITIES The effect of the gross bending on capacity is to reduce the capacity in the tension zones and to increase the capacity in the compression zones.
A similiar effect occurs with the sti f fness; that is decreasing stiffness with decreasing axial load.
In normal design practice, these ef fects are neglected since the decrease in capacity on one side of the shear wall is normally compensated for by the increase in-capacity on the other side.
In addition, as this effect is developing, the behavior of the structure is nonlinear, resulting in energy dissipation which can be approximated by an increase in damping.
Such an increase in dampino would result in lower forces in the walls than those obtained from the linear elastic solution with 2% damping for the OBE.
In order to address these ef fects realistically an extension of the state of the art of analysis in design of shear walls systems would be required.
However, in an attempt to evaluate the effects of this behavior a simplified analysis was per fo rmed on the R and N walls between elevations 45 and 61.
The simpified analysis considered the wall subjected to the vertical forces from the gross bending ef fect which causes tension on one side and compression on the other, direct dead load reduced by the vertical component of the earthquake, and the horizontal shear forces from the finite element analysis.
A new distribution of horizontal shear forces was calculated based on the stif fness reduction factors given in DE-8
Page 2 of 2 Appendix B of PGE-1020.
Considering the inertia loads directed north, the results of this approximation indicate a reduction in the load on the wall section between column lines 51 and 55 of 21% and an increase in loads on the wall section north of column line 51 o f 19 %.
The capacity to force ratio for these wall segments remain greater than one.
The load redistribution in the N wall was considered in a similar manner, a nd again the capacity to force ratios remain greater than one.
Since the capacity to force ratios are at or near their minimum values at this elevation and the effect of the redistribution did not cause the capacity to force ratios to reduce below one at these locations, it is concluded that the capacity to force ratios will not be reduced below one at other locations.
The OBE loads were not reduced for increased damping and the capacities were calculated as indicated in PGE-1020.
Similiar results would be obtained if the capacities were calculated using the single curvature approach.
DE-8
A.
Schwencer February 13, 1980 Page 1 of 8 VERTICAL SHEAR TRANSFER, AND SINGLE CURVATURE HORIZONTAL SHEAR CAPACITIES AND DISPLACEMENTS Vertical shear transfer at wall-column interfaces and horizontal shear capacities and displacements resulting from the single curvature assumption are discussed below.
Vertical Shear Capacity at Column Lines The vertical shear capacities of the walls at the embedded column lines arc calculated as the summation of the frictional component of the beam-column connections at the respective floor levels and the shear-friction of the horizontal reinforc-ing steel crossing the shear plane.
The capacity at any ele-vation is the cumulative capacity from the roof level down to that elevation.
The deformations associated with each of the two capacity components and their mutual compatibility are discussed in detail in Licensee's response dated December 22, 1979 to NRC Staff Question 6, datec October 2, 1979.
In that response it was also stated that in order to mobilize the entire capacity corresponding to the shear-friction of the horizontal reinforcing steel, the amount of relative vertical displacement is approximately 0.01 inch.
Figures 1 through 4 show the vertical shear forces for 0.15g OBE at the column lines of the major shear walls in the Control Building and their corresponding capacities calculated as above.
The vertical shear forces corresponding to a factored OBE condition are also drawn.
As can be seen, the unfactored OBE demand loads are less than the calculated capacities DH-10
Page 2 of 8 indicating thereby that the actual vertical relative defor-mation will be proportionately lower than 0.01 inch.
In calculating the vertical shear capacities, the openings in the wall panels have been taken into account.
The capacity reduction factor, 4,
of 0.85 and 0.9 have been applied to the capacity corresponding to the shear-friction of horizontal reinforcing steel, and that due to the restraining moment, respectively.
The full capacity of the frictional component of beam-column connection has been taken since the following conservative assumptions were made in calculating the capacities:
(a)
The clamping force due to the bolt tension equal to the proof load of the bolt was used in the capacity determination.
In the beam-column connections of the Complex walls, the Turn-of-Nut (1/2 turn) method was used to provide the bolt tension and the resulting bolt load corresponding to 1/2 turn is approximately 20 to 30% higher than the proof load (Ref.1).
Studies have shown that any relaxa-tion that occured in the inelastically deformed bolts has no effect on bolt performance.
These studies indicated that only about 5 percent of the initial bolt tension would be lost over the life of a structure (Ref.
2).
DH-10
Page 3 of 8 (b)
A friction coefficient of 0.ab was assumed.
Results of tests indicate that the slip coefficient for tight mill scale surfaces ranged from 0.34 to 0.57, with a mean value of 0.46 (Ref. 3).
It may also be noted that in the design of steel structures by using the principle of plastic design method, in accordance with AISC, Part 2, no capacity reduction factor is applied.
Although Licensee considers it inappropriate to apply a capacity reduction to the frictional component of the beam-column connection, the following representative examples illustrate the effect of application of a factor of 0.9 to this component:
Vertical Shear l
Total Capacity (kips) l l Wall l Column LinelElevationl100% Frictionall90% Frictionall l
l l
l Component i
Component l
l N l
46 l
117' l
240 l
220 l
l l
l 105' l
600 l
570 l
l l
l 93' l
1170 l
1120 l
l l
l 77' l
1310 l
1250 l
l l
l 61' i
1610 l
1520 l
i l
l 45' l
3510 l
3400 l
DH-10
Page 4 of 8 Horizontal Shear l
Total Capacity (kips) l lWalllElevationl100% Frictionall90% Frictional l l
1 i
Component l
Component l
l N l_
105' l
4710 l
4560 l
l l
93' l
4650 1
4500 l
l l
77' l
4670 1
4520 l
l l
61' l
4780 l
4620 l
l l
45' I
4890 l
4760 l
The calculated vertical shear capacities are greater than the OBE vertical shear forces at all locations, except at eleva-tion 93' along Column line 46 on Wall R and down to elevation 61' along Column line 46 on Wall N.
In those local areas a small amount of tension will develop to mobilize the excess shear resistance from below.
In both the cases the total vertical capacity along the entire vertical plane is more than the force developed due to the factored OBE.
Thus, for both OBE and SSE loading conditions there exist adequate capacities to transfer the vertical shear forces along all the intermediate vertical planes in these walls.
The verti-cal shear capacity considering the ultimate vertical resis-tance of the beam-column connections would increase approxi-mately 30% over those presented here.
DH-10
Page 5 of 8 Horizontal Shear Capacity Based on the Single Curvature Model The horizontal shear capacities of the shear walls in the Com-plex provided in Section 3 of PGE-1020 we' e calculated on the basis of a double curvature model of individual wall panels and applying the equation established in Section 3.4.2.2 of the above document.
The contributing parameters to those calcu-lated capacities were the percentage of vertical reinforcing steel and the effective dead load on the wall panels.
An alternative method of calculating horizontal shear capaci-ties of the walls was presented in the reponse dated July 6, 1979 to NRC Staff Question 43, dated May 18, 1979.
It was also stated in that response that the capacity evaluation based on the double curvature model was conservative in that the restraining effect due to the vertical shears at the two side edges was not included.
The method described in that response was based on a single curvature model of a wall panel with restraining moments provided only at the bottom of the panel by the vertical reinforcing steel and the effec-tive panel dead load together with those provided by the vertical shears given by the beam-column connection and the horizontal reinforcing steel across the two vertical edges.
The horizontal shear capacities of the major shear walls of the Control Building are calculated by this method and are shown in Figures 5 through 8.
The horizontal shear forces for both unfactored and factored OBE conditions are also shown in the figures.
Wall capacities at and above elevation 93' have been calculated neglecting the effect of dead load.
DH-10
Page 6 of 8 For capacities at elevation 77',
the dead load at elevation 93' is considered.
The wall dead load is reduced by 13% to account for the effect of the vertical earthquake and another 20% to consider the possible effects of creep and shrinkage.
Also, the vertical shear resistance at the panel edges is restricted to the frictional component of the beam-column connections only and not the ultimate values of the connec-tions.
The horizontal shear capacities considering the ulti-mate vertical resistances will be increased by approximately 30% over those presented here.
It may be noted that the contribution of dead load in the overall shear capacity as calculated by.his method is sig-nificantly lower (6% or less) compared to its percentage in the capacity determined from the double curvature model.
The Shear Wall Testing Program described in Appendix A of PGE-1020 provided information on capacity and stiffness degradation of wall specimens similar to the type present in the Trojan Plant.
By varying the axial loads and the vertical steel reinforcing ratios, the influence of these parameters on the capacity and stiffness of the specimens was established.
The results from the tests also showed the capacities when the failure mode was flexural as well as shear.
Section 3.4.2.2 of PGE-1020 correlated the flexural capacities to the flexural equation, which then formed the basis for the capacity evaluation of the Complex walls.
DH-10
Page 7 of 8 The majority of wall specimens were tested in a double curva-ture mode with the point of inflection approximately at the mid-height ( see PGE-1020, Table Al-1).
This type of boundary restraint was selected because of the simplicity it offered in the test set up while indicating the specimen's behavior under flexural and shear modes of deformation when subjected to lateral.oadings.
The mechanism of shear resisting capability of the specimen was a function of the summation of restraining moments de-veloped at the top and bottom of the specimen in the double curvature model. The single curvature model essentially follows the same mechanism excepting that while the restrain-ing moment at the top of the wall panel is absent, additional flexural resistance is provided by the vertical restraints generated at the two side edges.
The similarity of perfor-mance characteristics of these twc types of models is exhibited by the test specimens L1 and L2.
These specimens, described in detail in Appendix A of PGE-1020, were tested in single curvature with embedded steel columns to provide the restraints at the vertical edges.
Horizontal Displacements Corresponding to the Single Curvature Model Based on an approximate analysis, the horizontal displacements of the Complex corresponding to the single curvature model have been determined to be approximately 30% greater than those based on the STARDYNE elastic analysis of the modified DH-10
Page 8 of 8 Complex.
Review of the relative displacements between the Complex and Turbine Building reported in Licensee's response to NRC Staff Question 20 of May 18, 1979, and the relative displacements between the Complex and Containment reported in Licensee's response to NRC Staff Question 44 of May 18, 1979, shows that the gap between the structures is adequate to accomodate the displacements of the Complex corresponding to the single curvature model.
DH-lO
References:
1.
" Calibration of A325 Bolts," by Rumpf, J.
L.
and Fisher, J. W.,
ASCE Structural Division Journal, December, 1963.
2.
"Long Bolted Joints," by Bendigo, R.
A.,
- Hansen, R.
M.,
and Rumpf, J.
L.,
ASCE Structural Division Journal, December, 1963.
3.
"High Strength Bolting for Structural Joints," by Bethlehem Steel, Booklet 2867.
DH-10
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--- Force (1. 4 OBE) and Capacities
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Figure 6 Fbri:cntal Shear Forces and Capacities
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A.
Schwencer February 13, 1980 THE LOAD AND RESISTANCE PATHS BETWEEN TPE CONCRETE CORE A'ND THE BLOCK OF COMPOSITE WALLS The vertical shear trans fer of the composite block walls in the Complex at column locations is achieved by the bond be-tween the column and the core concrete, the frictional resist-ance of the beam-colunn connections and by the shear friction resistance of the continuous horizontal reinforcing bars in the continuous block.
Only the latter two mechanisms are used in the analysis.
The trans fer of shear by shear friction in the blocks is achieved by the direct trans fer of the in-plane shear stresses across the potential vertical slip plane at the column face.
Thus, there is no through-wall-thickness shear trans fer relied upon between the concrete core and the block.
The transfer of shear through the frictional resistance of the beam-column connections involves, first, the transfer of this frictional resistance by local bearing of the beam lower flange on a partial thickness of the wall, and secondly, the spreading of this local bearina stress to a lower magnitude, relatively uniform compressive stress over the total thickness of the wall at some distance away from the local bearing area, as shown in Figure 1.
The spreading of the local bearing stress from partial thickness to the total thickness of the wall induces interface shear stresses between the concrete core and the block.
For the case when the beam flange is wider than the concrete core thickness, such that the flange bears directly on both the concrete core and part of the block, the shear stresses DL-9
Page 2 induced at the interface of the core and the block will not be significant.
For larger beam sizes, even though the frictional resistance of the beam column connection is bigger, the larger shear and bending stiffnesses associated with the larger beam sizes tend to spread the local bearing stress over the wall to a larger distance.
Thus the peak load bearing stress may not be critical.
The peak bearing stress of a beam on a wall is most critical for the smaller beam sizes.
However, the slip resistance by the beam column connection is also smaller for the smaller beams.
As far as the interface shear stress is concerned, the critical case is judged to be the thinner walls where the width of bean flange is approximately equal to the core thickness.such that the beam flange bears directly on the concrete core alone with only a small distance from the outer edge of the beam flange to the core-to-block interface.
For illustrative purposes, the load and resistance trans fer mechanism between the concrete core and the block of the composite wall panel in Wall 55, at column line R, at el.
105' is presented.
This wall panel has a total thickness of 28" with a 12" concrete core.
The beam at el. 105' is W 27 x 94 having a flange width of 10".
The frictional resistance of the beam-colunn connection at this elevation is 190 kips.
This wall panel is judged to be one of the most critical panels for interf ace shear transfer.
DL-9
Page 3 (1) Compressive Bearina Stress of the Beam on The Wall The maximum bearing st ess on the concrete core cor-r responding to the frictional resistance of 190 kips is 4,650 psi (see Table 6-2 of Licensee's response to the NRC Staff Question 6 of 10/2/79).
This maximum bearing stress was calculated using the method of an elastic beam on a linear elastic foundation including the shear deformation of the steel beam.
The stif fnesses of the concrete encase-ment of the beam was conservatively ignored. The inclusion of the stif fness contribution would lower the maximum bearing stress as calculated.
The beam hearing stress distribution can be idealized by an equivalent triangular distribution as shown in Figure 2.
The distance of local bearing is dete rmined to be 8.2".
(2) Core-to-Block Interface Shear Stress Distribution When the concrete core is subjected to a uniform compressive bearing stress over a 10" core width as shown in Figure 3, the shear and compressive stresses along the core-to-block interface boundary, which is at a distance 1" away from the outer edge of the beam flange, can be computed using the method presented in Section 24, " Solution of the Two Dimen-sional Problems in the Form of a Fourier Series", chapter 1 of " Theory of Elasticity" by Timoshenko and Goodier. The resulting shear stress (t) and the compressive stress (ox) can be plotted as shown in Figure 3.
DL-9
Page 4 The net interface shear stress computed as the difference between the shear stress T and the compressive stress e
i.e.
ITI-o will be critical in determining x,
x, the capacity of the interface shear transfer between the concrete core and the block.
The net shear stress, (T-ex), is plotted in Figure 4.
As shown in that figure, the maximum net shear stress is approximately 0.2q, where q is the magnitude of the bearing stress.
Using a maximum shear stress of 150 psi between the con-crete core and the block, the maximum hearing stress qmax above which the net interface shear stress would he greater than 150 psi, can be calculated as follows:
150 750 psi q
=
=
max 0.2 (3) Reaion where Net Shear Stress Is Greater than 150 psi Assuming a triancular stress distribution and a 45* angle spreading of the local bearino stress in the concrete core as shown in Figure 5, the compressive stress at any elevation below the beam can be computed as follows:
1/2 g
= 1/2 a (y + d) g d
q=
q y+d o
DL-9
Page 5 For example, at a distance of 2 ' below the bottom of the beam, the compressive bearing q can be determined to be 8.2 (4650) 1184 psi
=
a
=
~
24 + 8.2 For this triangular compressiva stress distribution, the horizontal distance x, at wh ic': the compressive stress equal to the qmax of 750 psi can be determined as follows:
9 max y+d - x
=
q y+d For y =
2',
q= 1184 nsi, x is determined to be 750 24.0 + 8.2 - x
=
- x = 11.8" 1184 24 + 8.2 Following this procedure, the reaion where the net shear stress is greater than the maximum shear stress of 150 psi can be determined and plotted as shown in Figure C.
Figure 6 shows that this zone is very local, and there is ample reserve capacity in the wall to trans fer the 190 kips beam-column frictional force.
The above calculation is quite conservative.
In the actual case, the beam bearing stress would be smaller than that calculated previously since the beam is encased DL-9
Page 6 in concrete which would transfer part of the load to the face block directly, and the concrete material is non-linear, thus having the ability of stress redistribution.
Even though the elastic solution predicts shear stresses greater than 150 psi over a small local zone, it is the Licensee's judgement that the load trans fer capability will not be red uc ed.
DL-9
/
/
/
/
]
/
/
CORE I
i
- BLOCK 6
i i
t l
(!
oi
\\
ba!
s f
Y7111 tir
~ ~ _ _ __
o lg h j' 5 *
\\
4 3
=
=
x f
N x
N
/
\\e B
B, B__
x*
N j/
y+d E
B BI 8
fl A a a m
o a
a o 6 I
+
i e
i i
o FIGURE 1
6 L lo"_'
D l
a g,
$,, i s i
o l
0;Q ti1hh'i N
Yffk7 t
d = 8.2
\\
g.
,g.
g ;
o
=
FIGUPE 2
10"_
q C' G*
A tttt"t O. 23 g - '
-' 0.067cr N
'7 N
o.123g-v
-c.025 g L
llD'd'x
[ ^-
A
~
~-
B" l
/2" 8"
I l
l A
+ 6x COMPRES5/ON FIGURE 3 I!ITERFACE SHEAR A!!D COMPRESSIVE STRESSES ALO'!G AA
mi - L, A.
A o
1 1 o I
)
A
/
g3' f
'I
~1 :+ :
C.197 g O
.wei y
N
\\
l A.,..
A
_ s" 12" _ _6_
FIGURE 4 NET INTERFACE SHEAR STRESS ALONG AA
L
\\
b
\\
N r
, y
@o o
a x
=
3
'N N
(750 PSI
\\
a 3
x i
'N X
S+d.-X-s
\\
4+d
=
FIGURE 5 SPREADING OF REARING STRegg
6,7' EL 105 -53 j
/'-
p,.
g
[:,,'g,,
~Y ZoME IM WHICH THE NET
,7
- /;'d -
6HEA2 6TZE% EXCEED 5 ISO RS/
'l Y'-
v IG.1" 6"
"I i
i i
l i,
l 1
r a
E '. 95 - o J FIGURE 6
A.
Schwencer February 13, 1980 Page 1 of 4 DEAD LOAD REDUCTIONS DUE TO SHRINKAGE The amount of shrinkage strain in the composite walls of the Complex was discussed in Licencee's response dated December 22, 1979 to NRC Staff question 2 dated October, 2,
1979.
The restrained shrinkage stcain was established in that response as 84 x 10-6 inch / rich.
This value is considered to be appropriate.
If, however, a much higher value of 140 10-6 inch / inch is assumed generally in the walls, its x
effect on redistribution of wall dead load to the encased structural steel columns can be assessed as follows:
A
= Displacement of a wall panel unrestrained by the columns g
140 x 10-6 x 14 x 12
=
= 0.0235 inch, for a panel with a 14' effective height A
= Final restrained displacement, inch 4
I
= Moment of Inertia of floor beam, inches E = Modulus of elasticity of steel, ksi
= 30000 ksi 2
A = Cross-sectional area of steel column, inches 1 = length of column 192 inches
=
x = Length of separation of the beam, inch (See " Strength of Materials", II by Timoshenko, Chapter III, example problem 6) q = Uniformly distributed load on beam, kips /ft A
- P1 a
=
o AE X9 p=
2 DL-1
Page 2 of 4 4
0.0235 X"1 UX or,
=
24EI 2AE Por a W 24 x 68 typical beam and W 14 x 142 column, 2
1820 inches 4, A = 41.8/2 = 20.9 inches I
=
(2.29 x 10-5q) x4+
(4.59q) x -705 = 0 (1)
Case 1 Dead load = 100 psi Span
= 19.25 ft.
Thickness = 26 inches 100 x 26 q=
= 2.6 kips / inch 1000 Equation 1:
(5.95 x 10-5) 4+
(11.93) x - 705 = 0 x
or, x = 41 inches P=
2
= 53.3 kips Total load, W = 2.6 x 12 x 19.25/2
= 300 kips Hence, load trans ferred to column = 18%
DL-1
Page 3 of 4 Case 2 Dead load = 50 psi 19.25 ft Span
=
Thickness = 26 inches 50 x 26 q,
1000 1.3 kips / inch
=
Equation 1 :
(2.98 x 10-f) 4 x
+ (5.97) x-705 = 0 or, x = 60 inches 60 x 1.3 p,
2
= 39 kips W= 1.3 x 12 x 19.25/2
= 150 kips Hence, load trans ferred to column = 26%
Case 3 Dead load = 10 psi Span
= 31 ft.
Thickness = 35 inches x 35 q, 10 1000
= 0.35 kip / inch DL-1
Page 4 of 4 Equation 1:
(0.8 x 10-5) 4+ (1.6) x - 705 = 0 x
or x = 90 inches 90 x 0.35 p,
2
= 16 kips W = 0.35 x 12 x 31/2
= 65 kips Hence, load trans fe rred to column = 2 5%
Let us there fore, assume that a total dead load reduction due to the ccmbined ef fect of creep and shrinkage, for a total shrinkage strain of wall of 140 x 10-6 inch / inch, is 30%.
With this conservative assumption, the reduction in wall dead load and the consequent reduction in the wall shear capacity based on double curvature, the wall capacity will still be more than the demand load.
DL-1
A.
Schwencer February 13, 1980 CAPACITY-TO-FORCE RATIOS FOR NEW STRUCTURAL ELEMENTS Page 1 of 3 The following discussion provides capacity-to-force ratios for the new structural elements:
- 1. Concrete The capacity of the new concrete structural elements has been determined in terms of:
a) shear capacity b) flexural capacity In these capacity determinations, the new concrete walls were assumed to be isolated from the adjacent strucures and they were assumed to resist the applied loads without the assistance of these adjacent structures.
All loading combina-tions specified in the Trojan FSAR were considered in this evaluation, but only the capacity to force ratios of the governing loading combination are listed in this response.
The method used to determine the capacities is in accordance with ACI 318-77.
When calculating shear capacities, in the determination of v
reductions of axial forces (N) due to seismic forces c,
DL-2
Page 2 of 3 and shrinkage were considered.
Dead load was conservatively neglected.
In calculating V, the effective steel area s
used in the capacity determination was found by deducting from the total area of reinforcement provided, the area of reinforcement required for out of plane inertia forces, inter-story drift, and thermal gradients.
The calculations of flexural capacities were based on the combined effect of axial forces and moment.
The effective area of reinforcement was also obtained by reducing from the area of reinforcement provided, the area used for out of plane inertia forces, interstory drift and thermal gradients.
The capacity to force ratios for the new concrete structural elements are shown in the table below.
l l
l Factored OBE l
l l
l Capacity-to-force ratio l l Wall l
Elevations Ifor shear l for flexure l I
I I
I I
l N
l 45' to 65' l
1.35 l
1.64 l
l l
65' to 77' i
1.74 l
1.32 l
l l
77' to 93' l
1.77 l
2.64 l
1 1
I I
I l
R l
45' to 65' l
1.60 l
3.39 l
l l
65' to 77' l
1.72 l
6.06 l
1 1
I I
I l
N' l
45' to 65' l
3.50 l
1.67 l
1 l
DL-2
Page 3 of 3 used in the capacity determination was found by deducting The capacity of the 3 in. thick plate is governed by the welds.
It was determined that the minimum capacity to force ratio for the welds is 1.6.
The allowable stress in the welds was based on AISC Specification without an increase for seismic loads.
- 3. Bolts Although all the bolts are not loaded to their design load, for this evaluation it was assumed that they are.
Then the capacity to force ratio for each bolt would be 2, based on the factor of safety used in the formula uLF given in y_
FS Section 3.2.4.2 of-PGE 1020.
DL-2
A.
Schwencer February 13, 1980 Page 1 of 2 SLIDING RESISTANCE OF WALL PANELS Response to NRC Staff Question 11 of May 18, 1979 stated that the ultimate shear resistance to sliding at the wall-slab inter-face consists of the column resistance and the resistance re-sulting from the shear friction due to the vertical steel.
The shear friction resistance V was taken as =
V
= p(A fs y + N)
- where, u
= apparent coefficient of friction
= 1.4 area of vertical reinforcing steel A
=
s yield stress of rebar = 40 ksi f
=
y direct dead load on wall reduced for the effect N
=
of vertical earthquake (kips)
The applicability of this equation was confirmed by the test specimen L2, described in Appendix A of PGE-1020.
However, in order to provide conservatism to the above criteria, the sliding resistance of the west wall along column line R of the Control Building was reexamined with the shear friction resistance V taken as =
V = pA fsy+U The results summarized in Table 1 show that sufficient capacity remains available.
DH-9
Page 2 of 2 TABLE 1 SLIDING RESISTAtJCE AND SilEAR FORCES Ill KIPS WALL ALONG COLUMN LIllE R OBE Resistance against sliding-ultimate OBE Resistance Shear Forces Factor of Elevation Safety Shear Steel Total
=V
= 0.85V y
I I'4 Friction Columns
- 45' 5690 2398 8088 4910 2140 2.29 61' 3393 2398 5791 3516 2650 1.33 77' 3363 2028 5391 3273 2480 1.32 93' 4107 2028 6135 3725 2260 1.65
- Revised-from previous response only to reflect review of calculations.
Dil-9
A.
Schwencer February 13, 1980 TEMPORARY LOADINGS ON WALLS DURING MODIFICATION PROGRAM The maximum collar joint mortar shear stresses in masonry walls due to the out-of-plane loading combination of thermal loads, fluid concrete pressure, SSE interstory displacements and SSE inertia loads including those contributed by the fluid concrete, unsecured steel plates, rebar and forms are less than 18 psi.
DH-7
A.
Schwencer February 13, 1980 Page 1 of 2 RESISTANCE TO OVERTURNING AND SLIDING OF THE MODIFIED COMPLEX The following discussion describes our evaluation of the resis-tance of the Control Building to overturning and sliding.
Cri-teria used to verify resistance of the Complex to overturning and sliding are in accordance with the NRC's Standard Review Plan Section 3.8.5 of that document specifies a minimum factor of safety of 1.5 for an OBE and a factor of safety of 1.1 for an SSE be provided to resist overturning and sliding.
(See Licensee's response dated October 10, 1978 to NRC Staff Question 24.)
The method used for calculating structural overturning is as git,n in BC-TOP-4A, Rev 3, Section 4.4.1.
The method used in calculating resistance against sliding is as described in PG E-10 2 0, Section 3.2.4.4.*
1.
Overturning of Control Building The factor of safety is:
E_2 = 1330 Es
- where, E
= Energy required to overturn the structure g
E
= Maximum kinetic energy applied to th: Control Building s
- Areas that may be subject to uplift are excluded in evaluating the contribution from cohesion, C.
DH-8
Page 2 of 2 2.
Sliding The results of our evaluation of sliding resistance due to an OBE in the north-south direction ** are as follows:
Grade Beam Column to to Rock Shear Footing Shear Total Shear Base Factor Resistance Resistance ***
Resistance Shear of Wall (k)
(k)
(k)
(k)
Safety R
5,880 550 6,430 4,230 1.52 O
1,090 110 1,200 445 2.70 N'
2,130 170 2,300 720 3.19 N
5,500 1,000 6,500 3,040 2.14 H to L 7,790 7,790 3,220 2.42 Total 22,390 1,830 24,220 11,665 2.08 3.
Cenclusions Since the SSE loads are only approximately 2 porcent greater than the OBE loads, approximately the same factors of safety apply to SSE loads.
The above results demonstrate that the Modified Complex has sufficient capacity te resist sliding and overturning loads resulting form the 0.; 3g SSE and 0.15g OBE.
- Values for the east-west direction are not given because the factors of safety are higher.
- A factor of safety of 2 was used in conjunction with the coefficient of friction of 0.7 between the steel column base plates and the concrete foundation in order to provide addi-tional conservatism.
DH-8
A.
Schwencer February 13, 1980 POTENTIAL PREQUENCY SHIFT DUE TO GROSS RENDING AND VERTICAL g EAR TRANSFER MECHANISMS Gross Rendina Effects on Structural Frecuencies The influence of gross bendino on the natural frequency is normally neglected in conventional structural analysis and design.
It is dif ficult to evaluate this influence usina an ele, tic analytical model of the structure.
As described in Appendix B of PGE-1020, the stiffness of wall elements used in the finite element model of the Complex is a function of reinforcing steel ratio, horizontal unit shear stress, and vertical load.
The contribution of the vertical load to element stiffness and stiffness reduction factors varies considerably throughout the Ccmplex.
With gross bending considered to induce an effective vertical load on a wall element, the total vertical load on the eierent would consist of the passive dead load, the time dependent vertical load due to the vertical earthquake component and the time dependent gross bendir.g resulting from the horizontal earthquake component of the structure.
In terms of stif fness, these vertical load ef fects would be coupled with the effects of the time dependent horizontal shear forces acting on the wall eitmant.
To describe the overall behavior of stiffness changes in the structure due to the seismic direct vert ical load, gross bending vertical load, and horizontal shear load on individual wall panels, these effects can first be thought of in a pseudo-static sense.
Where the vertical load components produce an increase in compression on a wall element in a DL-9
Page 2 particular cycle of seismic loading, the wall panel would also experience some amount of stiff.iess increase.
Simi-larly, wall panels that are undergoing a decrease in vertical load woild experience the attendant decrease in stif fness.
With the change in stif fness due to changes in vertical load, the horizontal seismic shear forces acting on the structure would tend to be redistributed with increased shear on the compression side of the structure and a proportional decrease on the tension side.
Powever, since wall panel stiffness is also a function of unit horizontal shear stress, the shear redistributed to wall elements an the compression side would produce a decrease in stif fness in these elements.
Thus, the ef fects of cross bending on wall panel stiffness would tend to be self ccanonsatino to some extent.
When the structure is viewed more realistically in terms of ex pe ct ed dynamic per fo rmance, other considerations enter into the evaluation.
In order to provide a realistic assessment of the stif fness of the wall panels during a seismic event, the time dependent vertical load contribution to stiffness would be coupled, by statistical or other appropriate means, with the time dependent horizontal load contribution.
As nonlinear redistribution of shear forces takes place in the structure, energy is being absorbed resulting in increased damping and stif fness is experiencing an instantaneous re-duction.
The increase in damping would in turn reduce the inertia loads acting on the structure below those predicted by the linear elastic finite element analysis.
This nonlinear DL-8
Page 3 type of response generally results in a reduction in the am-p11tude and broadens the peaks on the floor response spectra and results in some shift in frequency.
The shift is smaller than predicted by comparing the instantaneous and initial stiffness.
The influence of all of the above factors on the potential shift in the natural frequencies of the Complex due to gross bending considerations cannot be realistically evaluated without the use of inelastic techniques beyond the normal state-of-the-art in building analysis and design.
However, a simplified analysis which icnores many of the factors which would tend to limit the effect of gross bending on frequencies can be pe r fo rmed.
Licensee has pe r fo rmed such a simplified analysis of the R and N Wall model described in the portion of this submittal titled " Effects of Gross Bending on Shear Wall In-Plane Capacities", and a conservative calcul-, ion of the bounding frequency shifts due to gross bending sas obtained.
The potential shi f t in the fundamental frequency of the Complex due to the gross bending ef fects was estimated by comparing the stiffness of the lower sections of the N and R walls with those obtained previously (see response to NRC Staff Question 23 of October 2, 1979).
In obtaining this new set of stif fnesses, the direct dead load was reduced for vertical carthquake and the vertical forces due to gross bend-ing were obtained from the finite element analysis.
Using the DL-8
Pace 4 stiffness reduction factors given in PGE-1020, Appendix B, a
reduction in stiffness of 18% was calculated, which corresponds to approximately a 9.5% reduction in frequency.
To obtain the frequency shift for the entire structure, the reduced stiffness of each wall would need to be calculated and then it would be necessary to average the values for all of the walls by some weighting technique.
The stiffness reduction due to gross bending is expected to be less at the higher elevations since the gross bending is a maximum at the base and reduces with height.
The cross bending effect in the east-west direction is less than for the north-south since the lengths of the walls in the east-west direction are greater.
Therefore, a conservative estimat.e of the reduction in stiffness of the entire structure due to cross bending would be 18%.
This approach neglects the influence of the reinforced concrete structures in the Fuel Duilding whose stiffnesses will re-duce less than the composite walls.
A corresponding predicted bounding shift in fundamental frequency of the Complex due to gross bending ef fects would be 9.5% in both the M-S and E-W directions.
Vertical Shear Transfer Effects on Structure Frecuencies As is the case for gross bending ef fects, realistic evaluation of the potential frequency shift due to moblization of vertical shear transfer mechanisms would involve nonlinear effects which cannot be accounted for in a simplified structural model.
DL-8
Page 5 Again, however, a simplified analysis was used to predict a
very conservative bounding frequency shift for vertical shear trans fer.
For example, concrete was assumed to have nc shear capacity and bond was totally neglected at the embedded steel frame to concrete core interfaces.
The vertical shear resistance between adjacent panels was obtained from the shear friction of the horizontal reinforcing steel and the friction resistance of the bean-colunn connection.
Associated with this resistance model is sone relative movement between ad-jacent panels.
If the full shear friction resistance is needed to meet the demand of the OBE, the relative movement could be 0.01 inch.
Since the entire resistance is not required, the actual relative movement will be less.
If this relative vertical movement were to occur, it would result in an in-crease in displacement which could have the effect of decreas-ing the fr ec uency.
The increase in the horizontal displace-ment was estimated by considering a panel to rotate about one corner to generate a vertical movement at the other edge of 0.01 inch.
Using the smallest capacity to demand ratio for the vertical shear, the shift in the frequency would be 7.9%.
If an aver-age ratio was used, a snaller reduction would be calculated.
In the east-west direction, the same approach was used.
How-ever, the Control Building is not the dominate element for de-termining the frequency.
The stif fness of the Spent Fuel Pool and Hold-up Tank Enclosure Structure has a shear area in the east-west direction slightly greater than the other resisting DL-8
Page 6 elements.
Using this along with the smallest capacity to demand ratio the calculated frequency decrease would be 7.8%.
SUMMARY
The calculated bounding potential shifts in fundanental fre-quencies of the complex due to gross bending (9.5% N-S and E-W), and due to vertical shear transfer (7.9% N-S and 7.8%
E-W) were based on simplified analytical models which do not account for inelastic performance of the structure and other factors which have been neglected.
If such considerations were taken into account, it would be expected that the fre-quency shifts would be less than those calculated by simpli-fied analysis methods.
The total calculated frequency shift due to these reductions would be considered to occur simultaneously, which results in a calculated frequency shift of 16.6% (0.905 x 0.921 =
0.834).
In the Complex there is earthquake monitorino equip-ment from which the natural frequency of the Complex can be determined under in-situ conditions in the event of a measur-able earthquake.
These experimentally determined frequencies can be compared with those obtained from any subsequent earth-quakes to determine if any shift in frequencies has occurred.
Based on the results of this comparison, any appropriate ac-tion can be taken.
Because of the information which can be obtained from the existing monitoring system and since the frequency shif ts described above are considered unlikely to occur, it is unnecessary to provide curve broadening beyond the 20% described in Licensee's response to Question No. 21, dated October 2, 1979.
DL-8
A.
Schwencer February 13, 1980 Atti.chment 10 NRC Ouestions (9/14/79)
O. 6 Page 1 of 17 Provide the results of your analyses showing that plates 1 through 6 are sufficient to sustain without detrimental ef-fects on plates 1-6, the structure, equipment, piping, or cable trays, the impact of plate 8 should a drop of plate 8 occur.
Include (a) a detailed description of all assumptions used in the analyses, and (b) detailed justification for all of the assumptions used in the analyses, all of the loads and all of the acceptance criteria relied upon.
Include an identical discussion for plate 7.
Answer:
In order to preclude any possibility of detrimental effects on Plates 1-7, the structure, equipment, piping and cable trays should a drop of Plate 8 occur, the maximum drop height of Plate 8 will be limited to approximately 4 inches by placing timber cribbing on top of Plates 5, 6, and 7 as shown on the attached Figures 6-1 and 6-2.
The timber cribbing will con-sist of two piles of 4" x 4"*x 4' long pieces stacked on top of each other, except that the last piece removed in each pile will be 2 in. thick.
A 1 in. thick, 4 in, wide, and 4 ft long HEXCEL pad, stabilized and procrushed to give a crushing strength of 750 psi, will be placed under each pile.
As the plate is being lowered, the 4 in. thick timber segments will be removed one at a time from each pile, thus limiting
- These are actual physical dimensions, rather than nominal dimensions.
NPC Ouestions (9/14/79)
Q. 6 Page 2 of 17 i
the maximum possible drop height of Plate 8 on wood to approx-i imately 4 in.
The HEXCEL pad will be removed last, thus fur-f ther reducing the drop height of Plate 8 on the lower plates to approximately 1 in.
i
[
The timber cribbing will consist of Douglas Fir or similar wood l
having a cross grain (perpendicular to grain) strength of about f
800 psi. It will be supported on the bottom by brackets attached to the lower plates.
The cribbing will be braced laterally by f
guide plates designed to prevent bulging and subsequent collapse
[.
of the cribbing.
The guide plates will be supported by the Tur-bine Building floor at el.
93',
by the floor girder and by the lower plates on the bottom.
Temporary lateral bracing will be added to the floor girder in order to resist the lateral forces induced by the bulging of wood piles due to a postulated drop of Plate 8.
Effects of 4 in. Drop of Plate 8 1.
Energy Absorption The kinetic energy associated with the 4 ir.
drop of Plate 8 will be absorbed by the crushing of either the wood or the HEXCEL pad, or the combination of the two, since the strengths of the timber and the HEXCEL pad are about the same.
If the timber strength is higher than the crushing strength of the HEXCEL pad, the HEXCEL pad would crush first and the excess energy would be absorbed by the wood.
If the last 2 in, wood
l NRC Ouestions (9/14/79)
O. 6 Page 3 of 17 segment has been removed and only the HEXCEL pad is Icft, then the kinetic energy of a drop of Plate 8 would have to be absorb-ed totally by the HEXCEL pad.
The kinetic energy resulting from a 4 in. drop of Plate 8 is:
k x4n
= 200 k-in.
i KE = 50 If this kinetic energy is to be absorbed by the HEXCEL pad alone, KE = fcr*AX 3 where f
= crushing strength of HEXCEL pad = 750 psi = 0.75 ksi cr A
= bearing area = 4" x 48" x 2 = 384 in.2 S
= depth of crushing required Substituting the values give 200 = 0.75 x 384 x S; S = 0.69 in.
Ther e fore, the required thickness to absorb the total energy is 0.69 in/0.7 = 1.0 in., where the factor of 0.7 is the ratio of total thickness to the crush depth recommended by the HEXCEL manufacturer *.
Hence, a 1 in, thick HEXCEL pad with
- Calculations for other uses of HEXCEL have conservatively assumed only 50% of the total thickness to be available for crushing.
However, that assumption has not been made in this instance since the energy from a drop of approximately 4 in-ches would be absorbed by both the HEXCEL and at least 2 in-ches of wood.
The maximum drop onto the HEXCEL alone would be approximately 2 inches.
i f
i' NRC Ouestions (9/14/79)
O.
6 Page 4 of 17 750 psi crushing strength is adequate for absorbing the total kinetic energy associated with a 4 in. drop of Plate 8.
The last timber segment removed from each pile will be 2 in.
thick, and the remaining bearing area will be 3.5 in. x 48 in x 2 = 336 in2 The drop height would thus be approximately 2 in.
Following the same approach as above, the required thick-ness of the HEXCEL pad is 0.56 in.,
which is less than the re-commended depth.
2.
Ef fect on Lower Plates The maximum vertical force, F,
acting on the lower plates induced by a 4 in. drop of Plate 8 is limited by the maximum crushing strength of the HEXCEL pad, i.e.
F= 1.3 xf x A = 1.3 x 0.75 x 384 = 374.4 kips cr where crushing strength of HEXCEL pad = 750 psi f
=
cr 2
A
= bearing area = 2 x 4 x 12 x 4 = 384 in
/fg and the factor 1.3 is the dynamic increase factor for the HEXCEL pad as recommended by the manufacturer.
This vertical force acts on the lower plates almost instantaneously and main-tains over a time duration td which is approximately equal to the time required to bring plate 8 to stop.
This duration can be calculated as follows :
NRC Ouestions (9/14/79)
O.
6 Page 5 of 17 V
= V2gh
=Y2x32.2x 4
= 4.63 ft/sec o
12 I
4.63x50
= 0.019 sec t
=
=
d
( F/m) 374.4x32.2 The frequency of the lower plates supported by the bolts on the wall can be estimated as follows:
l I
Vertical s t i f fn es s,
K, based on the group of 84 bolts, is deter-mined from the load-displacement characteristics of the bolt as-sembly.
Since the majority of the bolts are located in concrete, K is calculated on that basis.
214K 5
K=
x 12 in./ft x 84 bolts = 1.66 x 10 K/ft 1.3 in.
_/ f__
2 Mass of lower plates = M = 100/32.2 K-sec / ft
=
Ef E = 36.8Hz 1
Vertical frequency of lower plates on bolts = f 2H M
p Vertical f undan.2ntal frequency of R Wall = f
= 28.0 H y
z 1
2 1
1 Vertical frequency of plates on wall = f=
+
= 22.3Hz; f
f*2 2
- P Vertical period of plates on wall = T = 1/ f = 0.045 sec.
NRC Ouestions (9/14/79)
[
I I
O. 6 Page 6 of 17 i
t
- he dynamic load amplification for a rectangular pulse having a i
duration of td = 0.019 sec. acting on a dynamic system having a l
{
period of T = 0.045 sec. can be found to be about 1.9.
- However, for conservative purposes, a maximum amplification factor of 2 will be used, since the period T cannot be determined accurately.
Thus, the maximum vertical force acting on the wall due to the 8
2x374.4 =
4" drop of plate 8 on the HEXCEL pad is F
=
max I
748.8 kips.
Adding the initial dead load of the lower plates, the total force is 748.8 + 100 = 848.8 kips.
Referring to Figure 1-3 of Licensee's Res ponse to Question 1, Systems Branch, i
dated 8/17/79, the maximum vertical force of 848.8 kips acting on the wall is resisted partly by the vertical resistance of I
the bolts and partly by the frictional resistance developed at j
the interface of the wall and the lower plates.
It can be seen from Figure 1-3 of the referenced response that the ratio of the resistance by bolts to the resistance by friction is tan 0 to tan T.
It was determined that 0 is 15.24*
/
when the bolt is below the elastic limit, tan 0 = 0.27; and tan y = 0.7 is the friction coef ficient.
Thus, the maximum vertical force on bolts is tan 0 0.27 848.8 x
= 848.8 x
= 236 kips, i.e.
2.8 tan y + tan 0 0.7 + 0.27 kips / bolt The maximum vertical friction force on wall is 0.7 848.8 x tan y
= 848.8 x
= 612. 5 kips tan y + tan 0 0.7 + 0.27 i
NRC Questions (9/14/79)
O. 6 Page 7 of 17 Corresponding to the maximum vertical force of 2.8 kips per bolt, i
f the maximum tension, T, induced in a bolt is 2.8 2.8
10.8 kips T
sin 0
.26 The allowable working tension per bolt without any increase is 129.9 kips.
The allowable shear capacity for the bolts in masonry and concrete, as established from Tables 24-G and 26-G of the 1976 UBC and extrapolating to 1-3/4 in. diameter, are as follows:
k Concrete wall:
7.7 / bolt (with special inspection) k Masonry wall:
3.8 / bolt A
Due to the dynamic nature of the applied loads, the allowable
/[_
values can be increased by 1.33.
The extrapolation was performed in the following manner:
- 1. The allowable shear capacity for bolts with special inspec-tion in concrete given in Table No. 26-G were plotted, and a curve having the following equation was obtained:
VB = 4. 8 ( D.15 ) ki ps
- 2. The equation gives the following allowable sheat capacity for a 1-3/4 in. diameter bolt:
VB = 4. 8 (1. 7 5.15 )
7.7 kips
=
3.
Comparison of the allowable shear capacities for bolts in 3000 psi concrete (Table No. 26-G) and allowable shear
t NRC Ouestions (9/14/79) l Q.
6 Page 8 of 17 e
f capacities for bolts in grouted masonry (Table 24-G) shows that for larger bolt diameters (1-1/8 in.), the allowable s
j shear capacity in grouted masonry is equal to one half of the allowable shear capacity in concrete with special in-l spection. Consequently the allowable shear capacity in masonry was established to be 7.7 = 3.8 kips.
r I
2
{
The force per bolt of 2.8 kips, as calculated previously, is less
{
than the allowable values for bolts in either concrete or masonry, e.
- 3. Effect on Wall A
l LL The wall on column line R has been investigated to detern ine its ability to resist the applied impact load of 848.8 kips.
Both bending and axial loads, including the dead load of the wall, were considered.
It was determined that the maximum tension in the reinforcing steel is 4200 psi, the maximum compression in tae block is 273 psi, and the maximum compres-sion in the concrete is 660 psi.
I 4.
Effect on Equipment, Piping or Cable Trays Since the kinetic energy from a drop of Plate 8 would be ab-sorbed by the wood and/or HEXCEL, there would not be any de-trimental ef fect on equipment, piping o
- ble trays.
l t
1 NRC Ouestions (9/14/79)
{
Q. 6 Page 9 of 17 i
5.
Lateral Stability of Wood Piles j
l 6
Due to some initial eccentricities, the wood piles subjected to the vertical force induced by a postulated drop of Plate 8 may bulge, thereby exerting a lateral force on the lateral support system of the wood pile.
The lateral support system, which consists of vertical guide plates laterally braced by the Turbine Building floor on the top, the bottom flange of floor girder at a'pproximately the mid-s pan, and the lower plates on the bottom, must have adequate stiffness and capa-city to resist this lateral force such that the amplitude of bulging is limited and the stability of the wood pile is main-tained.
In order to evaluate this stability, the following parameters are used (see Figures 6-3 and 6-4):
a.
A maximum 2' initial out-of-squareness, a0, of the 4" x 4" wood section; b.
42 in. wide and 1-5/8 in thick guide plates; a, between the wood pile c.
A maximum of 1 in. initial gap, t
and the guide plate; d.
A maximum of 1/2 in. lateral deflection, d, of the guide plate at the point of lateral load.
This deflection will be checked later after the maximum lateral force is de te rmin ed.
NRC Ouestions (9/14/79) l Q.
6.
Page 10 of 17 i
e.
A maximum of 2 in. vertical load eccentricity, e,
on the j
I wood pile.
This is very conservative considering that the '
I thickness of the wood is only 4 in.
I i
i Using the maximum initial out-of-squareness impe r fection, and assuming that these imperfections are aligned in the worst possible configuration, the minimum radius of curvature, r,
of the initial bulging can be determined by 4"
r
=
4" 114.6 in.
2*Xw/180*
a0 The maximum available gap is; A = ai + d = 1.0" + 0.5" = 1.5 in.
Usinga and considering the geometry as shown in Figure 6-5, the worst possible bulging configuration involves a stack of 9 wood sections.
This is determined as follows:
r ( 1 -c os,0,)
,3 2
(114.6) (1-cos S) = 1.5; 6 = 18.56*
2 Thus, the number of wood sections within the bulge is 0 _ 18.56 = 9.28 say 9.
n-60 2
I r
6 NRC Ouestions (9/14/79) l Q.
6.
Page 11 of 17 I
l Based on the worst bulging configuration as determined above and using the maximum load eccentricy of 2",
the maximum lateral force R resulting from the vertical force F can be calculated by equilibrium as follows (see Figure 6-4):
374.4 x 1.5 + 2 = 72.8 kips for each pile A+*
l R
=
F
=
h/2 4 x 9/2 l
I t
The maximum deflection of the lateral support system, resulting from this maximum lateral force is determined to be 0.21 in.,
j which is smaller than the 0.5 in, assumed previously.
There-fore, the lateral support system has adequate stiffness to re-
+
sist this load and the wood pile stability will be maintained.
Effects of 1 in Drop of Plate 8 on Lower Plates t
1.
Ef fect on Lower Plates The maximum kinetic energy associated with a 1 in, drop of plate 8 is:
_j_
KE = 50 x 1/12 = 4.17 Kip ft.
Assuming this maximum kinetic energy is to be absorbed by the displacement of the bolts alone, the maximum vertical. displacement, X,
of the lower plates can be determined as follows:
.;itC Ouestions (9/14/79)
O. 6 Page 12 of 17 i
5 KE = 1/2 KX2; K = 1.66x10 K/ft 4.17 = 1/2x1.66x105(x2)
I X = 0.00709 ft e
l Thus, the maximum vertical force induced during the impact is:
l t
5 KX = 1.66x10 x 0.00709 = 1180 kips I'
Adding the initial daad weight of 100 kips, the maximum vertical I
force acting on the wall is 1180 + 100 = 1280 kips.
As shown previously, the maximum vertical force on bolts is I
I t
0.27 1280 x
= 356.3 kips or 4.2 kips / bolt 0.7 + 0.27 The allowable shear capacity for bolts in masonry, as shown pre-viously, is 3.8x1.33 = 5 kips.
i The maximum vertical friction force on wall is:
0.7 1280 x
= 923.7 kips 0.7 + 0.27 I
j The maximum tension, T, induced in a bolt by the vertical force t
j of 4.2 kips is:
i b
NRC Ouestions (9/14/79) l Question 6 Page 13 of 17 4.2 4.2 16.2 kips T=
=
sin e 0.26 The allowable working tension per bolt without any increase is 129.9 kips The above calculation assumes no less of energy during the impact due to the plate 8 drop.
In reality, a slight tilt or unevenness of plate 8 will lead to an impact condition closer to a plastic impact condition.
For a plastic impact condition, the total kinetic energy to be absorbed can be calculated as follows :
Velocity of impacting Plate =
V2x32.2xl/12 = 2.32 ft/sec.
V
= Y2gh
=
o Velocity of plates after impact =
'['
50 x 2.32 = 0.77 ft/sec.
V
=
V
=
1 m+M o
100+50 KE = 1/2 (m+M)(V )2 = 1/2 50+100 (0.77)2 = 1.38 k/ft 1
32.2 Thus the maximum vertical force induced in the plastic impact can be calculated as follows:
1/2 KX2 = KE 1/2x1.66x10 x(X)2 = 1.38; X = 0.00408 ft.
5 5
KX = 1.66x10 x0.00408 = 677 kips
NRC Ouestions (9/14/79)
,I l
O.
6.
Page 14 of 17 t
[
Adding the initial dead load of 100 kips, the maximum vertical f
force on wall during impact is 677 + 100 = 777 kips. Thus a plastic impact condition would induce a maximum vertical force f
on the wall which is smaller than that would result from the 4" I
drop of plate 8 on the HEXCEL pad.
g r
2.
Effect on Wall The wall on column line R has been investigated to determine its ability to resist the applied impact load of 1280 kips i
due to a 1 in, drop of plate 8 on the lower plates.
Both bend-f ing and axial loads, including the dead load of the wall, were considered.
It was determined that the maximum tension in
_/,
reinforcing steel is 10,600 psi, the maximum compression in the block is 460 psi and the maximum compression in the concrete is 1,180 psi.
3.
Effect on Equipment, Piping or Cable Trays The kinetic energy from a 1 in. drop of Plate 8 on the plates below is not large enough to cause a detrimental effect on equipment, piping or cable trays.
Effects of Plate 7 Drop 1.
Energy Absorption In order to preclude any possibility of detrimental effects on Plates 1-4, the structure, equipment, piping and cable trays
l NRC Ouestions (9/14/79) i l
0 6
Page 15 of 17 i
should a drop of Plate 7 occur, a HEXCEL pad will be placed on Plate 4 to absorb the energy of the drop.
The HEXCEL pad e
l will be 4" wide, 24" long, and 12" thick.
It will be stabil-
[
ized and precrushed giving a 750 psi crushing strength.
It will be attached to the top of Plate 4 as shown in Figures 6-5 and 6-6.
A " shoe" under Plate 7 will spread the load.
The Z bars shown in Figure 4-1 in Licensee's response dated Septem-ber 5, 1979 to Systems Branch Ouestion No. 4 will guide the plate.
The analysis to show the adequacy of this system is as follows:
Weight of Plate 7, W = 3 kips Maximum drop height, H = 14. 7 5 f t Maximum kinetic energy, KE = 3 x 14.75 = 44.25 ft-kips or KE = 44.25 x 1000 x 12 = 531,000 in-lbs The kinetic energy is to be absorbed by the deformation of the HEXCEL pad.
The required thickness is determined as follows:
t
= honeycomb core thickness c
S
= depth of crushing required A
= bearing area KE
=f
- ^* 8 cr f
= 750 psi er 2
A
= 24 x 3-1/2 = 84 in S
= 0.7t c
i t
j NRC Ouestions (9/14/79) i l
Q. 6 Page 16 of 17 t
531,000 = 750 x 84 x 0.7tc 531,000
= 12.0 in.
t
=
c 750 x 84 x 0.7 Thus, a 12-in. thickness will be adequate.
2.
Ef fect on Lower Plates Due to Plate 7 Drop on HEXCEL Pads.
The vertical force induced in the wall holding plates 1 through 4 would be F = 2x1.3 x f x A = 2x1.3 x.750 x 84 = 163.8 kips.
cr Adding the dead load of plates 1 through 6, the total force
/
is 163.8 + 97 = 260.8 kips. Using the same approach as for pj.
plate 8 in calculating the forces in the bolts, it was deter-mined that this case is not governing.
3.
Ef fects of 12 in. Drop of Plate 7 on Lower Plates.
The maximum kinetic energy associated with a 12 in. drop of plate 7 is:
KE = 3 x1 = 3 kip f t.
Since this is less than the kinetic energy due to plate 8 drop of 1 in., it is not governing.
i i
f NRC Ouestions (9/14/79) i O.
6 Page 17 of 17 4.
Ef fect on the Wall The forces induced in the wall due to the drop of plate 7 are smaller than those resulting from a drop of Plate 8 as determined previot. sly.
Therefore, the maximum stresses induced
+
in the wall will be smaller and likewise will not cause any jfg detrimental gross effect on the wall.
5.
Ef fect on Equipment, Piping or Cable Trays The kinetic energy from a 12 in. drop of Plate 7 on the plates below is not large enough to cause a detrimental effect on equipment, piping or cable trays.
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Schwencer February 13, 1980 1 Revised figures for Licensee's response dated June 29, 1979 to MRC Question 3 dated May 18, 1979.
Figure 3-5 Figure 3-6 Figure J-7 Figure 3-13 Figure 3-14 Figure 3-16
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