ML20147F096
| ML20147F096 | |
| Person / Time | |
|---|---|
| Site: | Trojan File:Portland General Electric icon.png |
| Issue date: | 10/16/1978 |
| From: | Richard Anderson, Johnson T, Katanics G PORTLAND GENERAL ELECTRIC CO. |
| To: | |
| Shared Package | |
| ML20147E636 | List: |
| References | |
| TAC-08348, TAC-11299, TAC-8348, NUDOCS 7810180207 | |
| Download: ML20147F096 (267) | |
Text
/~%
(_)
O UNITED STATES OF AMERICA NUCLEAR REGULATORY COMMISSION BEFORE THE ATOMIC SAFETY AND LICENSING BOARD In the Matter of )
)
PORTLAND GENERAL ELECTRIC COMPANY,) Docket No. 50-344 et al. ) (Control Building
) Proceeding)
(Trojan Nuclear Plant) )
)
LICENSEE'S TESTIMONY OF RICHARD C. ANDERSON, GEORGE KATANICS, THEODORE E. JOHNSON, AND WILLIAM H. WHITE ON l CAPABILITY OF TROJAN NUCLEAR PLANT TO WITHSTAND SEISMIC EVENTS 1
i I. INTRODUCTION This testimony addresses technical issues relevant to the l l
continuing safe operation of the Trojan Nuclear Plant, '
considering the discovery that the amount and arrangement of reinforcing steel in walls of the Control Building does not satisfy fully the criteria of the FSAR. This defi-ciency resulted from the discontinuity of some wall rein-forcing steel and misapplication of some design formulae.
The testimony first provides a description of the affected structure and the deficiency. We then describe the de-tailed evaluations performed since the discovery of the deficiency, which demonstrate the ability of the Control 73 N \S&'
Page 2 i
i Building,-as-built, to withstand a " Safe Shutdown Earth-
~
quake" (SSE). [1/ We next assess the " Operating Basis
' Earthquake" (OBE) which the as-built facility will meet.
Finally, to provide some perspective on the conservatism involved in seismic design, we identify and discuss sig-nificant factors of conservatism which underlie our confi-i dence that the Trojan Control Building will safely with-stand an actual earthquake with substantial margin above the SSE, and contrast the Control Building design with the i
seismic design requirements for other important buildings in the Portland area.
On the basis of these evaluations, we conclude that the Control Building is entirely suitable for safe operation of the Trojan Nuclear Plant, in that.the plant can be safely shut down under design basis earthquake conditions.
II. DESCRIPTION OF STRUCTURES AND BACKGROUND The Control Building is composed of a structural steel fl/ In Appendix A, we describe the results of re-lated investigations ' conducted to ensure that safety-related equipment is not adversely affected by the defi-ciency and that similar design deficiencies do not exist in any other Category I structure.
k Page'3 framing system with steel beams and columns supporting reinforced concrete floor slabs, and with shear walls de-signed to resist lateral seismic loading. (Reference 1, Attachment 2, Figure 2). f2/ The major shear walls are are located around the perimeter of the building and are generally composed of reinforced concrete core placed be-tween two layers of reinforced concrete block walls.
(Reference 1, Attachment 2, Figures 5 and 6). The block walls generally sandwich the structural steel frame so that the reinforced concrete core is partially or com-pletely interrupted by the steel frame members. (Refer-ence 1, httachment 2, Figures 3 and 4.) The major reason for this type of construction was the desirability of erecting a steel frame building to allow protection from the weather for the continued work inside. Later, the concrete block shear walls were erected and the reinforced concrete cores were placed between the two layers of l 42/ Attached to this testimony, and a part of it, are three documents, entitled " Reference 1," " Reference 2,"
and " Reference 3," which provide in more technical detail the basis for the testimony. Reference 1 is identical to the supporting documentation supplied to the NRC with the ;
Licensee Event Report (LER) of May 5, 1978. Reference 2 is identical to the package of supplemental information on the LER supplied to the NRC on May 24, 1978. Reference 3 is the " Trojan Control Building Supplemental Structural Evaluation" dated September 19, 1978.
I l
.k 1 . . -.
0, ss Page 4 reinforced concrete block masonry. Thus, the building is designed with the steel frame carrying most of the normal vertical floor load, and the block and concrete walls i
carrying most of the lateral loads caused by earthquakes.
The Fuel, Auxiliary, and Control Buildings (collectively the " Complex") of the plant are interconnected by their foundation systems and floor slabs. Up to an elevation 48 feet above grade, the lateral resisting members of the Fuel Building consist of the conventionally reinforced concrete fuel pool and hold-up tank enclosures, connected by reinforced concrete floor slabs. The upper portion of the Fuel Building is structural steel. The Auxiliary Building is between the Fuel Building and Control Building.
(Reference 1, Attachment 2, Figure 1.) The Auxiliary Building is supported laterally in part by both the Con-trol Building and the Fuel Building, with the reinforced concrete floor slabs acting as diaphragms to transfer lat-eral loads.
In Apri; 1978, during an investigation of the feasibility of cutting an opening and installing a security window in a wall of the Trojan Plant Control Building, it was
O 4
Page 5 4
determined that a deficiency existed in the original de-sign of the Control Building. Simply stated, the defi-ciency is that there is less continuous r einforcing steel
- in the Control Building shear walls than would be neces- I i 1 sary to fully satisfy the criteria of the FSAR. The defi-
- ciency is due to two distinct design problems. l 1
4 The first, and the more significant, problem contributing 1
to the deficiency was that some reinforcing steel embedded in the concrete core of the Control Building's shear walls
-l- was not continuous. That is, come of the vertical and horizontal reinforcing steel bars were not designed as continuous since they were interrupted at locations where the earlier erected steel beams and columns existed. We now believe that they should have been either welded to or s
run through the steel beams and columns or run outside the steel framing.
l l The second problem contributing to the deficiency was a 4
l misapplication of two formulae, one of which is a code de-sign formula used to determine concrete shear capacity, and the second of which is a mathematical expression used to determine the quantity of reinforcing steel to be
LOL
'Page 6 I 1
embedded in the concrete. /3/ '
43/According to the American Concrete Institute Code, 1967, the shear stress carried by concrete, V c , shall not exceed 2.0vf (fc is the specified compressive strength of the concr[ete) unless more detailed analysis is made in accordance with other provisions of the code to substanti-ate the use of a higher allowable concrete shear stress.
In this latter case, the value of the shear stress carried i by the concrete could be as high as 3.5/f[ . In the orig- l inal Control Building design, the capabilitLof the con- I crete to carry shear was assumed to be 3.5(fc . Absent i further jus value, 2.0 f c
'fication or additional analysis, the lower I
, should have been used. Assigning a higher shear-carrying value to the concrete resulted in a requirement for a lesser amount of reinforcing steel.
The proper formula to calculate reinforcing steel is :
1.4v
( , ")-Ve f
Y where 1.4 is the load factor, V u is the total applied de-sign shear. force, V is the shear force carried by the con-crete, f y is the yield strength of the reinforcing steel and & is the capacity reduction factor as specified by the code.
as:
This formula was misapplied in the building design 1.4(V u-v e)
$f Y Use of the correct formula would have resulted in a great-er amount of reinforcing steel-.
The effects of these misapplications (1) of the concrete capacity formula by assigning the concrete a higher allow-able shear stress and (2) of the' reinforcing steel re-quirement expression (underestimating the need for rein-forcing steel) were additive,-both tending to underesti-mate the need for reinforcing steel.
(3
%-)
Page 7 As a result of the combination o'f the two design problems, the amount of reinforcing steel available to transfer shear forces in the wall does not fully satisfy the re-quirements of the FSAR.
III. ASSESSMENT OF THE CONTROL BUILDING FOR " SAFE SHUTDOWN EARTHQUAKE" Appendix A to Part 100 of the Commission's regulations provides the criteria which guide the NRC in its evalua-tion of a nuclear power plant's design bases established in consideration of the seismic characteristics of a site.
The regulations describe the approach used to assess con-servatively a site's potential for seismic activity, the resultant effects of seismic events against which the plant must be designed, and the particular structures, systems and components of the plant whose design must re-flect those considerations. The basic parameter for seismic design is ground acceleration seen by the plant expressed as a fraction of "g" level (gravity acceleration of the earth).
Page 8 i
In NRC terminology, the " Safe Shutdown Earthquake" (SSE) defines that earthquake which has commonly been referred to as the " Design Basis Earthquake". Based upon an evalu-ation of the maximum earthquake potential at any site, the
- SSE is that earthquake which produces the maximum vibra-tory ground motion for which certain structures, systems, and components at a nuclear plant are designed to remain 4
functional. It is well recognized that the possibility of such a large earthquake occurring during the lifetime of a plant is extremely remote.
l The SSE for the Trojan plant as defined in Section 2.5.2 of the Final Safety Analysis Report (PSAR) is 0.25g.
Thus, the structures, systems, and components identified in Par t 100 as required to. remain functional during the maximum earthquake which could affect the plant must be designed to withstand a seismic event which causes a 0.25g ground acceleration at Trojan. /4/
In April 1978, following the discovery of the deficiency ,
in the Control Building, a detailed technical evaluation f4/The SSE ground acceleration of 0.25g accommodates an intensity of VIII on the Modified Mercalli Scale.
l Page 9 '
was undertaken of the control Building structure, as built. As is stated in attached References 1 and 2, this re-evaluation study utilized the design basis criteria of 0.25g, the design response spectra, f5/. damping values, methodologies, and general structural criteria including load combinations and load factors, all of which were originally used.in assessing the adequacy of the Control Building design and are defined and described in Sections 3.7 and 3.8 of the FSAR. Both the original analysis in 1971 and the re-evaluation study were based upon a fixed-base beam-stick model. f6/ (Reference 3, Appendix A.) The only differences between the original analysis Z5/ A response spectrum is a plot of maximum re-sponses of simple systems to an earthquake motion. Bach point on a response spectrum represents the maximum re-sponse to an entire earthquake motion of a system which is defined by its natural frequency and fraction of critical damping. Natural frequency is the frequency at which a system will vibrate freely with no externally applied force (after motion has been initiated in some manner as by displacement). Critical damping (as discussed in more detail in Section V infra) means 100% damping where the energy dissipation is so high that a vibratory motion will not be realized for the system.
the m[ ass associated with each level of the structure is6/ In a basic beam-stick m lumped into one concentrated mass and the concentrated masses are interconnected by vertical sticks and horizon-tal beams which represent the stiffness characteristics of the walls and floors, respectively.
i
(~T
\_/ ,
t 1
i
. Page 10
-(reported in the FSAR) and the re-evaluation study per-formed on the as-built structure are the following:
A. The concrete strength for the Control Building shear walls was specified in the FSAR. analysis to be 5000
, psi. A review of the concrete cylinder test results on the concrete actually in the Control Building shows a strength of 6000 psi. (Reference 1, Appendix C).
Therefore, 6000 psi was used in the re-evaluation study. (Reference 1, Attachment 3). l 1
B. Minimum yield strength for the reinforcing steel was specified in the PSAR analyses to be 40,000 psi. A review of the mill certificates of the material actu-
, ally furnished and used in the construction of the Trojan Control Building shows a minimum yield strength of 45,000 psi. (Reference 1, Appendix C.) A minimum yield strength of material actually used, 45,000 psi, was utilized in the re-evaluation study. (Reference 1, Attachment 3).
tO V
Page 11 2
C. The weight of the Control Building and the equipment it houses was reviewed based on as-built conditions l l
and estimated to be only 87 percent of that assumed in i the original analysis. This estimated weight was used for the re-evaluation study. (Reference 1, Attachment 3).
D. The capability of some interior walls in the Auxiliary Building--for which credit was not taken in the orig-inal analysis--was considered in the re-evaluation study as reducing the amount of shear force trans-ferred from the Auxiliary Building to the Control Building. (Reference 1, Attachment 3).
E. The criginal analysis combined the contributions of the building response modes by the " absolute sum" method rather than the equally acceptable " square root of the sum of the squares" analytical technique (SRSS). The SRSS technique was used for the re-evaluation study. (Reterence 1, Attachment 3.)
Based on the above, the re-evaluation study of the as-built Control Building demonstrated the capability of the l
O 1
I Page 12 shear wal'Is of the Control Building to resist the seismic shear forces-induced by the 0.25g SSE, even with the con-servative 5% damping value used. T" 3 asis of the conclu-sion was later confirmed by the Tt 'vsis. /]/
As part of the re-evalution study, ntrol Building structure was also examined by completely disregarding the concrete shear strength and considering only " dowel _ac-tion" of the reinforcing steel and the embedded steel col-umns, f8/ (Reference 2, Section 4.) The concrete was l
l only used in bearing to transfer the dowel forces. In this evaluation (Reference 2, Section 4) it was assumed )
47/The TABS Analysis, which was performed in June 1978, and is described in Appendix A to Reference 3, em- l ploys the program TABS (Three-dimensional Analysis of Building Systems). The model idealizes the building sys- !
I tem as an assembly of a system of independent plane frame and shear wall elements interconnected by floor diaphragms which are rigid in their own plane. The outputs of the TABS program are moments and shear forces on the walls and, as reflected in Table A-1 of Reference 3, indicated internal loads lower than the re-evaluation study.
[8/This analytical method is not normally required or performed to demonstrate the adequacy of a structure's seismic design. The analysis was done at NRC's request to provide additional confidence as to the adequacy of the Control Building.- The method of analysis is very conserv-ative because the assumed cracking in the concrete and as-sociated yielding in the reinforcing steel would actually increase the damping, thereby decreasing considerably the structure response and seismic loads. (See discussion of damping in Section V of this testimony.)
Page'13 that the reinforced concrete block walls are cracked all the way.through the most critical horizontal plane and that the. entire shear force continues to be resisted only by the reinforcement and steel columns fully embedded into the shear walls acting as dowels across the crack. The most critical location was examined to determine how much of the available energy was needed to resist the load.
Based on the required strength, about 12% of the available ;
energy was required. (Reference 2, Section 5.1). It was )
concluded that even with this unrealistically conservative l assumption, the structure has substantial excess ultimate capacity to resist SSE loads.
1 Additional confirmation of the validity of these assess-ments of the Control Building was obtained by use of fi-nite element computer analyses initiated in August 1978 to determine the responses of the building due to earthquake.
The STARDYNE computer program was used for these analyses.
STARDYNE is a finite element computer program provided by Control Data Corporation, used nationwide by engineers, analysts and scientists. The program is maintained by Mechanics Research Institute.
I
O Page 14 i
The finite element model for the Trojan Control-Auxiliary-Fuel Building Complex was originally developed over-a pe- )
riod of three months for the analysis of the as-built structure combined with the proposed structural extension at the north end of the Control Building. This model was intended to be used for the analyses required to design the planned modifications because of its mathematical so-phistication beyond that of the beam-stick model to evalu-ate interaction between the existing structure and the proposed extension. This finite element model was revised to omit the structural extension and, thus, to model solely the existing Control-Auxiliary-Fuel Building Com-plex in its present. condition. The finite element anal-yses were then able to provide an independent check of the existing structure. !
The beam-stick model emplo"ed in the original analysis and i re-evaluation study consists of lumped masses, sticks, and beams. In this model, all the mass associated with each of the floors in the Control Building, Auxiliary Building, Hold-up Tank enclosure structure, and Fuel Pool is lumped into one concentrated mass, and these concentrated masses are interconnected by vertical sticks and horizontal 4
- - , , , _ . . ,. w..,,.
q
(_/
Page 15 beams, representing the stiffness characteristics of the walls and floors respectively. The finite element model developed for the STARDYNE program has the building Com-plex represented by approximately 460 nodal points tied together by 685 plate elements representing walls and floors and 56 beam elements. This finite element model provides an excellent representation of the actual mast.
and s'iffness distribution within the Complex. The pro-gram solves simultaneously approximately 600 equations of motion which describe the dynamic responses of the Complex.
The basic results of the STARDYNE analysis compare very favorably with the results of previous analyses with re-spect to the overall shear force in the Complex (19,590 ,
kips total base shear for STARDYNE versus 26,140 kips for the original analysis and 18,480 kips for the re-evaluation study). The STARDYNE analysis did indicate increased seismic forces at the Control Building end of the Complex.
The analysis showed that the total base shear at the Con-trol Building end of the Complex is approximately 20%
greater than the load predicted in the re-evaluation study.
However, using criteria based on recent actual test data and a more complete evaluation of the strength (capacity)
I L _ __ _ __ _ _ _ _ _ _ _
Page 16 l of the walls, it has been shown that the Control Building has adequate capacity to withstand the predicted loads. i (Reference 3, Section 5).
It is to be expected that a finite element technique will show higher localized loads, since it takes into consider-ation more precise mass and stiffness distribution of the structure and consequently provides an upper limit indica- )
tion of the forces in the structure. When using this more i j
sophisticated elastic analysis tool, it is appropriate to l use realistic wall capacities based on test results. ;
(Reference 3, Section 4.) The more conservative code al-lowable stress limits, applied in the earlier analyses, were developed for use with conventional, less conserva-tive and more approximate analytical techniques such as the beam-stick model.
The STARDYNE analyses also allowed us to investigate some alternative hypothetical conditions which are far more se-vere than those which are actually anticipated. It should be emphasized that these conditions are not expected to occur. However,'their investigation develops better understanding of the structures and structural behavior A
O Page 17 and, thus, develops more confidence in the buildings' in-herent ability to resist earthquake loading. These studies, reported in Reference 3, Section 7, hypothesize situations in which cracking and building deflections oc-cur which are far greater than can be expected. These studies show that the building Complex has the capability to redistribute load and to withstand all of these unex-pected conditions without loss of function. These studies also demonstrate that the Control Building has substantial inherent strength, not fully accounted for either in the original analysis or in the re-evaluation study, which will come into effect in the hypothetical situation where major shear walls would not perform as well as expected.
These further investigations develop additional confidence that the structure will withstand the design basis earth-quake with a substantial margin of safety.
In summary, since the discovery of the deficiency, mul-tiple analytical techniques have been employed to assess 1
the capability of the Trojan Control Building and the interconnected Control-Auxiliary-Fuel Building Complex to resist the seismic shear forces induced by the 0.25g SSE. f In April 1978, the control Building was evaluated using
)
i Page 18 the same beam-stick model approach which had been utilized in the original design but with certain modifications in this re-evaluation, including recognition of the as-built condition and known properties of materials used in con-struction as well as use of the SRSS method.to sum build-ing response modes. At the same time, the structure was evaluated. postulating that the concrete walls cracked all the way through and the entire shear force was borne only by dowel action of the reinforcing steel and steel columns.
Most recently, finite element analyses of the Complex have I also been performed which confirm the results of earlier analyses, provide more detailed information on specific structural elements in the Complex, and demonstrate the capability of the Complex to redistribute loads under var-ious hypothetical seismic load conditions. All these ana-lytical techniques, taken collectively or individually, !
result in the same conclusion--taking into account the de-ficiency in reinforcing steel in shear walls of Trojan ;
Control Building, the structure will safely withstand the seismic shear forces of a 0.25 SSE.
i
(
t Page 19 IV. THE OPERATING BASIS EARTHQUAKE In addition to the design basis Safe Shutdown Earthquake, Appendix A to 10 CFR Part 100 provides for the establish-ment of an Operating Basis Earthquake (OBE).
Appendix A to 10 CFR Part 100 provides that if vibratory ground motion exceeding that of the OBE occurs, shutdown 4
i of the plant will be required. Prior to resuming opera- l tions, the licensee must demonstrate to the Commission l J
that no functional damage has occurred to those features necessary for continued operation without undue risk to the health and safety of the public.
For the majority of the nuclear, power plants in the United States, the maximum vibratory ground acceleration of the selected Operating Basis Earthquake is one-half the maxi-mum vibratory ground acceleration of the Safe Shutdown Earthquake. Recently, some nuclear power plants have been
O Page 20 licensed by the NRC with an OBE of less than one-half the SSE. 29/
In contrast, the OBE selected for the Trojan plant, 0.159, ,
is 60 percent of the SSE (0.25g). 'The Trojan plant thus is one of the few nuclear power plants in the United States for which the OBE selected is greater than one-half '
the SSE.
The evaluations of the existing Trojan Control Building shows that while the structure satisfies design require-ments for the Safe Shutdown Earthquake, its capacity to resist the Operating Basis Earthquake is 0.11g--less than the selected OBE of 0.159 (Reference 1, Attachment 3; Reference 3, section 2.) i This result illustrates that while the Operating Basis Earthquake is always by definition a less severe earth-quake than the Safe Shutdown Earthquake, the design of L9/For. example, the Byron and Braidwood plants.
These licensing decisions recognize that the provision in Appendix A to 10 CFR Part 100 for determination of the OBE to be at least one-half of the SSE (like the entirety of Part 100) is only guidance to the Commission in its evalu-ation of the suitability of plant sites and design bases which need not strictly govern in every situation.
/PT I
'/
(
I Page 21 l
I nuclear structures may be governed by the OBE and not the SSE. This situation results from the. application of more conservatism in designing and evaluating for the Operating
, Basis Earthquake. The principal additional conservatisms applied for the OBE are even lower damping values (2% i rather than 5%) and a load factor greater than 1.0, i.e., 1.4. /10/
As a consequence, in the case of the Trojan plant, the selected Operating Basis Earthquake of 0.15g results in i actual margins above the requirements of the 0.25g Safe Shutdown Earthquake. The capacity to resist the intended l OBE of 0.159 at 2% damping results in an SSE capability at 5% damping of approximately 0.34g--a margin of roughly one-third over the Safe Shutdown Earthquake requirement.
This margin will be restored, by appropriate modifications to the Control Building, through the restoration of the
/10/ Damping of a structural system, which is dis-cussed in greater detail in Section V of this testimony, is the inherent ability of a structure to dissipate the vibrational energy. Had less conservative but more real-istic 4% damping been allowed in the evalua* ton rather than the 2% value used, the capacity of t' control Building to resist the OBE would have been 0.15g. It should be noted that factored OBE loading associated with a ground acceleration of 0.15g would certainly be above working stress and a higher damping value would seem justified.
t
O >
Page'22
' building's capacity to resist an OBE of 0.15g, compared to its existing capacity to resist an OBE of 0.119 Since the structure satisfies the design requirements for the SSE and since the significance of its capability to resist an OBE of 0.11g, rather than 0.15g, is that it will be necessary to shut ~down the Plant and evaluate its in-tegrity following an earthquake with the smaller accelera-tion, there will be no undue risk to the health and safety of the public from operation of the Trojan plant in its '
as-built configuration.
V. FACTORS OF CONSERVATISM IN SEISMIC ANALYSIS AND DESIGN MARGIN Our conclusion that the Trojan Control Building can safely resist the Safe Shutdown Earthquake is further supported by placing in perspective the overall conservative ap-proach to nuclear power plant seismic design generally, and to the design of the Trojan Control Building specifically.
- bs
<J Page 23 A. Damping One of the major f actors of conservatism involves the values used for the damping characteristics of struc-tures. A structural system under vibratory motion has the ability to dissipate the vibratory eneroy. This can be demonstrated by the fact that when we induce vibrating motion in a spring by pulling and releasing it, the amplitude of the vibration becomes smaller and smaller from one cycle to another, and eventually the vibratory motion stops. This ability to dissipate vi-bratory energy is an inherent property of any real structural system and is called, in engineering terms, the damping characteristic of the structural system.
Damping in a structural system exists due to the pres-ence of the internal friction in the structural mater-ial itself, the frictions at the structural joints and connections, the air resistance, and the dissipation of energy intc the foundation soil. Furthermore, the degree of damping depends not only on the building ma-terial, but also on the severity of vibration to which the structure is subjected. That is, a structure under a more severe vibratory motion will manifest
O Page 24 higher damping ability. than that of the same structure under a smaller vibration.
The degree of energy dissipation or damping of a structure is measured, in engineering terms, by the ,
damping ratio, or more conventionally, the damping value. The damping value of a structure is usually expressed in terms of the percentage of critical damp-ing, for example, 2% or 5%. A higher damping value implies a higher energy dissipation ability of a structure. A 100% damping means critical damping--
which implies that the energy dissipation is so high that a vibratory motion will never be induced in the system. .
For building structures, the damping values are nor-mally in the 2% to 15% range, depending on the struc-tural materials, types, construction, and the level (amplitude) of vibratory motion. Steel structures usually have lower damping values than those of con-crete structures. Masonry structures usually have even higher' damping, i.e., in the 10% to 15% range.
_.w.rr , ..-., ,,,. -.. ..
- n. , .,
. ,,,.w. . .
O Page 25 The application of seismic criteria to the engineering design of concrete structures in a nuclear power plant is very sensitive to the damping value used. For ex-ample, if 4% damping is used instead of 2%, the Oper-ating Basis Earthquake, which the Trojan Control Building has the capacity to resist, would increase from 0.11g to 0.15g. The selection of 2% damping to design the Trojan Control Building for the OBE /11/
reflects substantial conservatism in that (1) 2% damp-
'ing for' reinforced concrete structures generally is I considered to be low; (2) the building is not a rein- \
1 forced concrete structure, but a combination of struc-tural steel, reinforced concrete, and reinforced and grouted concrete block walls which should have higher damping ability; and (3) stresses in reinforcing steel will exceed normal working stress levels during an OBE event with factored loads, which would contribute to higher damping.
/ll/The FSAR requires the use of damping values associated with certain stress levels. The values for reinforced concrete structures are 2.0% "at working stress level" and 5.0% "at yield point."
1 - -_. _ _
f' Page.26 B. Inelastic Deformation i
Building structures have the ability to deform in- '
elastica 11y, i.e., beyond the elastic limit, before l reaching ultimate capacity. In case of a severe earthquake, this inelastic deformation ability will increase- the ability of the structure to dissipate the vibratory energy. In addition, as structural elements exceed yield, the response of a building is lower than if they remain elastic. The seismic loads of the Trojan Control Building in the re-evaluation study and in the supplemental STARDYNE' analysis were determined based upon linear elastic dynamic analyses in which the energy dissipation ability or damping of the structure was conservatively specified based on the structural stresses and deformations within the
" elastic" limits. Reference 3, Appendix E, describes further the conservatism inherent in elastic analysis.
C. Strength of Materials Under a dynamic loading environment such as earthquake loadings, the dynamic strengths of building materials
.1- - , , , . , , . , . . ,,
L /~s O
Page 27 j are higher than the static strengths. Evaluations of the Trojan Control Building were conservatively based upon the static strength of materials. Also, the ac-j tual ultimate strengths are larger than the ultimate strengths of materials used in the re-evaluation study and supplementary evaluation.
4 D. Rigid vs. Nontigid Foundation i
The seismic analyses of the Trojan Plant assumed that 4
the plant is founded on rock with infinite rigidity.
i This is a conservative assumption because the infi-nitely rigid foundation does not allow any struct.tal 5
vibration energy _to be absorbed by or dissipated into j the foundation. Incorporating the actual elastic T
properties of the foundation material and the energy dissipation (damping) characteristics of the founda-1 i
tion' system in the seismic analyses would lead to lower but more realistic seismic responses.
1 i-Considering all this conservatism in seismic analysis and design margin built .into the Trojan Control Building, and considering all of the analyses and studies done on this 4
,m-<
/^)
(_/
Page 28 structure, it is our professional judgement that the Trojan Control Building, as built, would safely withstand an actual earthquake at least 50% larger than the Safe Shutdown Earthquake. The significance of this margin can be clearly shown by a direct comparison with the require-ments for other important buildings in the Portland area.
The existing Control Building has about twice the seismic l l
capability as is presently required for the newest local l hospitals, schools, and fire stations. /12/
l l
1 VI.
SUMMARY
AND CONCLUSIONS l
l Extensive evaluations have been conducted of the structur-al design and as-built condition of the Trojan Nuclear Plant Category I structures, following the discovery that the amount and arrangement of the reinforcing steel in the
/12/ Uniform requirements for seismic resistance de-sign of shear wall box-type structures are adopted by most building authorities in the western United States. Under the 1976 Uniform Building Code, the seismic capability of the Control Building is almost three times that required for normal buildings, and about two times that required for hospitals, schools, and fire stations. In other words, all other private and public buildings in the Portland area--including even the newest, most conserva-tively designed hospitals, schools, and fire stations--are required to be designed for only slightly more than half of the seismic loads which the Trojan Control Building will safely withstand.
{O '
t l t t
4' Page 29 '
i t .
I walls of-the Control Building do not fully satisfy the '
l r j criteria of the FSAR.
L i -
i
! Based on all.our investigations and evaluations, the:large '
j margins of safety designed into these structures, and the i
calculated strength of these structures, we conclude that 'l the Control Building.is entirely. suitable for continued 4
- safe operation of the plant.
I 4
5 i-1 l -!,
3 .
?~ r h
i i-2 I 2
5 I
r
-...s . -- , n. . , , -1. ,.a.-s. . - .e- . . - - --- --n .
O Page 30 3
Appendix A RELATED INVESTIGATIONS We have considered any potential effect the design defi-ciency might have on equipment housed in the Control Building. As shown in Reference 3, Appendix D, upper limit displacements in the Control Building during an SSE will be very small, about 0.5" maximum lateral displace-ment between the floors and less than 0.9" total displace-ment at the roof level. These displacements have been conservatively calculated based on cracked concrete prop-erties. fa/ A survey was conducted to determine the capa-bility of safety-related equipment, piping and electrical cabling to withstand displacements. Conservatively, the survey investigated the sensitivity of the safety-related components to take larger building displacements to confirm that even if much larger displacements were to take place, the function of such components will not be La/Use of cracked concrete properties in an evalua-tion means that it is assumed that concrete walls. crack in ;
tension and the load in the tension zone is carried only by the reinforcing steel. This results in a more flexible response by the structure and thus in larger calculated deflections.
1
()
Page 31 affected. The survey concluded that the safety-related equipment located in the Control Building can withstand large displacements (greater than 1 inch between floors and 3 to 6 inches between adjacent buildings) without loss of function.
i Since the finite element analysis (Reference 3) identifies slightly increased rigidity (i.e., higher first mode natu-ral frequencies), the original seismic qualification docu-ments for safety-related components, equipment and systems were reviewed to determine if the possible changes in the dynamic characteristics of the control Building have sig-nificant effect on the previously qualified equipment.
With one exception, /b/ it was concluded that the original seismic qualification of the safety-related components, equipment, and systems are not affected by the deficiency.
All of our analyses have shown that the control Building and the safety-related equipment inside will not be dam-aged by the Safe Shutdown Earthquake. Postulating, how-ever (Dr. McCollom question, Transcript 6589-90), that if Zb/It was found that a 3" cooling water line to some room coolers would require small restraints to change its resonant frequency into the acceptable range.
,~. l 4
us
) <
1
)
Page 32 some local structural failure were to occur, it would be no more than cracking of the walls or possibly some dis--
lodging of pieces of concrete from the structure. Postu- l l
lating even further that if some subsequent damage to electrical equipment inside the Control Building were to result, the plant can be safely shut down using manual controls in local areas outside of the building. As an illustration, were the electrical control equipment for the auxiliary feedwater pump (turbine-driven) to be dam-aged, that pump could be operated manually from a remote station located outside the Control Building. This pump in combination with the mechanical relief valves and safety valves (which require no signal from the control room) would be adequate to maintain the plant in a safe condition until controls for normal shutdown systems (sim-ilarly poctulated to be affected in the Control Building) could be reestablished.
-All other Trojan Category I structural designs were inves-tigated during the re-evaluation study to determine if similar design deficiencies existed that would in any way affect the continued safe operation of this facility. It
l 1
l Page 33 was found that no problem existed in other structures and that they meet the requirements for seismic design.
l l
l l
)
l 1
l
_ _ _ _ _ _ . _ __ _ _ _ __ m a
/~
V) i l
I l
l l
REFERENCE 1 TO l LICENSEES' TESTIMONY OF RICHARD C. ANDERSON, GEORGE KATANICS, l THEODORE E. JOHNSON AND WILLIAM H. WHITE ON CAPABILITY OF '
TROJAN NUCLEAR PLANT TO WITHSTAND SEISMIC EVENTS l
~
~ ~
OL '
i Attachment 1 Page 1 of 3 ORIGINAL STRUCTURAL CRITERIA Strength, load Determination, and Load Combinations
- 1) Factored load design using ACI-318, USD.
- 2) Load combinations 1.4D +'1.4E, and.1.00 + 1.0Es *
- 3) Seismic criteria:
a) Earthquake ground response spectra:
- Horizontal ground acceleration OBE = 0.15g and SSE = 0.25g.
4
- Damping OBE = 2% and SSE = 5%. -
- Vertical ground acceleration = 2/3 of horizontal.
b) Design based on one horizontal and one vertical component combined absolutely. Both north-south I and east-west directions were considered.
The original design required that only the Control and Fuel Buildin~gs were to resist lateral seismic loads with the exception of the Auxiliary Building walls which extend from Elevations 45 ft to 61 ft shown in Figure 5. Continuous floor diaphragms were designed to transfer the Auxiliary Building loads to the Control and Fuel Buildings. In the Control and Auxiliary Buildings, the structural steel was designed to
-carry vertical loads. -
Shear Walls _ _ _
. The determination of shear _ capacity was based on the following:
- 1) Assumed full concrete core plus 1/2 concrete block area. '
, r. ,- , y v. r - ,, .mp._ - - . . . -r, --
--em~
Attachment 1 i Page 2 of 3 J
- 2) Detemined: Vc =$3.'S&c bt (Eq. 1) 1.4(Vu-Yc)
- 3) . Reinforcing:. As" (Eq. 2) dy I
l where l
l Vc = the allowable shear resisted by the concrete i
+ = capacity reduction factor (.85) fj=concretestrength(5000 psi);designvaluestated on the design drawings .
b = unit length of wall ,
1 t = wall thickness based on Item 1)
A3 = reinforcing steel per foot of wall height- l Yu = the applied shear load 1
fy= reinforcing steel yield strength (40 ksi).
Seismic Analysis . . _ _
The seismic analysis was done using an extensive model which considered the Control, Auxiliary and Fuel Buildings, but the Auxiliary Building was considered to have no lateral resistance except for the walls. The analysis was based on the following: '
- 1) . The mass considered was based on the dead weight and 50 percent of the ~specified floor live load.
- 2) The stiffness was based on uncracked section properties. <
' Attachment 1 Page 3 of 3
- 3), The modal analysis spectrum response technique was used for the detennination of inertia loads.
F
- 4) 'The modal' responses were combined using the absolute sum value technique.
O 4
9
)
I m.
h- &
, _ _ _ . _ _ _ _ . _m . _ _ . . _ , . . - - - - -- - " ' * " - " " " ' ~ * - '" ' ^ ~ ~ - ' ' ' " ' - " " ^ ~ ~ - "
.- ~
.O-
~
Attachment 2 -
Page 1 of 8 ;
1 1
l STRUCTURAL DESCRIPTION -
, The general plant lay'out 'is shown in Figure 1. The Control, Auxiliary :
and Fuel Buildings are supported on a common foundation and connected by j floor diaphragms and shear walls. Figure 2 shows' the Control Building which is 93Ift long, 77 ft wide and 72 ft high. The Control Building is composed of a structural steel system of beams and columns and temporary i bracing. In addition, a system of shear walls were used which are l composed of a reinforced concrete core with two layers of reinforced concrete block. The construction sequence was as follows: -1
- 1) The column foundations were constructed,
', 2) The steel frame of columns and beams were erected, 4
4
- 3) The concrete grade beams, base slab, roof and floor slabs were constructed, and l
- 4) The reinforced concrete block and the concrete core were constructed.
This method of construction'was used so that a structure could quickly be erected and a roof installed thereby allowing the remaining construc-l tion to be performed with protection from rain. Typical wall reinforcing i was:
l
- 1) Block vertical reinforcing, #6 at 24 in., each' face continuous.
- 2) Core vertical reinforcing, #6 at 24 in., continuous in west and south walls; discontinuous in east and north walls.
- 3) Block horizo'ntal reinforcing, two #5 at 24 in., each face, bars j- in outside block are continuous; bars in inside block are continuous in east and west walls only.
. -%r ._ o - n, .- - v,
.,n.- -. ,.- -
-O !
Attachment 2 Page 2 of 8
- 4) ~ Core horizontal reinforcing varies'in size and spacing and is continuous in the west wall. The minimum reinforcing is
- 7 at 24 in.
Reinforcing details are shown in Figures 3 and 4. j The shear wall layout for Elevation 45 ft to 61 ft is shown in Figure 5, and the layout for Elevation 61 ft to 77 ft is shown in Figure 6 for the 4
Control Building and a portion of the Auxiliary Building.
The Auxiliary Building was constructed similar to the Control Building.
The Fuel Building used V;ical cast-in-place reinforced concrete j construction.
I
. I 1
l
\
1 a
y , * -- =v- - -
w a - w- - - - - - -
O. .
Attachment 2 Page 3 of 8 t
i FIGURE 1 1
1 STRUCTURAL DESCRIPTION j i
i t
j i
l l AT 2 _
l
.LN FUEL BLDG. l l
.i l dame . _ _
\
I >
l l
i e l l Aux. ,
I BLDG. .
CONTAINMENT .
I l
l'
\
! l s
! CONTROL
! BLDG.
immmmmmmmmmmmme TURBINE P
BLDG.
(
_ . . ~ . . . _ _ _ . . . _ . _ . . . . . . _ . _ .
am
o . .
l Attachment 2 Page 4 of 8 FIGURE 2
{
CONTROL BUILDING - STEEL FRAMIllG i
1
.1 i
l A .-
l EL. 117
[
/ /
EL. 93 / /
EL. 77 __
[ -.
EL. 65 '
/ / q, .
i*
EL. 45 -
3 9 31 = 93' ! .. - - _
.- . , _ ..y .
Q .
=
Attachment 2 Page 5 of 8 FIGURE 3 PLAN SHOWING TYPICAL HORIZONTAL REINFORCING EAST AND WEST WALLS 2-#5a24"
, -1 -
l v l j 1 ,
f
... _._._ _ _._..#6a24" /
e Q -
k ~ - . . . - .
i ;
,' e/ o I
. HT.' o [# hah"n s N.6.:.- ,
[_ !
.k'
....n9.' ; I '. 's o ;.. -D ;.v- = .
. VJ . Ul .
~. -. ..
lS J -
4-.
2-#5924" 4
,, e-+
Attachment 2 Page 6 of 8 FIGURE 4 TYPICAL WALL. SLAB CONNECTIONS I
-' ~
hM6924" ae .
. 'z, , i
"~~ ~
ed / !
i
, .u \ ..<.
~
'.* \
t
' s. ', I '
v, -
_ _ e. -.
- 6a24,, x _
h n
/7 - J, - . _ . .- .
j.
. gga2q"
. . a
= ' T 1 . .. . .
4 g
c = e- = .,
.c;
. ~
T 4 . ,
e
~> _
- 6924" y -
.4-
. ]- -( -
N
( '
Attachment 2 Page 7 of 8
, FIGURE 5 SHEAR WALLS - ELEVATIONS .45' TO 61' 41 46 55 W
^4 e*e .em-
- w*w._ w... ,, 4, .
R
O. .
Attachtnent 2 Page 8 of 8 FIGURE 6 SHEAR WALLS - ELEVATIONS 61' TO 77' 41 46 55
~
m 7
.4 N
e
- a R-4 T
-. ~. ,
O.
Attachment 3 1 Page 1 of 7 l INVESTIGATION Preliminary Review A preliminary review was made of the structural criteria and design of the Control Building. Based on this investigation, it was deter-mined that:
1
- 1) Due to the type of construction and amount of reinforcing, the shear strength of the concrete should be based on 2 instead of $3.5 as indicated in (Eq.1) of Attachment 1 (Page 2).
I
- 2) The load factor in 1.4 in (Eq. 2) of Attachment 1 should have only been applied to Yu and not both (V -Y u c*
- 3) The amount of shear wall pier moment resisting reinforce- ;
ment was potentially too low based on the technique used.
This technique was based on " Analysis of Small Reinforced ;
Concrete Buildings for Earthquake Forces" by Portland l Cement Association, 1955. This technique required that the flexural tension be based on Mc/I and then reinforce- .
ment by supplied to resist the total tensile force. l l
- 4) A review of the design and fabrication drawings indicated that some of the reinforcing was not continuous as illus-trated in Figures 3 and 4 of Attachment 2.
A review of the seismic analysis indicated that it was within the state-of-the-art at the time of design and it gave appropriate results for '
l lateral loads.
l i
l
i Attachment 3 l Page 2 of 7 Final Investigation General b
Even though the preliminary review indicated possible problems, struc-tures usually have inherent conservatism built in during the ' design stage. A new analysis must be made which accounts for all factors ;
including both unconservative and overly conservative ones when making a final judgment as to adequacy.
Capacity Determination .
l l
l The capacity of the structure was recalculated based on the following '
criteria:
l
~
- 1) Side wall shear capacity was based on e V = 2(bt (1.4Vu /*I'Y and A s" f
- c. Concrete thickness was based ~ - -
y on fuli core with 1/2 concrete block area to account for the reduced strength of the block. Only the continuous rein-forcing steel was used for capacity determination.
- 2) The moment capacity of the shear wall piers was based on I limiting the concrete strain to be .002 in./in. The strain in the outer reinforcing steel was allowed to approach twice yield. The general design criteria is based on the ultimate l strength principles of ACI-318. The limit of .002 in./in.
was used instead of .003 in./in. to guarantee ductile behavior. It would be unreasonable to base capacity of a beam or wall with reinforcing distributed over the depth )
by only allowing the outer reinforcing at yield, since a significant amount of other reinforcing exists in the section. Twice yield is a conservative strain. More ee
(:)
Attachment 3 Page 3 of 7 i details on this technique are given in Appendix 8.
In addition, very few of the shear walls are limited in capacity by the local bending manent.
- A detailed review of the Auxiliary Building revealed that there were several additional walls in both major of rections that can resist 3
seismic loading even -though they were not considered 'in the original analysis. These walls would directly add to the strength and not cause a change in inertia forces since the original fundamental frequencies were in the peak spectral response range. Any change in '
frequency would be an increase and result in even slightly reduced loads. However, this was not considered in the investigation. Since the natural frequency is a function of the square root of the stiffness, minor stiffness changes would only have a small effect on the calcu-lated frequency.
A review of the reinforcing steel mill certificates showed a lower limit yield strength of 45 ksi. A review of the concrete cylinder break strengths at 90 days concluded that the concrete strength is at least 6000 psi. See Appendix C for additional information.
Load Determination .. . . _ _
The dead load was recalculated and actual equipment weights were us'ed '
as identifiable live ic ' instead of the 50 percent specified live load approximation for equipment weight. This resulted in a more realistic mass which is 87 percent of that originally used.
The FSAR referenced Bechtel Topical Report BC-TOP-4, " Seismic Analysis of Structures and Equipment for Nuclear Dower Plants", April 30, 1971, for acceptable methods in combining modal responses. The topical states that either the absolute modal sunnation or the square root of the sum of the squares (SRSS) is acceptable. The SRSS gives a more realistic method since it agrees well with time hist'>ry solutions w
__ ._~. _. . __ _ _ _ . _ _ _ _ _ _ _
' Attachment 3 Page 4 of 7 for structures in the frequency range of the Control Building. The SRSS results in response values that are 80 percent of the absolute '
summation values.
Investigation Results i
The following terms are used in summarizing.the results of the investigation:
- 1) Absolute & SRSS Techniques - methods of combining the modal responses. The SRSS technique results in loads which are 80 percent of the absolute value technique.
- 2) Design Loads - based on originally calculated weight (or mass).
- 3) Actual Loads - baseo on as-built weight information, '
87 percent of original design loads.
- 4) Design Capacity - based on original Control Buf1 ding and additional existing Auxiliary Building walls. For i shear strength y f =40ksi,ff=5000psiand
+ = .85; for flexural. strength yf = 40 ksi and , = .90.
- 5) Actual' Capacity - same as Design Capacity except fy =
45ksiandff=6000 psi;resultsinastrength increase of 1.10 for shear capacty which governs design.
The following earthquake conditions were examined in detail. Eg is the Operating Basis Earthquake (OBE), E3 is the Safe Shutdown Earthquake (SSE) ,
8 is the damping and 1.0 and 1.4 are the load factors:
- 1) 1.4E gand a = 2% - this is the most conservative interpretation of the FSAR for the Operating Basis Earthquake condition.
- uur
-.a. a , -. _- ,, . . . . - ,
9 3
. u ~
Attachment 3-
, Page 5 of 7 '
, 2) 1.4E and g s' = 5% - the FSAR sta+.es in Table 3.7-1 that the damping may be 5 percent When the calculated
- stresses are near yield. Since factored load design is i being used, stresses are near yield and therefore this is a reasonable interpretation.
- 3) 1.0E sand a = 5% - this is the design condition stated in the FSAR for the Safe Shutdown Earthquake.
- 4) 1.0E and s a = 7% - this is a reasonable damping value
, for structures of this type when stresses approach yield.
, 5; 1.0E sand a = 10% - this is included to illustrate upper ,-
limit capacity when the structure is still expected to remain essentially elastic in overall response. '
i Compliance with the SSE design criteria is required'for public safety and the OBE design criteria to ensure operating capability.
Table 1 sumarizes the results of the investigation. This table contains '
results determined by comparing the predicted loads with the calculated capaci ty. It shows the ground motion that is acceptable for the structure to remain within the previously defined criteria:
Column h shows the values which were based on originally calculated loads and design strength. This uses high loads and low capacity. '
. Column is similar'to Column h but accounts for the as-built strength. '
. Column has included the actual as-built weight and strength.
h em
! ,. t l
Attachment 3~
Page 6 of 7 '
-* Column h is the.most representative as to the actual jL condition of the structure.
Columns h and h are included for information.
Row 1 of Column h - shows an OBE capacity of .11g; however, this is a conscevative interpretation of the FSAR. Row 2 of Column shows that for a damping value of a = 5%, the capacity exceeds the h
. required .15g. Row 3 of Column h illustrates that the capacity is sufficient to resist the specified SSE of .15g with the FSAR i damping of a = 5%. '
l 1
J l
s O
r)
. (J TABLE 1 IllVESTIGATION RESULTS "g" LEVEL CAPACITY ABSOLUTE SUM TECHillQUE SRSS TECllNIQUE DESIGil LOADS & DESIGN LOADS.& ACTUAL LOADS & ACTUAL LOADS & ACTUAL LOADS &- DESIGN LOADS &
DESIGN CAPACITY ACTUAL CAPACITY ACTUAL CAPACITY ACTUAL CAPACITY DESIGN CAPACITY DESIGN CAPACITY EAPlHQUAKE AND DAMPIllG @ @ @ @ @ @
1.4Eg .07 .08 .09 .11 .10 . 09 8 = 2%
1.4E g .11 .12 .14 .18 .16 .14 8 = 5%
l.0E s .16 .17 .20 .25 .22 .19 8 = 5%
l.0E s .20 .22 .25 .31 .28 .24 8 = 7%
l.0E .23 .25 .29 .37 .33 .29 s
8 = 10%
M R
se
~
~
r M
Attachment 4 Page 1 of 1 L
BUILDING LAY 30T
~
l J k l' . . . . .
AN ! FUEL BLDG.
i d
.LW f J
1 -
d I
! Aux. l i
BLDG.
f CONTAINMENT i i l i
I l l !
~ I l !
l i CONTROL
._._ _ _.._. __ i ,
I i n l. -
i g, \ b .
~
i i
STRUCTURAL TURBINE EXTENSION BLoe.
13 '~."
U .
e I
l l
APPENDIX A ,
1
SUMMARY
OF THE HORIZONTAL MODAL ANALYSIS l
l OF THE CONTROL, AUXILIARY AND FUEL BUILDINGS l
l l
l 6
4 O
b e
~
7
(3-V The Control, Auxiliary and Fuel Buildings form an L-shaped structural complex as shown in Figure A-1. These t'hree structures are tied together by common slabs at elevations 66', 77' and 93'. For the purpose of the seismic analysis, the structures were represented by a space frame supporting lumped masses at the various ficar and l l
roof elevations. The stiffness of various elements in the struc- l tures was represented by beam elements. The properties of the beam elements included bending, shear and torsional deformation. The model, giving the location of the lumped masses, is shewn in Figure A-2. The magnitude of the masses is given in Table A-1. The lateral resistance of the Auxiliary Building is neglected which results in the inertial loads being transmitted to the Control and Fuel Buildings through the co mon slabs. -
The frequencies, participation factors and the modal effective weights
- from the modal analysis are summarized in Table A-2. The first 15 modes were included in the analysis, and as can be seen from Table A-2, modes 1, 2, 6, 8, 9 and 10 are the dominant medes.
The sum of the effective weights of these modes are 63,450 kips and 61,905 kips for the x (E-W) and z (N-5) directions, respectively.
These dominant mode shapes are plotted in Figures A-3 through A-8.
e- <.,,,...,-m. ww-.. .
. *The modal effective weight = gr2=(acceleration of gravity)x (participation factor)2 A-1 .
. . -I
~
r L) . . -
I TABLE A-1 LUMPED MASSES MASS M NUMBER KIPS 1 5852 2 2521 3 6541 4 5740 5 8055 6 9318 7 5070 8 2710 9 3334 10 2268 11 4977 i
12 6834 1 13 3978 22 1600 Total: 68,798 .
O e
A-2
- suus.
o
i r
TA8LE A-2 RESULTS OF MODAL ANALYSIS
~
MODE FREQUENCY PARTICIPATION FACTOR EFFECTIVE WEIGHT (KIPS)
NUMBER (Hz) rx rg W N x z 1 6.2 2.408 6.1 63 6011 39384 2 6.9 6.329 2.571 , 41526 68 54 3 7.5 0.477 0.248 236 64 4 8.3 0.124 0.78 4 16 637 5 10.2 0.298 0.684 92 485 6 11.6 3.012 0.274 9407 78
.7 13.8 0.504 0.196 263 40 8 14.7 2.104 1.421 4590 2093 9 15.5 0.933 2.331 902 5635.
10 16.6 0.989 2.753 1014 7861
. 11 18.6 0.687 0.492 48 9 251 12 23.1 0.031 1.482 ,
1 2278 13 24.8 0.293 0.0 31 89 1 -
14 26.5 0.158 0.411 26 175 15 27.5 1.009 0.076 1055 6 Total Effective Weights: 65717 65842 e
A-3 2
e ampte W
O s
Control Building ) [
8 i .
I l Auxiliary i Building Fuel l l Building ,
i 1
.I s l R N r
+ .
@- D A FIGURE A-1 PLAN. VIEW OF CONTROL. AUXILIARY AND FUEL BUILDINGS l -
- 1. '
0; a 't E1. 116.0 40 4;22 El. 43.0 3o y7 21 g 10 Control Auxiliary ~Hoid-up Tanks ,
[ Building / Building 20 El. 77.0 2o j,6 E :; 9 13 eSpent Fuel Pool El. 61.0 5 19, 1 ,,. e ,, ;
7 8 '12 T
f, A \ \ !
\ C _ _ _\
\ \ .
\ w flinge # 'I \
, \ \ \
(Building \ 2 83.0' Outiine .. \
I N 18 \
g WK- N
! l \ \
- \ _ _. _\
96.0' _
'N__
! 179.b'
' j X -
i 'A l !\_ 207.0' y _s FIGURE A-2. SEISMIC HATilEHATICAL N0 DEL j l.
j-
-<4 i ; ; ,
i
-m...
1
'\/ -
/\
1 l
f,
/
s --~........_.._..__
,x . .
5 l 79
<w N"
8 .-
2'" l l
\ \ \ \ \
N
\ %, . .
- ~
/\ ,
1
.. l
't a, s / .
\
/
l A-6 en IO
l I
l l
I
\/
\
/ //
// \ _s/
/\
m
<u w%
5 .
$'"W
.E
\\
\\ xAv
/\
~wan
\ s. \/ _
A-7
- GD
em. - -
a
~ ~
O .
t
.\i :: .
/\
/
\ jj
/\ ,, r7 ... ..
' /\
- E!
.N S .'
u.
W g \
- \/ . . _ _ _ . _ _ . _ . _ . _ . .
/\
I
= .
, N/ ,
I I -
l A-8 '
O -e g 6
,m -
4 , w , , _ . -- , - - , . , - - - - - - - - - - - - - . - - . - -
M
~
O .
M 4 4- 4%d Me ne
/\
i - . . . - - - - . . - -.
/\
N 2".
N
@ k m W
W CC %
hb te W
8 x
-. - \/ - - - . _ , - . . . . - . . . . .
9
_ : .= .
\lJ ,
A-9 -
T
-- a , - a , , ,a , _ - , . - -- -,
l O .
I. -
5 1
3 4
i l
! c
}
\/ .. -
/ /\
4 4
4 1
4 3
. [ ....#m_+ e a+
..=.-me. ~.
/\ w
? =
h Ln.
$ b N
+ e m 5'
4 u
i . LaJ M %.
l-5 A
LD m
6 4
N O 4
E l
1
. =e g hu .* 'wM .4
- w-+-,
3-
/\
a l
5 4 4
i 1
4 5
T d
y m%
N/ -
j \
/
4 4
J f
e s
A-10 og
- sv.e +,e weemamsmepame++ , . ene.m * .neuene ., , ,
- **M - # -"our JI W
y , , a a _
O ll 1
I i l
'l l l
I 4
n l-i 1
l 8 i
.\ j . -.-. .
\
/
t / /\
I a
i h
i y - . - . . ..~ - - .
l .
x N a '
/s .
.g '
i = .
j w
. O '""
i s 4 M
$ N h k
l 8'*
~o 1 LA. m
! LAJ 1
8 x
gj .._. - - . _ . . . - _ . . . .
. /\
i j
i i
b i
i 4
4 l
1 .
- \/ /
\
t e
A-11 .
g ,
eemne e e . .
4 es eue
o U .
APPENDIX B BENDING CAPACITY OF SHEAR WALLS WITH UNIFORftY DISTRIBUTED REINFORCING STEEL l
l e
e e g I
O 4
In determining the capacity of a shear wall to resist lateral loads
-which lie in the plane of the wall, the s' hear capacity as well as the bending capacity must be determined. Since the reinforcing in the walls. is often uniformly ' distributed along the length of the wall, the bending capacity equations ~ developed for beams are too conserva-tive. To obtain a more realistic assessment of the bending capacity !
of shea'r walls, an expression for the moment is developed based on
)
the following assumptions.
- 1) Plane sections before bending remain plane during bending.
- 2) Compressive stresses in the wall 'are assumed to vary linearly since the stresses are small for an under-reinforced wall.
- 3) Tension is resi.sted by the reinforcing steel only.
- 4) The reinforcing steel is uniformly, distributed along the
. length of the wall.
- 5) No axial load, l
The strain diagram and associated stress diagrams for the reinforcing . i steel and the concrete block-concrete core wall (concrete / masonry) are shown in Figure B-1. In this figure:
. j ej=steelyieldstrain c
e = maximum strain in the concrete / masonry wall y '= ductility in reforcing steel i E, =' modulus of, elasticity of steel !
Eg = modulus of elasticity of concrete / masonry wall !
8-1 ,
O
. 59
o .
Using the forces shown in Figure B-1, (b) and (c), equilibrium is satisfied if:
J T) + T2 = C) + C2 The equation can be written in terms of the material characteristics defined above and the geometric quantities shown in Figure B-1.
1 From the resulting equation, the extent of the compression zone (c) 1 can be determined:
- - ~
c= 2X)-/2K) K2 d ~ ~ ~ ~ - ~
K)-K2 -
where: ,
X) = f,As (I ~ )-
, X2 " "*y(Es3 A + Ec t) where:
A3 = reinforcing steel per unit length t = wall thickness i d = wall width With the size of the compression zone known, the moment carried by the section is given by the following expression: l f A d2 -
l M= Y$ 2 (I ~ + I I~ * ~ ~ ~ ~ ~
. 3d(1-c/c) U ~ .
The design capacity is given by: l M' = 4M B-2 S
I i .-
g 4
0 .
The strain in the concrete should be checked with the following equation which i's derived from geometric considerations:
pcy C l c, = i-c < .002 For lightly reinforced walls or beams, cc is well below this limit.
Examole Determine the mcment capacity of a shear wall which has the following properties:
A3 = 0.441n2/ft = 0.0367 in:/in, fy= 40 ksi, u = 2 t = 24", E c= 3000 ksi, E = s
.30,000 ksi ey= 0.0013, d = 180" X) = 40(0.0367)[1 2(2)] = 1.10 K = 2(.0013)[30,000(0.0367) + 3000(24)] = 190.06 2 ,
~
c= 2(1.10) - /2(1.10H M0.06) 180 -
- 1.10-190.06 .
c=(0.097)(180)=17.38" 1
x M=h(0.0367)(180)2-2(.097)+(.097)2 3(2) + 3( f (2 -
= 18.221 k-in
= 1618 k.-ft 8 . ,
o
-i
_ .__ _. . . ~.. . . . .. ._ _ . . _ . _ - _ - . . . _ - .
O .
The design capacity is given by: )
M' = $M = (0.9)(1518) = 1367' k-ft Now check'the strain in the concrete:
- y" c =
- 2(0.0013)(17.38) = 0.00028 < 0.002 --
e d-c 180-17.38 e
i 1
e 1
e 8-4 .
9 I
O. .
d <
im I
1FN
~
e 1 . . . - . - . . ..
center of y -
N 2 "c -
outermost steel (a) Strain Diagram i l
T ,
1
'2 g 1 E s*c
- fy C
1
(.b) Steel Stress And Force Diagram l l
l 1
- ~ . _ _ _ . _ _ _ - . .
ec < 0.002 (c) Concrete Stress And Force Diagram FIGURE B-1 STRESS, STRAIN AND FORCE-DIAGRAMS O
B-5 m .
g
e g 8
9 1
l 1
l l
l l
l APPENDIX c l;
MTERIAL STRENGTHS e
i e
o e
I l
e
=== w e
O . .
7 C.. MATERIAL STRENGTHS This appendix surmarizes the actual strengths of materials utilized -
in constructing the Control Building.
C.1 Reinforcing Steel -
Table C-1 sume.arizes the average and minimum yield strengths of the Grade 40 reinforcing steel utilized during construction which includes the steel used in the Control Building. Strength values i
shown for the differene bar sizes were obtained from the manu-facturer's mill certificates. The average yield strength for 67
- lots of steel was 48.72 ksi with a standard deviation of 2.16 ksi. !
The minimum yield strength recorded was 45.08 ksi. Therefore, a reinforcing steel yield strength of 45 ksi is. considered appropri-ate for assessing the strength capacities of the Control Building.
C.2 Concrete 1
1 1
Two concrete mixes, 0-1 and D-2, were used in the wall cores of the Control Building, both of which had a specified design strength l of 5000 p'si. Mix D-1, with a maximum aggregate size of 3/4", was i
used in well cores 12" and less in thickness. Mix D-2, with a maximum aggregate size of 1h", was used in vall cores over 12" thick.
C-1
"'* 15'
j Table C-2 summarizes the 90 day strength properties of mixes 0-1 .
- i. and 0-2. These data were determined from the monthly concrete r
cylinder test record for a time period between April 1972 and June 1973, which corresponds to the construction period for ;
t the Control Building walls (February 1972 through February 1973).
The criteria'for determining the required average concrete strength .as outlined in the ACI 318-71 code was utilized in determining the usable concrete strength values. The three criteria are summarized as follows:
- 1) fcr =f;+1.282a
- 2) fer. " #$ + I '343# ~
- 3) fer =
f{+2.326a-500 ,
where:
f cr = average concrete strength (psi) fy = design concrete strength (psi) a = standard deviation for measured concrete strengths (pst)
Using the 90 day concrete strengths over the construction period l of the Control Building for mix 0-1 and mix 0-2, and lett'ing <
l f
cr = x (average concrete strength), the following values of fg are determined:
l l
l 1
C-2
. ~P
O. .
MIX CRITERION NO. 1 CRITERION NO. 2 CRITERION NO. 3
)
D-l 6444' 6390 6008 l 1
0-2 6206 6168 6068 i
1 Criterion No. ~3 controls and indicates a usable concrete I'
strength of 6000 psi.
t 4
0 e
e e
0 e
t 9
r C-3
- O
~~
TABLE C-1 YIELD STRENGTH OF GRADE 40 REINFORCING STEEL (FROM MILL CERTIFICATES)
BAR SIZE AVERAGE YIELD MINIMUM YIELD (NO.) STRENGTH (XSI) STRENGTH (KSI) 3 53.64 53.64 4 52.38 51 .7 5 5 .49.03 48.06 6 50.60 47.95 7 49.35 48.17 8 48.52 46.83 l 9 48.72 4 5. 50 !
10 46.32 !
45.08 11 47.08 45.19 14 47.78 47.78 -
- - {
- ~n TABLE C-2 AVERAGE CONCRETE STRENGTHS
~~~ *
~
. - . . . . - ~ .
OF TEST STRENGTH NO. OF DEVIATICNS TIME SPAN l MIX (DAYS) i (PSI) TESTS (n) o (PSI) 0F TESTS l
0-1 90 7593 159 896 5/72-6/73 ff=5000 3/4" Aggregate
~
D-2 90 6989 128 61 1 6/72 6/73 fj=5000 1 " Aggregate 9
4 4
. C-4
.
- ve
O.
v i k
l l
REFERENCE 2 q i
TO LICENSEES' TESTIMONY OF RICHARD C. ANDERSON, GEORGE KATANICS, THEODORE E. JOHNSON AND WILLIAM H. WHITE ON CAPABILITY OF TROJAN NUCLEAR PLANT TO WITHSTAND SEISMIC EVENTS y
8 I
- j /G v .
f- ,
May 24,1978 1
I i
4 1
i i
TROJAN IUCLEAR PLAtlT i
i i
Docket 50-344 a
T j
- SUPPLEMENTAL INFOR.'!ATION 1
! TO LER 78-13 i,
)
4 4
)
i 1
i I
I 4
i 1
i 1
J.
i i
O ,
TABLE OF CouTENTS SECTIC" PAGE
- 1. DETERMIPATION CF SHEAR RE3ISTED BY THE CONCRETE l-1
i 2.3 CONCRETE SHEAR CAPACITY CRITERIA 2-2 .
i 2.4 EFFECTOFSLASC0!$CRETECHSHEARCAPACITY 2-7 4 2.5 CONCRETE MIX DESIGN 2-10 1
2.6 CONSTRUCTI0il JOINT PREPARATI0H 2-12 !
l
- 3. SHEAR WALL REIHFORCI;lG 3-1 i
- 4. DETERMIllATION OF SHEAR CAPACITY BY DOWEL ACTION ONLY 4-1 ;
4.1 MATERIAL PROPERTIES 4-2 4.2 LOADS 4-3 4.3 DOWEL SHEAR CAPACITY CRITERIA 4-4
- 5. COMPARISON OF LOADS Ai!D CAPACITY BY 00$EL ACTICH ONLY 5-1 5.1 STRUCTURAL DISPLACEMErlTS 5-3
- 6. C0!! TROL BUILO!!!G TORMADO RESISTAt:CE 6-1 REFERE!:CES 1
e TG-LER78-13 ,
I 4 s
- 1. DETERMINATION CF SHEAR RESISTED DY THE CCNCRETE In determining the portion of the capacity darived from shear strength of l the concrete, several approaches are possible. The approach used in the original calculations is based on an equivalent :nickness equal to the thickness of )
l the concreta core and 1/2 the thickness of the blocks. This approach is based I cn cells in the block being fully grouted.with a concrete mix having a strength equal to or greater than that of the core and the cross-secticnal area of cells approximately equals 1/2 of the gross cross-sectional area of the block. An alternative approach is to consider the shear stress in the core equaY to 2/T7 and the shear stress in the block equal to 75 psi applied over c
the total area of the block'. These two methods for determining the shear capacity, Vc, can be summarized by the following:
Method 1 .
V c
= 2Ftc c +2F ct 1B ,
where f8 = compressive strength of the core concrete '
c t = ' thickness cf the con rete core tlB = bidth of one block = 7-5/8" M2thod 2 Vc = 2/TT c c t + (75 psi)t 2B where '
t2B " "n of tuo blocks = 15-1/4" .
1 -1 a o 4 e
' r
- r. .. .
d<s .
For f' = 5000 psi, the folicwing is obtained:
c ftethod 1: Vc
- I4It +c 141t)g ,
Itethod 2: Y = 14it + 150t IB c c 4
/
For f' = 5000 psi:
c Method 1: Vc = 155tc+ 155t)g - i l
Method 2: Vc = 155tc + 150t 1B .
l l
Since this is the range of values for f', the results will be essentially i 1
the scea. l O
e 4 e I
e e
1-2
.i
p t
- 2. CCNC'1ETE SHEAR CAPtCITY OF SI: EAR WALLS 2.1 INTRCCUCTICM The capacity of a shear wall to resist lateral loads which lie in the plane of the wall depends on the bending ar.d the shear resistance.
The specifics for determining the capacity when shear centrols is outlined belcw. The approach is applic.able to shear walls constructed o'f two concrete block wythes with a concrete core in between the biccks.
Along with this, an example is worked which considers both bending and shear behavior. -
2.2 DESCRIPTION
OF WALL The wali is constructed of two bicek wythes; the space between them is l filled with concrete. For the block' cells which contain vertical rebar, an open-er.ded block is used. This permits the bicek to be placed without threading it Ov2r the retar (dos Figure 2-1). In the courses which contain hori: ental rebar, a bond beam type bicck is used which removes the robar from the totar joint. For all walls, the vertical ' cells in the block are grcuted with a mix which has a compressive strength greater than er equal to that of the core concrete.
The grout fills all the areas in the bond beam as well. !
s I 9 2-1
N ,h .
1 2.3 CC.' CRETE SPEAR CAPACITY CRITERIA The shcar capacity is determined frcm the shcar capacity of the con-The shear tinuous hori:cntal stcol, the block, and the concrete core.
capacity in the block and the concrete core is obtained by taking an I equivalent thickness equal to the thickness of the core and 1/2 of the l total bicck thickness.
toff = t core +I2tbicch .
" The shear stress carried by the block and concrete core ever this equivalent thickness is limitad to 2/I[. The criteria for the capacity is expressed by the following equations:
cap " Y c +Y s ,
t.'h e re .
VC = 2 /I7C t QT T.(0.81W)$
Vs = Asy f (.81 w)$
and ,
f' = cc,Tpressive strength of the core concreto c
g,= length of the wall 4 = capacity reduction = 0.85 for shear A = continuous hori:cntal reinforcing steel fy= yield stress of reinforcing stcc1 .
o_?
,~\ .
V t
Examole J
l The follo. ting exa:::;13 is wori.ed using the above criteria and the bending capacity expressions su=arized in App.endi.i B of the Trojan
- Control Build'ng F.epcrt. The wall considered is sh'sm in Figure 2-2 As shewn in the figure, the hori: ental reinforcing steel is continuous in both wythes but the horizontal reinforcing steel in the core is not.
For determining V , nly the 4-H 0 2 7 are usei s
The vertical steel in the bicck is cont'incous past the upper and lower j
. edges of this wall segment, but the vertical steel in the ccre is not continuous past the upper edge. There is a steel beam along the upper edge spanning between the columns preventing the passage of the vertical steel. '
' 1
\
Shear capacity based on shear:
V cap' c
+Y s .
For pier A:
Vc + Vs = 2F terrc (0.82. w)t + A sy f 1(0.31.w)
S e 9 t
6 e
a 0
2-3
(n.J where f' = 5C03 psi c
teff = 15.25 + 7.63 = 22.03 in i
g = 12 ft 4 = 0.85 A
s
= 4-55 0 24" = o.62 in2/ f t f
y
= 40,000 psi V; + v3 = 2/Soco (22.83)(0.8)(12)(12)(0.ss)+(0. 62)(40,000)(0.a5)(0.s)(12) xio 3
= 316.8 + 202.4
= 519.2 kips For pier B: _
v+v= c 3 2 /3300' ( 22. 88) (0. 8 ) ( 30 ) (12 ) ( 0. 85 ) + (o . s 2) ( 40, coo) (0.s 5) (0. a ) ( 30 ) xic :
= 792.1 + 505.9
=.1293 kips The shear capacity based on bending also needs to be determined. The procedure used assumes both piers A and B fixed at the base of the ', tall and at the top of the doorway. The shear in the piers is obtained frcm:
M +N' y= T "
H whero M T and M g are the bending mcments at the top and bottom of the piers respectively, and H is the height of the pier. The bonding mcment is givcn by: -
A D2 5 ?"
!! = cf -y -- (21 - 2 a + 2 2)[]. )_av + 3Ti-a) (2 n 1)]
l 2-4
i
. f*
- The dovcicpment.of, this expressicn is su:a::ari:cd in Appendix B of the Trojan Control Building Report.
For determining the bending r.:oment at the base of the wall and the top of the docruay, the continuous vertical robar is the #6 bars in the 1
bicek and'the #6 bar in the core. This gives the folicwing steel effective for bending.
Block 06 0 24" = .44 Core - liG 0 24" = .22 "
.66 in2 /ft For pier A: *
~
a = 2K3 - /2K) K2 t
K)-K2 -
where K1 = f ys A (1 2a)
. X 2 " kCy(EA ss + Et)c and y =2 (alicws outer reinforcing to approach twice yield)
A.3
= 0.66 in2/ft = 0.055 in2/in f
y
= 40 ksi c
y = 0.00133 +
E = 30,000 ksi .
3 E
c
= 3000 ksi t = 22.80 in a = .110 2-5
- . e d ,
e
e o a i
V 1
l 1 2 1 M = 0.00(40)(0.50 u /1212 [1 - 2(0.110) F(0. ll S)2] [1 g + y( 0.11 j'-)g ( 2 1 2 ( ; _ g, 7)]
l l
0.G5(40b'O.66)(1?' ',0. /c ,
2
, 4) l l
= 1289 k-ft l
Since MT*HB l l
l V =
2k = 2(1289) = 322 kips l
i A H 8 .
1 I
For pier B:
a = 0.116 M = 0.90(40)(0.66)(30)2(0.754) 2 = 8060 k-ft 8060 i V * --'(_2_) * ,,C20
~
' M p,s l B 8 At this level of capacity determination, the capacity of pier A is 322 kips controlled by bending and the capacity of pier 8 is 1298 kips controlled by shear. , , j 1
1 l
t
=
j i
g .
l
. i I
l-2-0 .
t
- O li d ,]
- e. a ;
.. s r
1r M! -
l b f g .h
. d
-~
i i i Dimensions - (Inch) l I 'l -
c i
I I
a=7h f = if s .
'.i I i
I
- i. b = 1 r8 9=4 i -
5 c = 76 h=2
.i j k 3 )
(a) Open End Block d = 14 i = 14 j=6 e = 5B-5 k=18 1
s .
dZ ~
j z
e a
~
.l - . --.
l j
M:
! i l b f g h I
~~
)
. Dim 2nsions - (Inch) 7 r1 I
l t c a through k same as above d I
! l l- t=2 .
l I. i l t f l 1,
i j k ,
.(b)BondBeamBlock FIGURE 2-1 C0t: CRETE P. LOCK
- . T.
L. - - . , , , , , - ~ . . . , . , . , - - + . - , , , n- - , . , . ,.,c,.
, , + - ~, .- -,n,.,
6' .
_ -A A 8' ,
t -' 1- - I-1 -- ' '
30$ 8' 12' s..
8 h--- ~ #6 0 24" M (2-#5 0 24" '
. . cp- ,L
.y . - . s- - - - - - - < -
3 5 ,, _ - .-
, g 1 l 0 18"
. . _ ~ . _ , - co -. - -
S 7 .,
5 Section A-A 2-#5 0 24" FIGURE 2-2 -
TYPICAL SilEAR ilALL
- m _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ - - _ _ _ _ _ _ _ . - _ - _ _ _m_. ___. - - __-__m_m____ _ _ _ _ . _ _ _ _ _ _ _ _ _ _ _ _ _ _ -_m_ _ _ _ _ - _ _ _ _ _ - _ __m__-r. -
'f\
J 2.4 EFFECT CF SLA3 CClCP.ETE C.' SHE!O CAPACITY The shear wall core concrete and the concrete bicek wall cell fill grout are 5000 psi design mixes. In many areas, the cencrete slabs are a 30C0
- psi design mix. Figure 3-4 of the Trojan Centrol Eui'. ding F.cport shows typical details. The following illustratos the shear transfer frcm the.
walls through the lower strength concrete slabs.
i J4 l
"e-73" 5 Blocks t
c ,
Y
... . Wall .
~
- ' .' ,n . core 4 . .'
._m _ _ . _
m ,-.
s., - '
s,:a tg Stao ' ,
T.-
"a.
d HALL - 51.AB DETAIL V is considered as limited by principle te sien and 2Fc is used. The .
c shear ~ walls have used a capacity based on the core centrate plus only one block width; therefore, the allowable shear force per unit length is:
V Cw
= (t C-+ 7. 5)(b)2 /f'Cw where t
c
= core thickness b = unit length ,
Qu = compressive strength of wall l' <
2-7
/~'\ .
4 V.
The slab shear strength at tne interface can be determined by tne folicwing:
vc,, = (t_)(b)2 f'., .
3 t
where ,
t g
= total interfac2 thichness f's l C = compressive strength of slab 1
By equating V 3 Yew, the following expression results for the required cs slab strength to develop the wall capacity: l 1
1 i
V cs " 'cw l
(t g)(b)2/f'cs = (t c+ 7.5)(b)2/f'cw 2
(t+7.5) c f, f's C =
t 9 _
Cu .
l The largest walls which are limited by shear are 2.'-8"; larger' walls are narrow and are governed by banding leading to a reduced capacity. The 2'-8" (32")
walls have a core of 32-15 = 17" and the required slab strength using f'w C = 5000 psi is:
2 .
(17+7) 5000 = 2931 psi f'sc =
a2 Since the slabs are 3000 psi, they can develcp the wall capacity for design strengths. . Calculations which were dade on actual stron}th used f'g = 6000 psi and the required slab strength is: _
2-8
m.
~
= (17+7 O 5000 = 3517 psi fj3 3j' Based on 90-day cylinder tests, the usabic strength fer the 3000 psi slabs is.3504 psi. Since the 4/ alue is within .4" of required, the slab is considered adequate to develop the aall when using actual strengths.
4 4
e O
4 1
l l
e 4
9 e
0 h
e S
sl 2-9
. = - _.
en (s_-)
2-5 CONCRETE MIX DESIG1 Concrete mix designs furnished by the ar:nitect-engineer are listed in FSAR Table 3.3-17. Mix designs applicable to the Contr:1 Cuilding core wall concrete are the D-1 and D-2 mixes having a design fc = 5CCO' psi.
The 0-1 mix was used for core walls less than 12" thick, and the D-2 mix was used for walls 12" or greater in thickness. The floor slabs used the 3000 psi B-1 mix with the exception of the slab at elevation 77' which used the 5000 psi D-1 mix. For confirmation, a copy of the actual mix designs used in the core wall construction is attached.
The mix design used for concrete block wall cell fill is given in Table 2-5-1. Mortar used between concrete block units was designed in accordance with ASD1 Designation C476, Type Pl. Representative compressive strengths for the grout and mortar based on actual test samples are provided in Table 2-5-2.
4 4
2-10
iO .
' TABLE 2-5-1 COMCRETE BLCCK WALL CELL FILL GRCUT MIX DESIGN One Cubic Yard Batch
[ ,
\
Material Pounds-(S.S.D.) ,
Cement 940 j Sand . 1705 I Gravel (pea) 748 l Water .(51.0 gal) 425 W = 141.4 pcf TABLE 2-5-2 COMPRESSIVE STRENGTH OF MORTAR )
AND . BLOCK GROUT FILL ACI 318-71 Concrete Strencth Criteria
- Criterien Criter1cn Criterion Mix Ho. 1 No. 2 Ho. 3 i t i l Greut 6175 6142 6109 Mortar 2623 2579 2259 Critorion Ho. 3 controls, and indicates a usable
' grout strength of 6100 psi and a usable certar strength of.2300 psi. -
- 28 day 2-11.
4
v..uu. _ v. _v ,
l' R ev. 65 Date 9 /21/ :
r O s A s. ., s. . . u_ ,.:..... 2 L. .. . .
T ,,.
' C O N C R E T E ~ X O C S
- C'5 3_
JOS 6473 l.
B-1 Mix (3' 00 nsi: 3 /4" ma::imum sized accree?.:eh .
Abs. Vol.
d SSD W eight s /c. v, 455 2'. 31 4,84 sk 8S% Cemen: 0.40 62 l 12% Po olan 0,66 sk S.35 44% Sand 1344 d
1765 10.63 56 % 3/4" 4.63
- 0. 56 V/ater -
289 w/c 8. 3 os ----
- 1. 5 o= /sk WRA*
- 3. 3 os 0.68
- 2. 5% 0. 6 o= /sk AEAce
- 27. 00 cu. It.
d 3915 = 145. O!/cu It.
27 .
'. B-2 Mix (3000 osi; 1-1/2" maximum s' iced acgrecateh !
a SSD V/cichts /c. v. Abs. Vol. l 434 2.20
- 4. 62 sk 8S% Cement 0. 33 12% Po solan 59
- 0. 63 sk 8.24 43% Sand .. 1327 '
955 5.75 30% 3/4" 5.17 27% 1-1/2" ,- 869
) 0.58 V/ater 2S6 4.SS 1 w/c 7. 9 os ----
~
- 1. 5 ' o:/sk VfRA9 0, 6 o=/ sk AEA** 3. 2 c: 0. 6a_
2.5% *, 27. 00 cu. h.
3930 =- 145. 66 /cu. f t.
27 0 When ambient temperature is below 700 F use Type A (1. 5 c:/sk)
' When ambient temperature is above 'iOO F usc Type D (2. O c:/sk)'. *
- Adju:t in ficid'to obtain air centent between 2-45 Section 2 ,
Attachment -
Og &
em
- W =
Y ~
Sheet 4 of 6 I r R ev. !!5 Dat e 9 /. ' 7 TROJAN NL C '_EA.7 PIANT ,
m CO .s.*p =x
.-. =' *- ==.
--- %..e *...
- , *- c7 :
-. 7.
r=* t
- u. % ! a' ,
JOB 6473 9
D-1 Mi:-: (5000 osi: 3/4" m2.::imum size accrecate):
SSD Weichts /c. v. Abs. Vol.
- 6. 60 sk 88% Cement 620 3.15 O. 90 sk 12% P,o::olan 85 0,55
. 40% Sand 1137 .,, - 7.06 60% 3/4" 1760 10,60 1 w/p - 0. 44 Water 310 4.96 i
- 1. 5 c /sk WRA* 11. 3 os ----
- 2. 5 % '
- 0. 6 o /sk AEA** 4. 5 os 0. 6 B ;
27.00 ca.?.
' 3919 ~
=
- 144. 9 E /cu. ft. .
- )
_. ,27, -
\
- i
- i
,. , i 1
D-2 Mi:: (5000 n si: 1-1/2" ma::imum si= e age- ere.te): '
I 1
l
. SSD Weich:s /c. v. Abc. Vol. '
.6.16 sk 88% Cement *579 2.94
- 0. 84 sk 12% Pozzolan -
79 0,50 35% San:1 1096 6.81 32% 3/4" 953. 5.74 I
~
30% 1-1/2" 904 5.38 ]
w/c 0. 47 Water 309 4.95 l
- 1. 5 o=/sk WRAo 10. 5 o:
- 2, 5% *
- 0. 6 c:/sk AEAcc 4. 2 o n 0.63
- * . 2 7. 0 0 cu. it.
3920 27
= 145. 2f /cu. ft. ,
o When ambie it temperature is below 700 F use Type A (1. 5 o /sh.) '.
When ambient temperature is abov: 70 F use Type D (2. O o:/sk) k
- o Adjust in ricid to obtain air content between 2-4%. <
.Section 2 - -
Attachment l .
. . .. i l-n
, 2.6 C0t STR!lCT!0:1 J0!!iT PREPARATI0ti
- i Concrete work for Trojan Plant safety-related structures was perfern
- 2dlin accordance with the provisions described in FSA?, Section 3.8.3.5.
- It was confirmed that construction methods used to develop bonding of the' l
concrete at joints between lifts were as described in paragraph 3.3.3.5.5.2 (attached).
' i y l j
- l 4 0 6
}
4 4
4 I
1
- I 1
1 l
1 0 s 6 l
I i
l 2-12
- w.,.
cm=
tr~ i G'
Excerpts fre: Trojan FSAR Ter :he Cen:ainmen: vall, :he inside for was fixed, made up of the a:iffened linar pla:es.
For o:her stru::ures conven:i:nal f:rning 't:s pr:vided , in :::crdsn:a with che previsions of AC:-347.
During :he pr:gress of :he werk, cen:re:e was sampled and tas:cd far slump, .
air c:n=en:, tempera:ure, uni: weight, and cr pressive streng:h as described in Sectien 3.3.3.3.1.7. Cencrete slump and compressien :es: samples were taken a: the batch plan:. Pariodic correlation samples were taken at the point of placement.
3.8.3.5.6.2 Sonding of Concrs:s Between Lif:s Mori: ental cons: rue:1:n joints were prepared for rocciving the nex: lif:
by sandblas:in;, b-, cu::ing wi:h an air-wa:er je:, er by bush hammering.
Surface wet retardan: compounds were not used. When wet sandblasting I was employed, it was centinued until all laicance, coating, stains, debris, 4 I
and o:her foreign esterials were recoved. The' surface of che concrete , l was washed thoroughly :o remove all loose sacerial. .
When air-va:er cu::1ng was used, it was performed af:er initial se: had j taken place bu: before the :ncrete had :sken its final se:. The sur- (
\
f ace was cut w1:h a high-pressure air-wa:er j et to remove all laitance l s
and to expose clean, sound aggregate. but not to undercu: the edges of the larger particles of aggrega:e. Af:er cu::in;, :he surface l
was washed and rinced as 1:ng as :here was any trace of cloudiness of the wash wa:er. Where necessary to ra cve accuculated lai:ance, coa:ings, s:ains, debris, and other foreign 22:erial, wet sandblas:ing was used before placing :he nex: lift, to supplement air-water cu::ing.
Hori: ental surfaces were we::ed and covered wi:h approxt:acely 1/4 in.
of mor:ar of :he same cemen:-sand racia as used in :he a:ncre:e 12:ed-iately before the c:ncre:e va: placad. Vertical join:s vera sandblas:2d or. bush ha==ered, cleaned, and wa::ed bef ore placing concrete.
3.8-151
O
_'\ )
Excerpts from Trojan FSAR The above-described pr:cedure provided bend be:veen the old and new con:rece equivalen: in :ansile and shear strength :: :ha: Of concrece placed nonoli:hicly.
3.8.3.5.6.3 In:erior .Ceatings for Nuclear Service The liner plate enpesed to the C:ntainten: in:erior was primed wi:h in-
- organic zine primer, carbo:ine 11. The final finish en the liner place v
consistad of 5 mils of epoxy - phenolic finish generally to a heigh: of 6f above :he floors and 2 :o 3 mils inorganic tcpecat above that.
The exposed strue: ural s: eel insida :he Con:ain=en: was pri=ed'wi:h organic :inc. The coatings were subjected to tests designed to ensure
,that the coa:in; would suffer no significant less of adhesion er ieceriors-tion that could centribu:e particles capable of in:crference with the free flow of the Emergency Core Cooling System. -
The criteria for the tests were as follows:
- 1) Test solution: Demineralized 'w'acer 0.23 molar H330)~ ~
(3000 p;= 3eron) (U . S . Borax, Scecial Quality Grade)
Add Reagent Grade NaCH to make pH 9.5.
- 2) Test Temperature-Pressure: Expose to the following ti=c-ce:perature-precsure profile as accurately as possible:
Temperature Time 'F Pressur(=}
psig 3 120 0 O
1 see to 30 min 300 70 30 =in to 96 hr 250 30 4 days to 100 days 200 10 Record ac: al test temperature and pressure versus :ine.
(a) Sa:uratad s: cam that will yield p:essures of approxi=ately 120 psig at 300*7 is accap:able. Ambien: pressures =ay be used at 200*F for 4 :hrough 100 days.
Amendment 1 3.8-152
(.'! arch 19 7 3 )
l __ _
( .
- 3. S!! CAR WALL REIMFORCI"G I
The wall reinforcing between clavation: 45' and 77' is as fo)1cws:
- 1) Block vertical reinforcing is a d6 0 24" each face in all shear walls except the North wall- between elevations 45' and 65' . At this loca tion ,
a f5 0 24" was used on each face.
The vertical steel is either continuous or embedded in the slab at the bottom of the wall or embedded at the top of the wall.
- 2) . Core vertical reinforcing is a #6 0 24" in all walls between elevations 45' to 61' and 65'. The reinforcing is continuous in the West and
. South walls and laps with .8" shear studs on'the bottcm flange of the steel beams at elevations 61' and 6'5' in the East and North walls -
respectively. At elevations 65' and 77', the core reinforcing is a
!S 0 24" in the South' and West walls and is continuous.
- 3) ,
Block' horizontal reinforcing is 2-!5 0 24" in each face of all shear l
walls between elevations 45' and 77', except the North wall. The l 1
reinforcing in the M:rth. wall is a i!6 0 24" in each face. The !
reinforcing in the block on the out:ide of all walls is continuous.
'The reinforcing in the inside block is cont'inuous with the following exceptions:
1 4
. . l l
3-1 i
,n -
\
(./ .
a) Both bars are discon:inuous in the South and liest walls between elevations 40' r.rd Gl' and in the '! cst wall at elevations 61' to 77'.
b) On the East wall, only one #5 of :the two is centinuous on the inside between elevations 45' and 77'.
- 4) Core horizontal reinforcing varies in size and spacing. The reinforcing is continuous in the ',lest wall between elevations 45' and 61' ena in the South wall between elevations 61# and 77'4 The reinforcing at
'these locations is a ill 018" and a #110 9" respectively.
0 0
0 4 4
9 e
B 4
e 4
e 4 0 3-2
O
- 4. CETERMINATION OF SliEAR CAPACITY BY CO',iEL ACTIC:l C JLY The structure was also conservatively evaluated by ccmpletaly disregarding the concret2 shear strength, VC , which has been empirically developed by testing. Only dowel action of the reinforcing steel and the embedded steel columns is considered. The concrete was only used in bearing to transfer the dowel forces. The safety margins based on strength are given in Table 4-1. The most critical location was examined to determine how much of the available energy was needed to resist the load. Based on the required strength, about 12% of the available energy was required.
This entire investigation is quite conservative since the loads were based on elastic analysis which results in upper limit loading with low dampi ng. In reality, if this dowel action mechanism were required, then the structure would be nonlinear, higher damping would result, and also the responsc would be lower than predicted by elastic analysis .with low da:npi ng.
In conclusion, this investigation has shown that there is sufficient-capaci'ty to resist the Safe Shutdown Earth:;uake by dowel action with no reliance upon the concrete shear capacity.
l .
4-1
4.1 MATERIAL PROPERTIES The following material properties have been used in the investigation.
Structural 'rengths have been based on the ASTM A-35 specification:
J l
Yield = 36, Ultimata = 5.
Reinforcing stes ngths have been based on the mill certificates:
l
- 5 reinforcing: l Ultimate = 76,000 psi (average) /
Ultimate = 73,200 psi (low)
- 6 reinforcing:
Ultimate = 30,900 psi (average) l Ultimate = 75,7C0 psi (1cw) 911 reinforcing:
Ultimate = 76,900 psi (average) i l
Ultimate = 73,600 psi (low) 1 i
e 4-2 1
O ,
t 4.2. LOADS The loads for the cesign conditions are given in Tcble 4-1.
1 TADIe d-1 LOADS (KIPS)
Elevation 45' & Above To Below Elevation 61' North-South East-West 1 i
13,500 14,100 1.4Eo 0 a = 2%
9,890 10,500 1 1.0Es 0 s = 5%
' Elevation 61' & Above To Below Elevation 77' North-Scuth - East-West.
g = 2% 10,200 11,900 l 1.4Eo 0 1.0E3 0 s =.5% 7,580 8,880 ,
i Required Capacity Using SRSS And 13". '!eight Reduction:
e i
1
(
i l
-l 4-3
\
4
- . . ~ . .
. (~'s, - . . . . . . . . . . . - - . . . . . - . . - . . . . ... - . - . -.
l l
U 1 l
l l
l 4.3 00'.lEL SHE??. CAPACITY CT, ITER:A This section defines the criteria that was used to evaluate the 1
ultimdte strength of the structural members and the reinforcing in dowel action shear.
1 l
4.3.1 Structural Steel The capacity was based on the icwer value of either concrete ..
bearing or steel shear strength. ,
4.3.1.1 Bearing The capacity was limited to a bearing value of fj using a . triangular bearing distribution as shcwn below.
1 f, Fc =f[(W)(2W)7 c
F *W p c c
/
m where:
f[=compressivestrengthof concrete R
W = width of member in the
" direction of loading 4-4
U 4.3.1.2 Steel Shear Strength a) For the loading parallel to the web:
Fs " Ati(~}(#u 3
11
]
ll i t
Load ,% ; ,- ;
a H vihere:
p sa A,,, = vie b a re a
. eu= ASTM minimum ul timate tensile 1
strength b) For the loading parallel to the flanges:
F s = 2Ap ( 2) f *u)
Load >- ] "
vihere: .
A = single flange area i f
au= ASTM minimum ultimate tensile i i
strength l l
l O
l l
l l
e a
- l l 4-s l
l l
. r 4.3.1.3 Effective Columns The figure Th7,an b21 w indicates t.:iich col =ns were -
'~ '
considered effcctive. ,
~
Of rection Of Lead n
MMy i D. fi' .
This out:ide.c:le.n Inta al eclu n: k This cutsido colt =n considered as consicered as not considered.
parttally effective fully effective ar.d only 50% of and full strength., ~
strcncth c:cd. j
.(See Appendix 4A) 4.3.2 Rei,nforcing Steel i
The reinforcing st:-el capacity was bh. sed en the following:
F3 = A3 (.7)(ay) .
where: .
As = area f reinforcing 'bar c9a average ultirate tensile strength for bar si:e to e
se
. 6 4
4-6
O v
~
4.3.3 Miscellaneous
- 1) The determination of capacity has relied upon the side walls resisting all the applied shear loads and the end wall resisting the gross applied bending mcment.
- 2) The lap and embedment length of the reinforcing steel was in l accordance with ACI 318-63, and vertical lap splices were 1
made at the same elevation.
- 3) The embedment length for shear transfer is icwer than that l for tensile strength transfer. Nelson studs (Reference 1) .
l typically have a depth equal to 5 to 8 times the diameter and develop full ' shear strength which results in a stud failure. Studs typically have a strength equal to 90~,
95% of the tensile strength.
- 4) Tests are presented in Reference 2 in which flat bars are leaded 2
to failure, ranging in size frcm 1 to 4 in . For square bars, the embedment length was 6 times the width of the member.
Load-displacement curves for these tests indicated a propor-tional limit between .04" to .09" and ul timate di spl acements were equal to or greater than .28". The shape of the load displacement curves was very similar to that obtained frem testing Nelson studs. Therefore, it is concluded that the curve shown in Appendix A is typical for shear embedments 2
ranging in si:e from very small diameters up to 4 in in area.
4-7
L) f*
J The shear strength reduction factor of .7 was based on the test data for square specimens (1 in 2) from Reference 2.
- 6) For all the columns which were considered effective in fosisting shear, the column is fully embedded in the wall.
All reinforcing steel which was considered effective satisfies the embedment length requirements for the calculated shear transfer.
l l
l b
l l
S 4-8 1
. .- ~ . . .
D APPENDIX CAPACITY CF SHEAR EMBEC.tiEM'TS WITH OME-SICE CONTACT i
- 1. GENEP.AL The folicwing documents testing of liner plate anchors embedded in concrete blocks. The basic tests and results are documented in Bechtel Topical Report BC-TOF-1, " Containment Building Liner Plate Design Report", Revision 1, December 1972.
- 2. PILOT TEST Prior to running the tests documented in .BC-TOP-1, a pilot test was performed on a block of concrete which is shown in Figure 1. The concrete block was 12" wide, 8" deep and 15" high. The plate was 1/4" thick and the angle was 1/4"x2"x3". The angle was welded to l l
the plate using a (3/16") A"-12" stitch weld en both sides. The block I was restrained frcm rotating by having hold-down bolts en the side opposite the load. Shortly af ter the application of load, concrete cracking' developed as shown in Figure 2. The load was increased and
. the final capacity of the ancher was about 2,5 kips /in. In this te.st a member was tested with only one-sided embedment or contact.
i l
l 4A-1
. 9 $ *W
i
/~'\ i U
l
- 3. FINAL TESTI:!G
)
Based on the results of the pilot test pregram, the tcst setup was modi fied as shown in Figure 3. Steel plates were added to the top and bottcm with bolts to prevent the premature cracking of the concrete bicck. These tests are documented in SC-TCP-1. Other then the modification previcusly described, the plate, angle and welding was the same for Specimen I. Specimens II, III and IV had different wold and angle orientations. Specimen I, which had the same orientation of angle and weld as the pilot test, failed at 4.94 kips /in.
- 4. CONCLUSION Based on the pilot test and the final test of Specimen I, it can be concluded that the strength of a shear embedmont with only one-sided contact has about 50's the capacity of an embedment which has two-i sided contact.
1 l
)
r l
( .
-. I
? . ,
1 l
4 J
i l L :d 3 1
- - \' ,r ,
a
~
Crack 1 l
m p \
n ,1 >
TT .1
,r')
h tatien N g ,
j r.:. : : n 1 , t - - L-- r -
.k :I %, :/ .r.' s l
6 t ) s- ncta s. .
i
% u.
- c : -
-I M Angle '
- a *. g 3
, . s, ,
- . .. c i
i
- - ; mv < . ;n;;- -
- m w, ,-m n - ;
1 Figure.1 Pil *. Test -
Figure 2 Pilot Test 1 1
l l
4 .
Load i 1
l 4
.~ . .
mA min' I. .
Eleck a. ..I* l C!::pinD . 4 .'.7 ~ .
Ba s 1 j .
. t .
. s. .
. s .3 .
%..a i
il rD, b 1 Rat.:tien kut**0 h10% ,j. ,u.
I ei i
,,---/,u,,,...,..,,
i Ffgre 3 Final Test
- l
'4A-3 l
,f- .
(
9
- 5. CCMPAR! SON CF LCADS AND CAPACITY 3Y CO'.-!EL ACT 0N CNLY Cased cn tua conservative critoria stated in Section 4, th shear capacity was determined at various elevations in the structure. This capacity was only based on the dowel strength of the reinforcing steel and the col umns . Tacle 5-1 summari:es die capacity from ooth the reinforcing steel and the structural steel columns. The margin was cetermined by div'iding the total capacity by the .25g SSE load given. in Table 4-1.
The minimum margin based on strength is 1.4 or the structure is loaded to about 70% of its capacity. The CBE that can be resisted with a load factor of 1.4 and a margin of 1.4 is as fo11cws:
s CBE = capacity + 1.4 (.15g) loao o (.159)
For elevation 61' in the North-South direction:
i OBE = (10,600 + 1.4) (.15g) = .11g gggg These margins based on strength are really not representative of the margin that actually exists since seismic induced loads are not static loads. Many structures rely upon inelastic energy absorption to resist seismic motion. Usually, s, hen considering inelastic seismic response, the structure is idealized as an elasto-plastic system and the ductility ratio is determined 'ahich is required to resist the motion. This dowel action mechanism would result in a load displacement curve similar to the one given in Appendix 5-A.
5-1
I i
Since the curve is nonlinear and not elasto-plastic, the ductility l l
required cannot be evalua ted. The total area under the curve !
represents the total energy that can be resisted. At a particul ar. 1 load level or required capacity, the area under the curve is the l
amount of energy used. Table 5-A-1 shows the energy used as a .l l
percentage of the total energy available for various conditions. 1 For the condition of loading to 705 of ultimate,12", of the available energy is used and, based on energy, a margin of about i 1
8 exists. This value is more indicative of the safety margin than a simple strergtn ratio.
l l
1 l
l I
1 4
4 9
6 j 5-2
i 5.1 STRUCTURAL DISPLACEME!!TS Based on the ccpacities required to resist the .25g SSE and the load displacement curve in Appendix 5-A, the interstory and gross displace-ment can be estimated. The following assumes that cracking and I
slip will only occur at the major construction joints. From elevations 45' to 75', considering the four construction joints results in a displacement of .24". The upper stories will have lower displacements due to reduced loads. Increasing the .24" displacement by two and rounding off results in 1/2" as an estimate of the displacement at the top of the structure. Since there is a 3" clearance between buildings, the 1/2" is considered acceptable.
The maximum interstory relative displacement occurs between elevations 61' to 75' and is .15". This value is quite low and -
an examination of the equipment and piping has verified that this displacement should not cause operational or structural failure of l
any of the systems. Table 5-1 shows the maximum interstory displace-cents for the varicus floor elevations.
I 1
I l
l I
~
l I
5-3 i
- 1
TABLE'5-1 REI?tFORCI' 3 00', EL A :D
=
COLUM:1 SHEAR CAPACITY (KIPS) ,
1 i
Rainforcing Col =,n Elevation fTotal ::rgin
]
3 45' 18,000 2,200 20,2C0 0.C5 i
59' 13,100 2,100 15,200 1. 54
.c 61' 8,200 2,400 10.600 1.40 65' 10,000 800* 10.B00 1.42
. S 70' 10,000 li400 11,400 1.50 75' 9,600 1.400 11,Ct0 1.45
. 45' 14,300 ,
13,200 27.600 2. 63 j 4
l 59' 10,500 13,3C0 23,800 2.f3
- 1. -
i 3 Gl' 12,800 10,700 23,500 2.65 l
, .o 5,600
}0 65' 11,200 4,700 16,800 15,900 1.89 1,79 70' 11,200
. 75' 11,200 8.600 19,200 2.23 ,
l Note:' Margin is based on dividing the available capacity by the .25; SSE loading given in Table 4-1.
- Lower column strength results frcm column splice.
4 5-4
l
'i f > \
q .
TABLE 5-2 MAXII'.Ul1 I;;TERSTORY 1
DISPLACEMElli (Ill) !
l
. Ocyatien I ibrgin I
V/V u Dis pl acc.7/;nt 45' 2.05 .49 .0J"
> 59' 1.54 .65 .05" Gl' l.40 .71 .08" 75' l.45 .69 .07" 0
1 1
l i
l l
)
l l
l i
l I
. 1 l
so '
t 9
5-5 l l
., , . _ . _ . . . , , . . , __..-.....__.__,.m-.
._-_ ej , , .-
f%
V APPEUDIX 5-A 1
LOAD-DISPLACE.':ENT CHARACTER!STICS FOR 3/4" HEACED NELSON STUDS Figure 5-A-1 shows the shear load versus slip curve for a 3/4" headed Helsor stud. The ultimate shear load, Yu, is 32.5 kips with a slip of .3?".
l I
Figure 5-A-2 shows the same curve as in Figure 5-A-1 with the shear load normalized to V/Vu = 1. Figure 5-A-2 is used to determine the relation-chip between the applied shear load as a percentage of ultimate and the percentage of available energy used. l Table 5-A-1 illustrates this load-energy re'lationship. Cumul ative l percentages of energy used are shcwn for six load points:
V/Vu = .5, .6, .7, .8, .9, and 1.0. ,
e e
n 5h-1 l
4 d
1 5 35 1
- Vy = 32.5 ,
~ ~ -
30 (
25
/
. /._
/ i
/ --
{ 20 j
z i .
- 15 4
[ '
. 9 B
,f i m
- y u
,0 8 -;l' -
- I i 3 1 _
. 5 .
f 0"
. Os .05 .10 .15 .20 .25 . 30 .35
- Slip (in.) .
1 FIGURE 5-A-1
- LOAD-SLIP CURVE FOR 3/4" HEADED NELSON STUD
- e 5-A-2 99
l (}
4 m-4 1.0 .
- h ,
h 7 f f
/l \
t
.. /
/ -
_t l.
/
/ I l- .
~J p,
i .5 J u
- ) . l t
.1-i I
O!
s 0 .05 ,10 .15 .20 25 .30 .35 Slip (in.)
FIGURE 5-A-2 NOF.!'ALIZED LCAO-SLIP CURVE FOR 3/4" HEACED l'!LSC'! S~L'D TABLE S-A-1 SitEAR LOAD - OtERCY USED V/V u .5 .6 .7 .8 .9 1.0 Cum. % 2.2 5.2 12.1 . 25.1 52.6 W, Of SA' 1
B 4
( 5-A ,
's ,
_q V.
s
- 6. CCHTRCL GUILDING TOR:lACD RESISTAhCE The Centrol Puilcing has been checked to catermine if it is adequata to resist'a desiin basis ternado as specified in the FSAR.
Tne gross bending movement and shear due to velocity pressure are only about 250 of the loading frem the 0.25g SSE, and tne structure has adequate capacity to resist these loads.
The walls can resist the local bending and shear due to the differential pressure from external pressure drop.
Based on the modified Petri formula, the walls have sufficient thickness 4
to prevent perforation and spalling when impacted by the FSAR specified tornado missiles.
V e
0 4
6-1 .
- -- - -- =
s REFERE!!CES
- 1) Desicn 0:1ta, !!elson Cencrete Ancher. Studs fnr Securinn Steel to Concrete, Ccpyright 1961, Gregory Industries, Inc., Lorain, OH.
- 2) " Containment Liner Plate Anchors and Steel Embedments Tcs:
Re sul t s" . P. L. Chang-Lo, et al., 4th International Conference of Structural 14echcnics in Reactor Techno1cgy, San Francisco, CA, August 19, 1977.
O
~
l e
4 4
O REFERENCE 3 TO LICENSEES' TESTIMONY OF RICHARD C. ANDERSON, GEORGE KATANICS, THE0DORE E. JOHNSON AND WILLIAM H. WHITE ON CAPABILITY OF TROJAN NUCLEAR PLANT TO WITHSTAND SEISMIC EVENTS
. ()
i e
i i
i TROJAN CONTROL BUILDItlG SUPPLEMENTAL STRUCTURAL EVALUATION SEPTEM8ER 19, 1978 i
i l
l I
d I
I l
f} ..
i TABLE OF CONTENTS
- 1. INTRODUCTION 2..
SUMMARY
AND CONCLUSIONS
- 3. SEISMIC ANALYSIS AND LOAD DETERMINATION
- 4. CAPACITY DETERMINATION
- 5. COMPARISON OF LOADS AND CAPACITIES
- 6. MOMENT AND VERTICAL SHEAR TRANSFER
- 7. REDISTRIBUTION OF FORCES
- 8. TRANSFER OF SHEAR FORCES TO ROCK FOUNDATION Appendix A - Comparison Of Seismic Analyses Appendix B - Shear Capacity Criteria Appendix C - Strain Compatibility .
Appendix D - Displacement Determination Appendix E - Inelastic Behavior i
(
/
LIST OF FIGURES FIGURE NUMBER TITLE 3-1 "STARDYNE" Model 3-2 Wall Key Plan For Elevation 45'-61' 3-3 Wall Key Plan For Elevation 61'-77' 3-4 Fundamental Mode Shape At Top Of Structure, North-South Direction, Plan At Elevation 117' (Not To Scale) 3-5 Fundamental Mode Shape At Top Of Structure, East-Wes t Direction, Plan At Elevation 117' (Not To Scale) 4-1 Shear Wall Capacity Criteria 8-1 Foundation System 11
f-Q) r LIST OF TABLES TABLE NUMBER TITLE 3-1 Shear Forces, N-S Motion, Elev. 45' , 8/17/78, SSE = 0.25g, 8 = 5%, Flexible Base 3-2 Shear Forces, E-W Motion, Elev. 45 ' , 8/17/78, SSE = 0. 25g, 8 = 5%, Flexible Base 3-3 Shear Forces, N-S Motion, Elev. 61', 8/17/78, SSE = 0.259 ,
8 = 5%, Flexible Base 3-4 Shear Forces, E-W Motion, Elev. 61 ' , 8/17/78, SSE = 0. 25g, 8 = 5%, Flexible Base 3-5 Shear Forces, N-S Motion, Elev. 45 ' , 8/ 24/78, SSE = 0. 25g, 8 = 5%, Fixed Base 3-6 Shear Forces, E-W Motion, Elev. 45' , 8/24/78, SSE = 0. 25g, 8 = 5%, Fixed Base 3-7 Shear Forces, N-S Motion, Elev. 61 ', 8/24/78, SSE = 0.25g, 8 = 5%, Fixed Base 3-8 Shear Forces, E-W Motion, Elev. 61 ' , 8/24/78, SSE = 0.25g ,
8 = 5%, Fixed Base 3-9 Dynamic Characteristics - Fixed Base 4-1 Capacities In N-S And E-W Directions (Elevation 45'-61')
4-2 Capacities In N-S And E-W Directions (Elevation 61'-77')
/
iii
73 .
V TABLE NUMBER TITLE 5-1 Force-Capacity Comparison, N-S Motion, Elevation 45'-61',
Fixed Base, SSE = 0.25g, s = 5%
5-2 Force-Capacity Comparison, E-W Motion, Elevation 45'-61',
Fixed Base, SSE = 0.25g, s = 5%
5-3 Force-Capacity Comparison, N-S Motion, Elevation 61'-77' ,
Fixed Base, SSE = 0.25g, s = 5%
5-4 Force-Capacity Comparison, E-W Motion, Elevation 61'-77',
Fixed flase, SSE = 0.25g, s = 5%
6-1 Gross Moment Tension Forces For N-5, SSE = 0.25g, s = 5%,
Elevation 45' And 61' 6-2 Vertical Shear Forces And Capacity For N-S, SSE = 0.25g, s = 5%
7-1 Base Case Shear Forces, N-S Motion, Elevation 45' 7-2 Base Case Shear Forces, N-S Motion, Elevation 61' 7-3 Case 1 Shear Forces, N-S Motion, Elevation 45', Based On Limited Capacity Of Wall 1 From Elevation 45' To 61' 7-4 Case 1 Shear Forces, N-S Motion, Elevation 61', Based On Limited Capacity Of Wall 1 From Elevation 45' To 61' 7-5 Case 2 Shear Forces, N-S Motion, Elevation 45', Based On Limited Capacity Of Wall 3 From Elevation 61' To 77' 7-6 Case 2 Shear Forces, N-S Motion, Elevation 61', Based On Limited Capacity Of Wall 3 From Elevation 61' To 77' l
1 l
iv
~. . . . . - - _.
., ([
i
. TABLE-NUMBER TITLE i
- 7-7 Case 4 Shear Forces, N-S Motion, Elevation 45', Based On Limited Capacities From Table 4-1
- 7-8 Case 4 Shear Forces, N-S Motion, Elevation 61', Based On Capacities From Table 4-2 7-9 Case 5 Shear Forces, N-S Motion, Elevation 45', Based On
- Limited Capacity Of Wall 1 From Elevation 45'-61' 7-10 Case 5 Shear Forces, N-S Motion, Elevation 61', Based On Limited Capacity Of Walls From Elevation 45'-61'
! 7-11 Summary Of Cases. Elevation 45'-61' 7-12 Summary Of Cases, Elevation 61'-77'
! 7-13 Elastic Displacements 4
! 8-1 Sliding Resistance And Base Shear i.
i t ,
1 j
i 1
i 6
1 ,
4 e
9 v
. (I ~
9/19/78 4
- 1. INTRODUCTION
- The original seismic analysis of the Trojan Control-Auxiliary-Fuel Building complex was completed by Bechtel in May 1971. This analysis was updated by ,
Bechtel in April 1978, when the weight of the Control Building was re-estimated based on the as-built condition and responses were re-calculated using the' square-root-of-the-sum-of-the-squares (SRSS) technique. In June 1978, an analysis performed by an independent consultant using the TABS program verified that the prior Bechtel analyses were. conservative. More recently, 4
in August 1978, additional analyses were performed by Bechtel using the j finite element STARDYNE program to predict seismic loads. This report 'provides 4
j additional information, analyses results and conclusions associated with these t
finite element analyses.
l 4
I 1
i
}
W 4
1-1
9/19/78 j
- 2.
SUMMARY
'AND CONCLUSIONS Four seismic analyses have been performed on the Trojan Control-Auxiliary-Fuel Building complex: 1) the original spectral response analysis using a stick model, in 1971 (re-evaluated and modified by application of SRSS and use of estimated as-built weights, in April 1978); 2) an analysis by an independent consultant utilizing the TABS program, in June 1978; 3) an analysis utilizing a finite element model and the STARDYNE program with flexible-base, in August 1978; and 4) an analysis utilizing a finite element model and the STAROYNE program with fixed-base, in_ August 1978. These analyses are described j
and compared in Appendix A to this report.
Section 3 of the report provides information on loads predicted by the STARDYNE analysis. This ana. lysis is highly sophisticated and comprehensive; walls and slabs are modeled by finite elements. In some walls, the loads predicted by STAROYNE are higher than previous analyses predicted. Due to the use of elastic I finite element analysis, the loads predicted by STAROYr E are upper limits.
l Section 4 provides criteria for the shear wall capacities used to evaluate the capability of the Control Building to resist loads predicted by the STARDYNE analysis. These capacities are higher than those used in the previous evaluations.
Justification for use of higher capacities is provided in Section 4 and Appendix B, along with the developments of the criteria employed to determine the capacities and their correlation with tests. Strain compatibility of the shear walls is discussed in Appendix C.
2-1
HO:
4 Section 5 of the report compares the STARDYNE-predicted loads with the shear wall capacities. Only in a limited number of small walls does the load demand exceed the capacity. The contribution of those walls to the capability of the -
I total structure to resist seismic forces is very small.
Section 6 describes the capability of the Control Building to resist gross bending moments and the transfer of shear fran side walls to end walls. The gross bending moments are primarily resisted by the end walls perpendicular to the direction of earthquake motion. Sufficient vertical shear capacities exist to permit transfer of load from side walls to end walls.
4 Section 7 documents a detailed investigation which distributes seismic loads as a function of both stiffness and capacity. The investigation demonstrates that the Control Building can accommodate the seismic loads, even with small walls exceeding yield, due to multiple shear walls which have a total capacity higher than the total load.
Section 8 describes the load transfer mechanism from the Control Building to the foundation rock and demonstrates that a cufficient factor of safety exists against sliding.
On the basis of the investigation and analyses described in this report, it is concluded that the Trojan Control Building can resist, with a margin of safety, the forces due to a 0.259 SSE with 5% damping. Since the seismic forces resulting from a 0.25g SSE with 5% damping and a O.11g.0BE with 2% damping are the same, it follows that the 0.11g OBE for the Trojan' Control Building continues to be appropriate.
lI i
i 2-2.
- 3. SEISMIC ANALYSIS AND LOAD DETERMINATION Additional seismic analyses have been perfonned by Bechtel using the STARDYNE finite element computer program. The first analysis simulated foundation flexibility and the second has used a fixed base condition. The analyses were based on the following:
- 1) The entire complex of Control, Auxiliary, and Fuel Buildings was modeled.
The model considered all significant wall and floor slabs which were simulated by plate finite elements. Access openings were modeled but not pipe penetrations and other small openings. The stiffness of the elements was based on elastic properties. The model is shown in Figure 3-1.
2)
The stiffness of the foundation for the flexible base case was based on the techniques presented in BC-TOP-4 A [Ref.1). A shear wave velocity of 5500 fps, as defined in the FSAR, was used for the foundation rock.
The foundation flexibility was orily considered under the Control Building.
Radiation- damping conservatively was not considered.
- 3) The .25g SSE response spectra with 57, damping presented in the Trojan FSAR was used in the analyses.
- 4) The seismic analyses were performed by a linearly elastic modal analysis spectrum response technique, utilizing the connercially available STARDYNE computer program. The modal responses were combined using the square root of the sum of the squares technique (SRSS).
3-1
- . . _ - - ~ . , , ,
A Key Plan for elevation 45'-61' is shown on Figure 3-2 and the Key Plan for elevation 61'-77' is shown on Figure 3-3. These Key Plans are for reference and do'not show all openings.
Tables 3-1 to 3-4 show the shear forcesat elevations 45' and 61' for both the i N-S and E-W directions for the flexible base case. Tables 3-5 to 3-8 show the shear forces at elevations 45' and 61' for both the N-S and E-W directions for the fixed base case.
A comparison of these two sets of tables indicates only a 3% difference in gross response between the flexible base and fixed base cases. Since the foundation shear wave velocity is high, the fixed base case is an appropriate representation of the response, and evaluations given in this report are based on this case.
The mode shapes for the fundamental and most dominant modes are shown in '
Figures 3-4 and 3-5 for the top of the structural complex. The most significant i modes are documented in Table 3-9 which shows frequency, participation factor and modal effective weights.
REFERENCE
[1] Bechtel Topical Report BC-TOP-4A, " Seismic Analysis of Structures and Equipment for Nuclear Power Plants", Bechtel Power Corporation, San
~
Francisco, CA, Revision 3, November 1974.
3-2
, [.._,.y - -'
i
()\
l x
t i 3 l
l t
i flit liUILDING G) Nil 3)L BUILDING .
AUXILIARY BullDialG .
[
.. t- : :
\.- , 'f. 4 ~~~~~~~. -
.t- \ - ,
ls' . -
hI , -; l -
t
.__s.s x/- \ - -~ . 1. .
l s_- .-
.w
.s
- N' ;; _ -,-- I ! :
. . . . . . - ~ e.
I . .
< 'l -
.l . ,I t g.
,:- g J_,. .- .
sQ,7 ......-
,,:,..3.-
.._ - i _,-s,
_ ;- g-l-
'- - t .g .
l s v N'i '
-x..v- . -, m- ' 1 _
1- ,. -1 s, r _ . . ;-
.. '~~
, . - -y \(.-~ .- ._.' w -' 7- [- -- - " . .. 1 N- g7
. ;g
+:,' lg s 'i
?- .- ' I
-;j N'.g l k;;'9 i . 3 ~_ -
~~~
l ,
-q .
I
'~
1
~
.) [, _! . -' ;h-[, j., . ~ M:I
!. ,_ .-:: :- ~--2 _
M~s-... ~. grkI e '=^-e d l.y. ,- :-
, hJ~-
-c ~; * . .'~; - - ] K- [
s_-:
. st - _ I .l . .
s..v-
.s , N
- .L- t t
N._
k
__,. a .-i
- e. ...
l'Q ~
,e . c, f
/g - h-
.z.
x_ .. -i. } f _.'[-L.,-..
4:.ESt
-: ;.::- m.%
'N s.
'% h 1
i - :% 3._. -M.; 5: 1 p j' , .~l
- ': 22LN- ^-f. . ~~.
\ s, .
- 7. ' ; . ,' ,l -
-n -
p" p_,_. - O. _. .M~.- M. -- .M..
!s 4
A
_- 9
- i 3
-c 3- pt-I -"M.O. . ;" s,.s ss.- 5, s 5 '
-Q"* ,
5 3 .--
_- + . - .# ,.
. . . - .. _ s._.:~.--
s -
x, - _, _:-;i_, s
______.,;<__- r.- -
- __... -,a - .--
- -
s ,.'$
.l
' ss
.t -
_- g
,'S!
'~~i . ;4. q \'" ' .,
N- 6N -+ j.~.-
g, ,4 *-i'. I '
lE 9___ ,,h-- *-
's s;(-l ~_ql' . .
- I - .r;. - M .: . ,
v t
.fy
.s 1
_;- u iss 3
j5 3
' . 3._
g
~-
- l
\,,
st i _ i . ..
,s ,
N i '~ \
s s
- 'I
\ '
' I l
- \l
'- . - ' ~
[.
- -} e , N 1 -
. llGURL 3-1 "SIARDYl4E" PDDIL ,
t I ,
m & _ m _ _ _ _ _ _ _ _ . _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ . ._ _ _ _ _ _ _ _ _ _ _ _ _ __ u.- _ _ -.--.
I
@ l
@1
@ I 9@ @ 9 I i 1 l l I i l l l
g_ __
l.__ .q._._
I l l ! l
' .I _._.p.
f
~
l ! I l I ,
l ,
g _ {_._.
I
'.__. l 1 I
I l l
g_ _ j. _ . __ . _ _._____I _ _ _ _ _ . _. 2 1
I I
D, @
(- - . ,,
O g_-.__._ _ . . _ . _ . _ ._.
i
_Q._.._ < I 00- y r-0 @h :
.g~
hl I
& ._.l j W
_l l g__ @ Q ,
._._.i .. @
_;.' ._. . _ . l G_ l l
l o ._. _1 v
g __
.rf Figure 3-2 Wall Key Plan For Elevation 45-61 l
1 l
l
O -
i
' I - '
l i l l l l l l I i
@p_
i b
_.q._ _.__._l____._
I ! ! l l
_.__l.__ .-__._I.-.. -____L_
Qu_ . . . . . _.__
l i
i I l 1 -
g_i _ . _ _ .
l i g_ j . _ . _ . _ __ _ ._ _ _I ;
i
('7j _. L._ _._ . _ _ . _ _ . . _ _
p._ . _ . __ ._
Q.)_ _ '
7._
I
! i p@ '
gy_ p . __ . __ . _ . _ _i f _ _ . _
l
@ @ @' 9 g_ p. .__ _ . . _ _ .
" V. _ .
W9 i
(p/ __
0 ; i
~
^ , ,
- A@ i g_ _.__.____L._._.__ . _ _ . _ . -
h I l gy_ ,.-._.._.,1 l -
. I l g _. _ . __ . __ . _4 . __ . ___ . __ .h _ . _ . _
l l g_ v i i v h .. __ . __.
_@ h Figure 3-3 Wall Key Plan For Elevation 61-77
'()
jX2_bP '
__.y -5 -4 t
i I l g
i I g i N' \
g
. t \ . .\.
'f ~i f ~ ~ ]
i l 1 1 I I I 1 ,
l ' -r T~
't I l I g i 1 I g i l
,_- , m t -- - .4 f__I g i l
i l
i l
i I
g i i f = 6.8 cps
~
a- e n
T I i i l I !
X3 I I
', ', X1 l
_a e _. - 4-g l 1 l
l l I g l I I I t i
$ eL_.y._y:-y=-f=~h i
\
l i i 1 1
. \
l I I g i l i
- . i _
i 1 . I _. I _, I p _. __ e _ m. _. _ c "_ - *- ~ _ 4 - -Y I l 1 I I I ;
i 1 I l \ 1 ;
1 I l I I I g y _ -. p y e. - n - _ * ~ -x I i l l l l l 1 1 I I I I I I I I I I I I m 4.- c _ _.
_ ,. ,__.*% _. 4 I l I _. _ b l l l I I I I I i I I i i i i I I e _ _ e _ x. _ _. , ~_ _ ,_ ;_. _ + - -. x ,
Figure 3-4 Fundamental Mode Shape At Top Of Structure, North-South Direction Plan at Elevation 117' (NotToScale)
O l
a yt y--y--y- g
' I I I ; ;
r r e i i
l l x- - - u - - x - - x l l
_ _ yl l l l l l )
y t i i t l l l l l n .- ---p -j --q m . .
t i i l
p !
l l l l
>g.__A___.4 -s t s ,
i t t 1 f
n.
= 8.5 cps - L '. L I
, - - n - - :p
_ 1 i - >Itl 1ll ,
1 l i i I 43- - p - - v - d - - dl x1 !
il 11 11 0 y -
1 lI i i lt i
+--r--
--p-- L, --
L, J A, J
l 1 il I i l )
i Xi x; ;, :
pf$---h-- b- p i
- ,4 - 1 I I I I I ji ,i g i f-- f-- '--k-f __ _
i I I I t i
- g I g 23 %, 2, .,
.:I
- l m, . - -l *- - - x- - -l +i a
lI i.
i i i i i Ti, w I g i l i l i
, ,, ,, .I g
L-i k- - k_ - k, _ __,4., _
l 1
x ,
x x x : ,
Figure 3-5 Fundamental Mode Shape At Top Of Structure, East- West Direction, Plan At Elevation 117' (Not To Scale)
69 V
Table 3-1 Shear Forces, N-S Motion, Elev. 45',
8/17/78, SSE = 0.25g, 8 = 5%
Flexible Base SHEAR FORCE WALL NUMBER (KIPS) LOCATION 1 4380 N-S WALLS 2 510 3 440 4 2020 5 3550 6 390 7 600 8 320 l l
[=12210 9 1670 E-W WALLS 10 1280 11 140 12 130 13 1610 14 40 15 290
l
[N O
Table 3-2 Shear Forces E-W Motion, Elev. 45',
8/17/78, SSE = .25 9, 8 = 5% i Flexible Base !
SHEAR FORCES WALL NUMBER (KIPS) LOCATION 1 920 N-S WALL 2 180 l l
3 190 l 4 400 5 650 1
6 60 7 140 1 8 60 l
l l
9 1870 E-W WALL '
10 1820 ,
11 440 12 260 13 4700 14 48 0 15 930
[=10500 I
Table 3-3 Shear Forces, N-S Motion, Elev. 61',
8/17/78, SSE = 0.259 , s = 5%
Flexible Base SHEAR FORCE WALL NUMBER (KIPS) LOCATION 1 4170 N-S WALLS 2 330 3 3000 4 2130 5 510 6 660
[=10800 7 1530 E-W WALLS 8 1570 9 830 10 130 11 490 12 300 9
- . . - -- .. . . _. . , .~. -
{
i- t i
l Table. 3-4 ' Shear Forces, E-W Motion, Elev. 61',
i- 8/17/78, SSE = .259, 8 = 5%
Flexible Base a ,
I
-SHEAR. FORCE-WALL NUMBER (KIPS) LOCATION iV 1 1110 N-S WALLS 2 160
. 3 870 4 370 5 170 ,
6 200 d
7 1810 E-W WALLS 8 3560 9 .850 10 380 11 1000 12 1400 l =.9000 l
f A
.~~v , , -y. < ..mv- -
('T
%.)
Table 3-5 Shear Forces, N-S Motion, Elev. 45',
8/24/78. SSE = 0.25g, s = 5%
Fixed Base SHEAR FORCE WALL NUMBER (KIPS) LOCATION 1 411 0 N-S WALLS 2 780 d
3 560 l 4 2240 5 3050 6 340 7 540 8 290 1 1
[=11910 1
9 1540 E-W WALLS 10 970 11 110 12 130 ,
1 13 1260 14 30 15 240 l
n d
Table 3-6 Shear Forces, E-W Motion, Elev. 45',
8/24/78, SSE = 0.259 , s = 5%
Fixed Base SHEAR FORCE WALL NUMBER (KIPS) LOCATION 1 910 N-S WALLS 2 220 3 180 4 420 5 570 6 70 7 180 8 60 9 1700 E-W WALLS 10 1680 11 510 12 320 13 4620 14 450 15 870
, [=10150
g.
O s i
Table 3-7 Shear Forces, N-S fetion, Elev. 61',
8/24/78, SSE = 0.25g, 8 = 5%
Fixed Base SHEAR FORCE WALL NUMBER (KIPS) LOCATION 1 3910 N-S WALLS 2 560 3 3140 4 1910 5 470 6 600
{=10590 7- 1480 E-W WALLS 8 1450 g 9 650 10 130 11 440 12 220 y
w m. r. - w,,,, -,, ,-ca
l3 LJ Table 3-8 Shear Forces , E-W Motion, Elev, 61' ,
8/24/78, SSE = 0.259 , s = 5%
Fixed Base i
SHEAR FORCE WALL NUMBER (KIPS) LOCATION .
I 1080 N-S WALLS 2 170 3 750 1 4 31 0 ,
l 5 150 1 6 220 l
7 1670 E-W WALLS 8 3560 9 790 10 350 11 950 12 1310
[=8630 l
in
()
Table 3-9 Oynamic Characteristics - Fixed Base 1
EFFECTIVE ;
FREQUENCY PARTICIPATION WEIGHT l (CPS) FACTOR (KIPS) DIRECTION 6.81 1.68 31,100 N-S l
9.40 1.11 12,500 N-S l
11.80 .92 11,800 N-S 8.53 1.79 50,300 E-W 12.47 2.07 9,900 E-W
O i
i
- 4. CAPACITY DETERMINATION The allowable capacities in the codes are usually set anticipating a certain level of sophistication when determining the applied loads. Both the ACI and UBC codes have not significantly changed their shear provisions for several years. It appears that these codes have not considered that the user would be' applying techniques as sophisticated as an extensive finite element analysis. !
l Only recently have computer programs come into use which can consider the '
flexibility of an entire complex of walls and floor slabs and mathematically I
distribute the loads throughout the elastic system. l The code provisions for determining shear capacity in walls are based on walls which have a height sufficiently large when compared to the base dimension 1
so that 45* type diagonal tension cracks can develop, which essentially run to i both outer edges of the wall. In the case of the Trojan Control Building, many of the walls are quite short in height compared to their length and this situation cannot develop. Therefore, particularly as to these walls, the code provisions are extreniely conservative.
Recognizing the considerations described above, a set of criteria was developed to evaluate the capacities of the shear walls. The detailed develognent of these criteria is provided in Appendix B, and the strain compatibility of the masonry walls and the concrete core is illustrated in Appendix C. These criteria t
are given in Figure 4-1. .The wall capacities based on these criteria are given in Tables 4-1 and 4-2.
The tables also include comments as to which criteria governed the capacity. Af ter evaluating the capacities, it was found that none of the walls that were evaluated by criterion (b) were governed by it. They all had lower shear capacities limited by bending. The Key Plans given in 4-1
.-v- , . - . - - + * - -e-* r
l1 V
Figures 3-2 and 3-3 are applicable to these tables. The values given are the lowest values obtained by the criteria between the elevations stated.
I 1
l l
4-2
/ . i V ;
Figure 4-1 Shear Wall Capacity Criteria Basic Criteria:
.2 V
u = [261-84 h + af] WT - 5h5 1.0 V = [218-41 + 1.0
]WT 5h3 3.0 l
V
= [150-19 h + ] WT 3.0 $h5 4.0 a
V 4.0
= [75+ -[] WT 2h ph
- Pv 1 002; py 1 0007 and ph 1 0013 4
forhj.2useh=.2 ,
where:
Vu = allowable.shearforce(lbs) p h = horizontal reinforcement ratio H = wall heig'nt (in) o = vertical reinforcement ratio W = wall width (in) a" = axial stress (psii
(+ for compression; T = gross wall thickness (in) for tension) !
j l
In order to qualify for the basic criteria, the walls must be subjected to dead load and have core reinforcing steel. Walls which have structural steel columns completely interrupting the core shall use an f ratio which limits W to the distance between columns.
Additional Criteria:
a) Walls which are ' subjected to dead load, but for which no core reinforcing steel is assumed, may be evaluated by the basic criteria; however, the stress is limited to 150 psi.
._ . _. . . . . .- . _ _ _ _ _ . . - . .. -._ . _ _ .-._ _ ~_. . .. _. .
i
[ figure 4-1 Continued l b) Walls which are not' subjected to dead load, such as interior walls
! between floor slabs, shall be evaluated using the following shear :
criteria used in the May 5,1978 submittal:
i r
- Af V
=t(2yt,ff.+- {2)(.8W) 1 where: ,
& = .85 capacity reduction factor for shear f' = concrete cc.apressive strength (6000. psi) teff = total wall thickness -8" or t gross-8a i As = minimum of horizontal or vertical reinforcement (in /ft)
- f y= yiel'd strength of reinforcement (psi)
} Other terms have same definition as given under the basic criteria.
4 I
c) All walls must be checked for combined bending moment and dead load effects.
5
-w - rn 9 - - , me. r, .- , . - , . - - , . -
O' l
'i i
Table 4-1 Capacities In N-S And E-W Directions (Elevation 45'-61')
CAPACITY WALL (KIPS) CONTROLLING CRITERIA 1 5390 Basic criteria j 2 470 Bending Moment 3 490' 5
C 4 3810 Basic criteria S " "
5 5 5970 a
T x
6 11 0 Bending Moment 7 420 8 60 l=16720 i i
9 4730 Basic criteria I 10 5560 Basic criteria
. 1 11 240 Bending tioment
- 5 p 12 420 " "
8 5
o 13 9350 Basic criteria T
w 14 170 Bending. Moment 15 760
["21230 l
j
l3, .
V Table 4-2 Capacities In N-S And E-W Directions (Elevation 61'-77')
CAPACITY WALL (KIPS) CONTROLLING CONDITION 1 5100 150 psi limit 2 190 Bending Moment O
{ 3 4520 Basic criteria w
y 4 2240 Bending Moment 9
=
5 650 6 750
["13450 7 4440 150 psi limit 8 9340 Basic criteria
=
8 9 2820 t3 E 10 1380 Bending Moment c
2 11 2390 a "
12 2390 "
[ " 22760 l
l
- 5. COMPARISON OF LOADS AND CAPACITIES This section compares the f%d base case loads, as determined by the STARDYNE :
computer analysis described in Section 3, with the wall capacities detennined employing the criteria described in Section 4. The comparisons are given in Tables 5-1 through 5-4 for both directions at Elevations 45'-61' and 61'-77'.
The loads were developed based on elastic analysis which assumes that each wall i in the system has a yield strength higher than the load. When the load is higher than the capacity, then the load is fictitious and cannot develop further in that particular wall. The most important consideration is that the sum of all the wall capacities is greater than the sum of the applied loads.
I I
Table 5-1, for elevation 45'-61', and the North-South direction, shows a total I capacity 40% higher than the total applied load. Walls with an elastic load higher than capacity will yield and the excess load will be carried by other members. As indicated in Table 5,-1, only the small walls (Walls 2, 3, 6, 7 and l
- 8) have an elastic load greater than capacity and the combined small wall capacity is only 9% of the total capacity.
1 Table 5-3 for elevation 61'-77' and the North-South direction hews a tota l !
capacity 27% higher than the total applied load. In this case, only one small wall has an elastic load greater than capacity.
l For the Ent-West direction, as shown in Tables 5-2 and 5-4, the total capacity j is very high compared to the total load. There are some small walls with the i
elastic load greater than capacity, but again, they will only yield at their '
capacity and their effect on the system is essentially negligible. Based on the large excess capacity in the East-West direction, thic direction will not be 5-1
O l
considered fur ther in this evaluation.
Section 7 (Case 4) examines a more realistic distribution of forces in the N6rth-South direction, when the capacities are considered.
O e
5-2
O ;
l l
l l
I' Table 5-1 Force-Capacity Comparison, N-S Motion,
. Elevation 45'-61', Fixed Base, ;
SSE = 0.259, B = 5 SHEAR FORCE CAPACITY CAPACITY WALL NUMBER (KIPS)- (KIPS) LOAD l l
1 4110 5390 1.31 2 780 470 .60*
3 560 490 . 88
- 4 2240 3810 1.70
$ 5 3050 5970 1. 96 af w
6 340 110 .32*
e
'
- i 7 540 420 . 78
- 8 290 60 .21*
[ = 11910 [ =16720 1.40 9 1540 4730 10 970 5560 11 110 240 g .12 130 420
~
N 13 1260 9350 2
6 14 30 170 15 240 760
- Ratios less than 1.0 indicate the load is fictitious since the load cannot exceed the capacity.
.O i
Table 5-2 Force-Capacity Comparison, E-W Motion, <
Elevation 45'-61', Fixed Base, SSE = 0.25g, s = 5%
SHEAR FORCE CAPACITY CAPACITY WALL NUMBER (KIPS) (KIPS) LOAD 1 91 0 5390 ,
2 220 470 3 180 490
"} 4 420 3810 9E 5 570 5970 v7 3c 6 70 11 0 7 180 420 8 60 60 9 1700 4730 2.78 r
10 1680 , 5560 3.31 11 510 240 . 47
- 12 320 420 1. 31 m
gj 13 4620 9350 ' 2.02
- 2 2 14 450 170 .38*
a 15 870 760 .87* .
[=10150 [=21230 2.09
- Ratios less than 1.0 indicate that the load is fictitious since the load cannot exceed the capacity. -
l l
{
L 4
4 _ , - . . . . _ - , - , - ,
O Table 5-3 Force-Capacity Comparison, N-S Motion, Elevation 61'-77', Fixed Base SSE = 0.259 , 8 = 5%
SHEAR FORCE CAPACITY CAPACITY WALL NUMBER (KIPS) (KIPS) LOAD 1 3910 5100 1 30 2 560 190 . 34
- m 3 3140 4520 1.44 a
3 4 1910 2240 1.17 5 470 650 1.38 6 600 750 1.25
[=10590 [=13450 1.27 7 1480 4440 8 1450 9340
$ 9 650 2820 2 10 130 1380 11 440 2390 12 220 I 2390
- Ratios less than 1.0 indicate the load is fictiticus since the load cannot exceed the capacity.
(3 V
Table 5-4 Force-Capacity Comparison, E-W Motion, Elevation 61'-77', Fixed Base SSE = 0.25g, s = 5%
SHEAR FORCE CAPACITY CAPACITY WALL NUMBER (KIPS) (KIPS) LOAD 1 1080 5100 2 170 1 90 g 3 750 4520
!! 4 310 2240 m
d: 5 150 650 6 220 750 7 1670 4440 2.66 8 3560 9340 2.62 9 790 2820 3.57 m
gj 10 350 1380 3.94 3 11 950 2390 2.52 12 1310 2390 1.82
[=8630 [ = 22760 2.64 1
l l
l
. ~ - _
O l
l 6. H0 MENT AND VERTICAL SHEAR TRANSFER The gross bending moment is primarily resisted by the end walls (perpendicular to tne earthquake motion direction). Table 6-1 shows a compari. son of the ten:. ion in the end walls due to a North-South earthquake of .259 SSE at 5%
damping for tha fixed base case and the total dead load. The values shown are the totals for the walls from' Column Line N to R. As indicated in the table, there is no gross tension predicted in the end walls.
Sufficient vertical shear capacities are required so that the side walls can transmit load to the end walls and the end walls can resist the major portion of l the gross earthquake moment. The vertical shear capacities have been determined
.i by considering the combined strength of the structural steel beam-to-column connections and the dowel strength of the horizontal reinforcing steel. The connection capacity was based on twice the AISC, Part I allowable capacity. The reinforcing dowel strength was determined using 90% of the dowel ultimate stress times the cross-sectional area. This method of detennining capacity is ,
1 conservative since it neglects any transfer of shear by the concrete. A comparison
^
of total capacity and total applied: load from the fixed base North-South earthquake of .25g SSE at 5% damping is given in Table 6-2.
As shown on the table, the minimum margin is 1.55.
6 _ _ _ _ - a , - . .
Oc
\_)
Table 6-1 Gross Moment Tension Forces For N-S, SSE = .259 , S = 5%, Elevation 45' And 61' EARTHQUAKE TENSION DEAD LOAD WALL NUMBER FORCE (KIPS) (KIPS) m 9 3000 3450 j 13 4310 4560 7 2440 2470 s
j 8 2410 3330 1
.f.
-- .. . - - .. . - . - - . .- . = . - - .
O !
I Tabl'e 6-2 Vertical Shear Forces And Capacity For N-S, 'l
-SSE = .259, 8 = 5% j l
l CAPACITY- -
VERTICAL SHEAR REINFORCING CONNECTION TOTAL l LOCATION FORCE (KIPS) (KIPS) (KIPS) ( KIPS) MARGIN l l
R & 41 1910' 1430 1530 2960 1.55 N & 41 1030 1850 1080 2930 2.84 N & 55 2070 2980 2330. 5310 2.57 I R & 55. 2660 3650 1575 5225 1.96 l
l i
t t
> . . , , - . . . a .- . - _ , - .
- ( )-
- 7. REDISTRIBUTION OF FORCES
)
i I
7.1 General
{
As shown in Section 5, (Tables 5-1 and 5-3), the STARDYNE analysis predicts load demands in some walls higher than their ctracities. These loads are fictitious since the load developed cannot exceed the capacity. This section will examine in detail how dependent the overall system is on .the capacity of large walls and it will also examine the force distribution in the system after yielding of small walls. The analyses presented in this section are j very conservative since the applied upper limit loads based on linear elastic response are applied as static loads. When yielding occurs in a 1
system it becomes nonlinear and the response decreases. This is illustrated i i
in more detail in Appendix E.
I In order to study the redistribution ability of the Control Building, the following evaluations were made: I i ,
i
- 1) The effect of limiting individual large wall capacity and allowing other walls to develop the required capacity witnout limits, illustrating which other members would be most affected (Cases 1 and 2).
- 2) The effect of limiting vertical shear transfer (Case 3).
7-1 l
I
O :
- 3) ~The redistribution of loads within the entire system with limited capacities (Case 4). This is considered the most realistic case studied.
- 4) A final analysis, limiting the capacity of the' West wall (Wall 1), to illustrate that the Control Building's ability to resist the applied loads is primarily dependent upon having sufficient gross capacity and is not even significantly affected by single major walls (Case 5).
All analyses presented in this section are for the North-South direction with loads that simulate the .25g SSE at 5% damping. !
7.2 Analytical Techniques The redistributio.n analyses were performed by applying an equivalent set of static loads to the existing structural complex and determining the corresponding forces in the structural members. The equivalent set of static loads was obtained by applying a system of loads proportional to the first N-S mode inertial loads. The scale factor for these inertial loads was obtained by equating the base shear in th'e Western one-half of the structural complex (from Column Line H to R) to the corresponding bdse shear from the SRSS forces, which is 11,910 kips as shown in Table 3-5. The force in the various members due to the equivalent static loads is shown in Tables 7-1 and 7-2. These results are to be compared with those shown in Tables 3-5 and 3-7. The comparison shows excellent agraement for all N-S walls at both levels. For the E-W walls, the comparison is not as good, but this is to be expected since, for these ells, the first N-S mode is not the only important mode as it is for the N-S walls. In all the redistribution analyses, an equivalent static load system of 1.05 7- 2
O~
V times the first mode' inertial loads was used. In assessing the effects of
' ~
redistribution, the comparisons should be made with Tables 7-1 and 7-2, which are the base cases for redistribution.
The analysis' techni.que used a bilinear capacity or resistance relationship.
Reference values wili'be given;for capacities; this can be considered as the value where the elastic limit is reached. After exceeding this i
value, there is a slight increase in force developed since a finite slope l following the yield -is used in the analysis.
l 7.3 Detailed Description Of Cases The following cases were analyzed using the technique described. in Section 7.2.
j 1
a) Case 1: The capacity of Wall I was limited to 2800 kips between elevation 45' and 61'. This low value was chosen to investigate whether single major walls have a significant effect on the overall system of walls. All other walls had no limit on capacity. The results of this case are shown in Tables 7-3 and 4 and can be compared to 1
Tables 7-1 and 7-2.
l b) Case 2: The capacity of Wall 3 was limited to 2000 kips between j elevations 61'and 77'. This low value was chosen to investigate if i-single major walls have a significant effect on the overall system of walls. All other walls had no limit on capacity. The results of this case are shown in Tables 7-5 and 7-6 and can be compared to Tables 7-1 and 7-2.
7-3 e +,
y - _ - .p.
O v
c) Case 3: The effect of vertical shear transfer was examined. The intersection of Walls 1 and 13 from elevation 45' to 61' was separated so that vertical shear could not be transferred. The results are discussed in Section 7.4 below.
d) Case 4: This case considered all walls limited to the capacities presented in Tables 4-1 and 4-2. The results of this case are presented in Tables 7-7 and 7-8 together with the base case loads and the capacities.
l e) Case 5: This case considered all walls limited to the capacities i 1
presented in Tables 4-1 and 4-2 except for Wall i between elevations 45' and 61' which was limited to 3350 kips. This case assumes a local I I
lack of transfer to the foundation for the North pier in Wall 1.
This condition is highly unlikely since the pier is attached to the grade beam which can, in turn, transfer force to :the compression zone.
The results of this case are presented in Tables 7-9 and 7-10 together with the base case loads and the capacities used in this analysis.
7.4 Results For Case 1, (Tables 7-3 and 7-4), which initially limited the shear capacity at elevation 45'-61' for Wall 1 to 2800 kips, increases occurred in other shear walls oriented in the same direction fran elevation 45' to 61'. The total shear force was reduced indicating that it is resisted in the Eastern portion of the Auxiliary Building with a slight increase at the Fuel Building; but increases in the Fuel Building walls are insignificant compared to their capacities. The shear forces increased in the major E-W walls.
The effect at elevation 61'-77' was basically a' decrease in the force on Wall 1 at that level and an increase in forces on' other walls. The general 7-4
O l I
conclusion is that, as the capacity of one wall is decreased, others will resist the load.
! ]
, For Case 2, (Tables 7-5 and 7-6), the structure behaved similarly to Case 1
- 1. with other walls increasing slight ly to resist the applied loads. -l
- For Case 3, the total vertical shear throughout the height of the structure
, decreased from 1680 kips to 1280 kips, and the membrane force in Wall 13 increased by the difference in shear. The overall moment in Wall 1 increased from 101,000 k-ft to 112,000 k-ft. As illustrated, the system is not particularly sensitive to local shear transfer.
, i Case 4, (Tables 7-7 and 7-8), illustrates the effect of small wall yielding.
The major change occurred in Wall 1 at both elevations 45' and 61'. The force increased from 4170 kips to 4320 kips at elevation 45' and from 3990 to 4100 kips at elevation. 61' . This results primarily from the limited
- capacity of Walls 2 and 3 at elevation 45' to 61'. The slight increase 4
in capacity in the small walls resulted from the fact that some stiffness was used in the computer analyses for these walls even though the initial capacity was set at the values shown in the table. 51ight decreases in 4
the total shear for the sum of the N-S walls indicates that the increment i
of shear is being resisted by the E-W and/or N-S walls toward the Fuel Building end of the complex.
Case S .(Tables 7-9 and 7-10), which limited outer Wall 1 to 3350 kips at elevation 45' to 61', showed that the increment of force previously resisted by this wall is now being resisted by other parallel shear walls and also.the perpendicular walls which aid in resisting torsion.
/
i 7-5 e
1, Tables 7-11 and 7-12 surmiarize the results of the Base Case and Cases 4 and.5 for ease of comparison. The general trends in load redistribution can be seen from these tables. Capacities are not given for the East walls, since with the exception of some smiler walls, the large walls are
]
lightly loaded relative to their respective capacities.
l 7.5 Displacements 1 Table 7-13 shows the elastic displacements for the various load cases. The values at Column Line 55 are for the South-West corner with motion in the North-South direction. The values for Walls 2 and 3 are' the relative values between elevations 45' and 61'. The values are primarily presented -
for compar'ison purposes, since they do not include the cracked stiffness of the structure. Displacement predictions considering concrete and )
masonry nonlinear behavior are determined in Appendix D.
Comparing the fixed base value to Case 4 at elevation 117' indicated an increase by a factor of 1.08. From these values, the amount of yielding can also be estimated for the small walls with capacities in the range of 400 kips.
Wall 2 had a capacity / load ratio of .60 (Table 5-1) for the fixed' base case at a displacement of .026" (Table 7-13). Therefore, the elastic displacement when the capacity would equal the load is
(.60)(.026) = .016". For Case 4, the displacement is .030" or
(',
) = 1.88 times the value when capacity is reached. Based on the bending capacity criteria, the outer reinforcing steel is allowed to be at two times yield. Therefore, the maximum strain is estimated to be 2(1.88) = 3.76 times yield. Doing the same type of calculations for Case 5 leads to 6.4 times yield for the outer fiber reinforcing steel.
~
7-6 m
,e ' :x 7.6 Summary
'A variety of cases have been considered. The structure dernonstrates an excellent ability to redistribute loads, in the event that some walls yield. The analyses have shown that even after redistribution, the displacements- only increased by about 20%. Finally, the analyses show that the small walls, governed by moment, will reach capacity first and then yield. However, the outer reinforcing steel will only be about 6 times yield, and the compres'sion stresses in those walls will be quite low relativ'e to the ultimate strength of the concrete.
7-7 o.
4 .i6 n
-..-- -- ,.-- ,--,- . . -. y
- h. F Table 7-1 Base Case Shear Forces, N-S Motion, Elevation 45' i
SHEAR FORCE l WALL NUMBER (KIPS) LOCATION j 1 4170 N-S Walls i
, 2 800 s 1
3 5%
l 4 2270 5 3030 i
- 6 330 1 7 470 8
4 260 j
[ = 11920 4
e 9 960 E-W Walls l
, 10 880 t
11 70 12 110 4
13 660 4
14 0 a-i 15 21 0 4
hp O
f f
f
- -. . .. .. -. .. - - . . . . ~ - - _ _ _ _ _ _ _ _ _ _ _ _ __
... a e .
L i
. )
b Table 7-2 Base Case Shear Forces,-N-S Motion Elevation 61' l
' I 1
T SHEAR FORCE :
i l
WALL' NUMBER (KIPS) LOCATION
)
3-1 3990 N-S Walls !
i 1 2 20 i 3- 3130 1
4' 1940 l l ,
l 5 430 l': ;
1 i.
6 550
- i 4
[=10620
}- 7 1030 E-W Walls
' l
- 8 1060 !
.t . ,
j 9 630 i
d 10 80 4
11 420 12 230.
- 4 4
1 1.
E 3
b' 2
4 1
I r:
_ w
Table 7-3 Case 1 Shear Forces, N-S Motion, Elevation 45',
Based On Limited Capacity Of Wall 1 From Elevation 45' To 61' SHEAR FORCE WALL NUMBER (KIPS) LOCATION 1 2840 N-S Wall s
, 1 2 1080 3 700 4 2450 5 3190 6 340 7 480 8 270
{=11350 9 1340 E-W Walls 10 1040 11 60 12 21 0 13 800 14 20 15 210
- n. . I
. l 1 ,
Table 7.4 Case 1 Shear Forces,.N-S Motion, Elevation 61', .
- Based On Limited. Capacity . 0f Wall 1 From i
, Elevation 45' To 61' i
j SHEAR' FORCE l
WALL 11 UMBER (KIPS) LOCATION 4
1 3630 N-S Walls I
2 680
' l i 3 3300 i
- j. 4 2040
(-
! 5 440 6 560
[=10650 4
7 1190 E-W Walls l 8 1130'_
- l
, 9 680 i
10 90
{
- 11 430
.l 1 12 280 '
l-i i
4 i
)
1 I i i l il I
J 4
, \
.L
, 1 l
i l
.-1
. - _ _ , -. _ . _ . ___ ._ . _ _ _ - _ _ _ ~ . _ . - . _ -
r.
L i
l l'i Table 7-5:- Case 2 Shear Forces, N-S Motion, Elev. -45', ~ j Based On Limited Capacity Of Wall 3 '
- , From Elevation 61' To 77' i
SHEAR FORCE WALL NUMBER (KIPS) . LOCATION 4
1 4230 N-S Walls 4
2 81 0 3 580 1
, 4 2050 t
5 3080 !
i !
- 6 330 3
7 490 8 270 1
[ = 11840 l
l 9 940 E-W Walls 10' 890 1
- 11 260 l
12 110 l 13 670 h
- 14 10 4
15 220 d
I l
1 Table 7-6 Case 2 Shear Forces, N-S' Motion, Elev. 61',
Based On Limited Capacity Of Wall.3 From-Elevation 61' To 77' SHEAR FORCE WALL NUMBER (KIPS). LOCATION !
i 1 4050 N-S Walls 2 610 I
3 2040 4 2240 5 470 6 580
[=9990 7 1020 E-W Walls 8 1080 9 68 0 10 140 11 460
'12 280 t
P erwe -
f- 1 f - 4 , _ -. -
,, -m L)(
Table 7-7 Case 4 Shear Forces, N-S Motion, Elevation 45' Based On Capacities From Table 4-1 i
WALL BASE CASE SHEAR NEW SHEAR CAPACITY NUMBER FORCE (KIPS) FORCE (KIPS) CONSIDERED (KIPS) LOCATION l
1 4170 4320 5390 N-S Walls 2 800 51 0 470 3 590 500 490 ;
4 2270 2410 3810 5 3030 3250 5970 6 330 120 110 7 470 440 420 8 260 70 60
[=11920 [=11620 [ = 16720 9 960 1200 4730 E-W Walls 10 880 680 5560 11 70 90 240 12 110 20 420 13 660 590 9350 14 0 5 170 15 210 200 760 4
's Table 7-8 Case 4 Shear Forces, N-S Motion, Elevation 61' Based On Capacities From Table 4-2 WALL BASE CASE SHEAR NEW SHEAR CAPACITY NUMSER FORCE (XIPS) FORCE (. KIPS) CONSIDERED (KIPS) LOCATION 1 3990 4100 5100 N-S Wall s 2 580 260 190 3 3130 3190 4520 4 1940 1820 2240 5 430 400 650 6 550 520 750
{=10620 [ = 10290 [=13450 7 1030 820 4440 E-W Wall s 8 1060 830 9340 9 630 690 2820 10 80 120 1380 11 420 420 2390 12 230 180 2390
[=22760
. .. - - - .~. .. . . . . . . -.
\}
l i
Table 7-9 . Case 5 Shear Forces, N-S Motion, Elevation 45' 8ased On Limited Capacity Of Wall 1 From Elevation 45'-61' l
WALL BASE CASE SHEAR NEW SHEAR CAPACITY NUMBER FORCE (KIPS) FORCE (KIPS) CONSIDERED (KIPS) LOCATION
]
1 4170 3410 3350 N-S Walls 2 800 480 470 3 590 490 490 1 l
, 4 2270 2720 3810 5 3030 3500 5970 s .
6 330 110 110 i 7 470 430 420 8 260 60 60 i [ = 11920 [=11200 [=14680 9 960 1390 4730 i
E-W Walls
- 10 880 1080 5560 11 70 90 240 12 110 80 420 i
13 660 890 9350 14 0 20 170 7
15 210 170 760 L
I
- . . -- - - . - - . - . .. . .. _ = _ - -
( )'
Table 7-10. Case 5 Shear Forces, N-S Motion,- Elevation 61' '
Based On Limited Capacity Of Walls From Elevation 45'-61' WALL BASE CASE SHEAR NEW SHEAR CAPACITY NUMBER FORCE (KIPS)' FORCE (KIPS) CONSIDERED (KIPS) LOCATION 1 3990 4200 5100 N-S Walls 2 580 210 1 90 3 3130 3300 4520
'4 1940 1850 2240 5 430 430 650 :
6 550 520 750 ,
[=10620 [=10500 [ = 13450 7 1030 1150 4440 E-W Walls 8 1060 960 9340 ,
9 630 750 2820 ,
10 80 130 1380 11 420 400 2390 12 230 210 2390 l
i l
1 I
I
)
l 1
V)
Table 7-11 Surnary Of Cases, Elevation 45'-61' (All Values In Kips)
WALL BASE CASE CASE 4 CASE 5 NUMBER LOAD LOAD / CAPACITY LOAD / CAPACITY l 4170 4320/5390 3410/3350 2 800 510/470 480/470 3 590 500/490 490/490 i
4 2270 2410/3810 2720/3810 ;
m '
5 3030 3250/5970 3500/5970 h
'd
- 6 330 120/110 110/110 7 470 440/420 430/420 8 260 70/60 60/60 11920 11620/16720 11200/14680 !
9 960 1200 1390 10 880 680 1080 11 70 90 90 m 12 110 -
20 80 a
t
$ 13 660 590 890 l f; 14 0 5 20 15 210 200 1 90
i g(~s Table 7-12 Summary Of Cases Eleva tion 61'-77' (All Values In Kips) 4 i
WALL BASE CASE CASE 4 CASE 5 j NUMBER LOAD LOAD / CAPACITY LOAD / CAPACITY l l 3990 4100/5100 4200/5100 2 580 260/190 210/190 m 3 3130 3190/4520 3300/4520 4 1940 1820/2240 1850/2240 v,
s 5 430 400/650 430/650 4
6 550 520/750 520/750 l 10620 10290/13450 10500/13450 1
7 1030 820 11 50
- 8 1060 830 960
., v, 9 630 690 750 !
a s 10 80 120 130
- x u's 11 420 420 400 12 230 180 21 0 k
l
("\ l
/ i l
l 1
I Table 7-13 Elastic Displacements i
I FIXED BASE BASE CASE CASE 4 CASE 5 4
South-West Corner i
Elevation 61' .036 .038 .039 .090 l 77' .072 .076 .078 .118 117' .145 .156 .156 .187 Wall 2 At Elevation 59' .026 .027 .030 .0 51
, Wall 3 At Elevation 59' .022 .023 .027 .0 41 All values in inches. l
(Vh
- 8. TRANSFER OF SHEAR FORCES TO ROCK FOUNDATION The Control Building, like all other Category I structures in the Trojan plant, is founded on rock. The columns of the steel framing system for the Control Building are supported by spread footings and the shear walls by reinforced concrete grade beams. The grade beams are interconnected by the reinforced concrete ground floor slab. The foundation systen is shown in Figure 8-1. The dead load of the floors is carried mainly by the steel framing system and the spread footings. However, since the shear wall s were built from the ground floor up, it is apppropriate to assume that their weight is carried by the grade beams. Floor loads imposed after the shear walls were built are carried partially by the shear walls and partially by the steel columns. 1 In the Control Building, lateral forces are resisted by the shear walls and are transferred to the rock foundation through the following means:
- l. Grade Beam To Rock Shear Resistance The greatest resistance to sliding is provided by shear resistance between the bottom of the grade beams and the rock. Shear resistance
, to sliding (S )pcan be expressed using Coulomb's formula as:
Sr = CND where:
C = cohesion at zero confinement (psi) u = coefficient of friction 0 = dead load (psi) 8-1
O 1
The above formula is used by the U.S. Corps of Engineers and U.S.
Department of the Interior, Bureau of Reclamation to determine the l
resistance to sliding of concrete dams (Ref.1] . It is also used by other organ.zations in concrete gravity dam design. )
l i
To develop the value of C, it was conservatively assumed that the foun-dation under the Control Building is composed entirely of flat-lying volcanic tuff having an. unconfined compressive strength of 1225 psi, the average of the actual laboratory test results. (See FSAR, Section 2.5.1.5, " Foundation Evaluation".) Based on the average compressive strength of 1225 psi, the design cohesion value of the material is estimated to be in excess of 130 psi. This conservative assessment is based on laboratory results of the unconfined compressive tests of volcanic tuff materials and on triaxial test results on rocks of similar type arc strengths. This approach is very conservative for the following reasons:
- 1) The in-situ rock at the plant site is much stronger than the volcanic tuff materials used in the laboratory tests.
2)' No credit was taken for favorable attitude of the rock units underlying the foundation. .
- 3) The 130 psi value would be used for normal desigr. conditions. i Usually, higher values are accepted for the more extreme conditions of a large earthquake (SSE). That is, these are normal design working stress levels, not ultimate.
8-2 ,
O' The coefficient of friction, u, for this foundation material is estimated to be at least 0.7. This value is a reasonable value for this rock with jointing and laminations.
l The dead load, 0, used in conjunction with the coefficient of friction, l l
includes the weight of the shear walls and the grade beams, but includes no contribution from the dead loads and live loads of the floors.
It should be noted that Winterkorn and Fang (Ref. 2] indicate allowable cohesion stress values between 0.03f{ and 0.05fg. By using the average of 0.04ff, a value of 120 psi is obtained. This compares well with the 130 psi value used in this analysis.
1 l
i Available resistance to sliding from shear resistance is given in Table 8-1 under the heading " shear".
i
- 2. Column To Spread Footing Friction Friction between the steel columns and the concrete spread footings provides additional resistance to sliding. A coefficient of friction equal ;to 0.7 between the steel column base plates and concrete was used in calculating the available resistance to sliding. This is given in Table 8-1 under the heading " footing".
The results of the analysis of the Control Building presented in Table 8-1 reconfirm that the seismic forces can adequately be transferred to the rock foundation.
8-3
(
REFERENCES
[1] Design Standards No. 2, Treatise on Dams, Chapter 9 " Gravity Dams",
U.S. Department of the Interfor, Bureau of Reclamation.
[2] Winterkorn and Fang, Foundation Engineering Handbook, pp. 610-611.
1 l
l l
l l
1 l
l 8-4
O I
@ g
. ' I' ;
I; l llll l
t g
l ll 1 l. -
l
' Il I
l
' l il l g ,
't
. __ a _ _ _
i li g----,:----g,g u~!i j
";I m I II l
I l )
i.
C\
.l i I '
_. _ _ _ J i ,
N t
l l l 1 l lll ! I rw4,
.,, l -
g Giy , , ,.:.___.,. I J.
t
, '! l
'jM -
!_T - - .
lT L_4i
$w f l '
i
( 7 ""^ " " il,l
. i
,y 4'
g _
illD L.11 g
'-l'
,'y l , ii i
i i
il!!
l
.;l l l l i I
i ,
j l 1 i i: 1 I
i--t ;
I .1 i i s i L.l ,4 8 il i i 1 ;i9:
I l ." l li l l j
[
! I 1: l I j ll b Ii I g 1 , .
-l .
l .
I - - -- - - - I I - -- -
p - - - - i ;_
i !
- 9
+
~
n /
777- l I Figure 8-1 Foundation System
n.
U Table 8-1 Sliding Resistance And Base Shear - North-South!
l WALL IDENTITY RESISTANCE (KIPS)
BASE PER PER SHEAR 2 FACTOR OF FIGURE 8-1 FIGURE 3-2 SHEAR FOOTING TOTAL (KIPS) SAFETY R 1 4490 860 5350 4320 1. 24 0 2 1090 225 1315 51 0 2. 58 N' 3 1010 225 1235 500 2.47 I N 4 4680 1350 6030 2410 2.50l 8etween L&N 5,6,7,8 7790 ----
77 90 3880 2.00 Between K&L3 l
TOTALS: 19060 2660 21720 11620 1.87 Notes: IValues for the E-W direction are not given because the factors l of safety are higher. !
2 See Case 4, Table 7-17.
3These two walls were modeled in the STARDYNE analysis as a single wall designated No. 8 and moved to Column Line L.
(
APPENDIX A COMPARISON OF SEISMIC AtMLYSES
- 1. INTRODUCTION Seismic forces due to the horizontal SSE were determined for the Trojan Control-Auxiliary-Fuel Building Complex based on four separate seismic analyses and l one re-evaluation of the original seismic analysis. Due to the differences in
, the analyses, model assumptions and methodologies, different analyses resulted in different response forces in the structures. This appendix explains the differences in the analyses and discusses the differences in the responses.
s The four seismic analyses and the re-evaluation are designated for the purpose of discussion as the original, the re-evaluation study, the TABS, the STAR 0YNE flexible-base and the STARDYNE fixed-base. The original analysis was completed in 1971; the re-evaluation study in April 1978, TABS in June 1978, and the STARDYNE flexible-base and the STARDYNE fixed-base analyses in August 1978.
All analyses were linear elastic analyses based on the uncracked structural stiffnesses and used the modal spectral response technique. The differences among the analyses are described in the following sections and summarized in Table A-1.
- 2. DESCRIPTION OF SEISMIC ANALYSES 2.1 Original Analysis The Control, Auxiliary, and Fuel Buildings form an L-shaped structural complex. These three structures are tied together by common floor slabs at elevaticas 61' , 77', and 93' . The original seismic analysis was perfonned based on a 3D fixed-base beam-stick model consisting of A-1
O essentially four sticks representing the Control Building, the Auxiliary Building, tne Fuel-Building Hold-Up Tank, and the Fuel Building Spent 1
Fuel Pool. These sticks were tied together laterally through beam l
elements representing the common floor slabs. i The lateral and torsional stiffnesses of each stick were determined based on the uncracked structural stiffnesses of each respe:tive structure, except for the Auxiliary Building stick, which was ass;med to have no lateral resistance. Therefore, the lateral resistance of the entire building complex is shared between the Control Building and the Fuel Building.
i The masses of the structures were lumped at each floor level for each stick. The total weight of the lumped masses included in the model was 68,790 kips.
l The modal analysis of the model indicated the fundamental node frequencies of the structure at 6.2 cps for the N-S direction and 6.9 cps for the E-W direction. The modal effective weights of the fundamental modes account for about 60% of the total weight of the structure.
Spectral response analysis was used to obtain the modal responses. The total responses were determined by combining the modal responses by the sum-of-absolute-values ( ABS) method.
A-2
. ~ . .- - ,
O l
The total base shears for the N-S 0.259 SSE with 5% structural damping )
l were 14,210 kips for the Control Building and 11,930 kips for the Fuel l Building, giving a total base shear of'the entire structure of 26,140 kips which is 38% of the total weight.
I The output of the original seismic analysis was converted to force and
)
moment in walls and floor slab by hand calculations. Floor shear forces were distributed according to the relative rigidity of the individual walls; the center of the mass and center of rigidity were considered.
2.2 Re-Evaluation Study l The original analysis was reviewed in April 1978. The weight of the Control Building was re-estimated based on as-built conditions and found l
to be 13% lower than the values used in the original analysis. The !
original analysis could have used the square-root-of-the-sun-of-the-squares 1
(SRSS) method, but used the ABS method. In the re-evaluation study, the SRSS method was used. The total response of the Control Building was i
reduced from the original by 2.0% to account for the difference between the l ABS and the SRSS modal response combination techniques. Applying these two reduction factors to the original analysis base shear resulted in 1
(187)(.80)(14,210) = 9890 kips. The Fuel Building had a 28% reduction '
due to SRSS and the ba:e shear was (.72)(11,930) = 8590 kips.
l l
1 I
A-3 i
j w ,
O 2.3 TABS Analysis This analysis was perfonned using the computer program TABS (,T,hree-Dimensional A,nalysis of B_uilding Systems) developed by Dovey and Wilson of U.C. ,
Berkeley [Ref.1] .
I The model used for this analysis was in accordance with the basic assumptions of the TABS program; that is, the building system is idealized as an 4
assemblage of a system of independent plane frame and shear wall elements
, interconnected by floor diaphragms which are rigid in their own plane.
This simplification reduces the global degrees-of-freedom of the building system to three per floor: two horizontal translations and one twisting rotation. The output of the TABS program are moments and shears in the walls.
Since shear walls are treated as independent plane elements, the flange effect of cross walls or the beam-like behavior of box-type shear wall systems such as the Fuel Building Hold-Up Tank and Spent Fuel Pool cannot be modeled completely.
The masses of the ' structures were lumped for each floor at the center of mass of that floor. The total weight of the lumped masses included in the model was independently estimated as 70,200 kips. The modal analysis of the model indicated four dominant modes, having frequencies at 5.2, 7.2, 12.1 and 13.5 cps, respectively. The first two dominant modes have nearly equal participation in the N-S and E-W directions. The third mode is predominantly an E-W mode while the fourth mode is a N-S mode. The modal effective weights A-4
O' of the four modes are 21, 29, 3 and 26%.of the total weight, respectively, for the N-S direction; they are 24, 23, 33, and 0.3% of the total weight, respectively, for the E-W direction.
4 The seismic modal responses were obtained based on the spectral response technique. The total responses were obtained by combining the modal responses using the SRSS method. The total base shears for the N-S, 0.25g SSE, 5% damping were 6830 kips for the Control Building and 8260 kips for.
the Fuel Building, giving the total N-S SSE base shear for the entire structure complex at 15,090 kips which is 22% of the total structural 4
i weight. When the results were reviewed, it was concluded that the original l and the re-evaluatic.n study forces were conservative.
'I 2.4 STARDYNE Flexible-Base Analysis i
This analysis was performed based on a 30 finite element model of the entire building complex. All the shear walls and floor slabs were modeled by STARDYNE plate elements. Thus, the floor flexibiiity effects, in-plane and out-of-plane, were included in the model. !
To. account for the effect of rocking due to the foundation flexibility, vertical foundation springs were used under the Control Building to connect the base nodes of the finite element model to a fixed boundary. The stiffnesses of these springs were derived based on the BC-TOP-4A (Ref. 2]
formula. The entire model consists.of 685 plate elements and 56 beam elements.
i The masses of the structure were re-estimated in detail and were lumped at 347 mass points distributed throughout the structure. The total weight of the lumped masses in the model was 71,160 kips.
8 A-5
x
,r% ,
C) '
The modal analysis of this model indicated five dominant modes having frequencies at 6.4, 8.4, 9.3,11.7 and 12.4 cps, respectively. The modal l l
effective weights of these five modes are 41. 5, 0. 3, 19,17 a nd 0". of the !
total weight, respectively, for the N-S direction; and are 1.2, 69, 0, 0, and 15% of the total weight for the E-W direction. Therefore, the first, third and fourth dominant modes were N-S modes while the second and fifth dominant modes were E-W modes. The first N-S mode was also a twisting mode pivoting about the more rigid Fuel Building with major modal deflection I at the Control Building end in the N-S direction, l
The seismic response was determined based on the spectral response technique l
using the SRSS modal response combination method. The total base shears for the N-5, 0.25 ,9SSE, 5% damping were 12,200 kips for the Control Building and 7870 kips for the Fuel Building, giving the total N-S, SSE base shear for the entire structural complex at 20,070 kips, which is 28% of the total structural weight.
2.5 STAROYNE Fixed-Base Analysis This analysis was based on the same model as the STARDYNE flexible-base analysis described previously, except that the fixed-base condition was prescribed and the vertical foundation springs under the Control Building were eliminated.
A-6 i
O !
l The modal analysis in this case indicated essentially the same modal behavior as that for the flexible base. The total base shears for the N-S, 0.25g SSE, 5% damping were 11,910 kips for the Control Building and 7680 kips for the Fuel Building giving a total of 19,590 kips for the entire structural complex which is 27% of the total weight. The total shear obtained from the fixed-base analysis differs from that of the flexible-base analysis 'by only 3%, indicating that the foundation flexibility effect was negligible. l 1
- 3. DISCUSSION l
3.1 Comparison Of Model Behaviors All analytical models used predicted a relatively consistent fundamental
{
l mode frequency which is between 6 and 7 cps, except for the TABS model which predicted the frequency at 5.2 cps. This indicates that the TABS model is softer in lateral stiffness. This is due in part to the TABS !
model assumption that shear wall elements are plane elements independent of each other, i.e., the " flange" effect of cross walls and the composite i
behavior of box-type wall systems cannot be representatively modeled. '
All models used also showed a consistent trend that the horizontal seismic response (N-S or E-W) was governed by one predominant (N-S or E-W) mode, except for the TABS model which showed three daninant modes, two of them having nearly equal participations in both the E-W and N-S directions.
l l
J i
! l 2
- A-7 1
The STARDYNE flexible-base and fixed-base models are the most mathematically refined models among all models considered. The models can consider not 4
only the gross structural behavior but also the local element behavior of 1
) individual walls, fiame members, and wall systems. The STARDYNE model I also include the floor flexibility effect with more mathematical preciseness 4
than the other models and directly predicts the elastic distribution of response forces.
3.2 Weight Determination i
The original seismic analysis used a total weight of the model for the
- entire complex of 68,790 kips. A weight calculation of the as-built
! condition for the re-evaluation study was made in April 1978 for the 1
l Control Building area. This calculation indicated a weight 13% lower than that used originally. The weight determination for the TABS analysis was made independently based on the drawings and live load requirements. This determination was considered as approximate since it did not consider
, actual as-built conditions, such as the actual equipment weight. Since a STARDYNE analysis was going to be used to predict the interaction of the total complex (existing Control, Auxiliary, and Fuel Buildings), and the new structural support system, a new weight calculation was made for the entire complex, and the information was used to determine the mass in the STARDYNE model. A comparison of the weight used in the STARDYNE analyses with that used in the original analysis is as follows:
8 J
A-8 I I l~kw , - -
_l ) The Control Building weight in the STARDYNE analyses. is 8.2% lower than that used in the original analysis. !
- 2) The Auxiliary Building weight'in the STARDYNE analyses is 11.3%
higher than that used in the original analysis.
{
- 3) The Fuel Building weight in the STAR 0YNE analyses is 4.6% higher than that used in the original analysis. .
- 4) The total weight of the complex in the STARDYNE analyses is 3.5%
higher than that used in the original analysis.
3.3 Foundation Flexibility Effect The comparison of the STAROYNE flexible-base and fixed-base results shows that the foundation flexibility effect is insignificant, and that a fixed-base analysis is appropriate considering the fact that the foundation rigidity of the Trojan Plant is very high (the shear wave velocity of the i
site rock is 5500 ft/sec).
3.4 Comparison Of Results "
i The original analysis predicted the highest total N-S SSE base shear (26,130 kips) which is mainly due to the sum-of-absolute-value (A85) modal response combination technique. The re-evaluation resulted in a total base shear of 18,480 kips. The total base shear of 18,480 kips is close to the 19,590 kips predicted by the:STARDYNE fixed-base analysis. The. TABS analysis resulted in 'the lowest total N-S SSE base shear (15,090 kips).
All analyses predicted a. higher proportion of the total base shear to be resisted by the Control. Building than by the Fuel Building, except the TABS i
l A-9
>- ,l
O.
analysis which predicted a higher proportion of base shear to be resisted i by the Fuel Building. This appears to be directly related to the particular model behavior of the TABS model as explained previously. Despite this 4
higher proportion, the absolute value of the base shear resisted by the l Fuel Building as identified by TA8S is still lower than base shears predicted -
- by the other analyses.
, The STARDYNE analyses, both the flexible- and fixed-base cases, resulted l in a higher proportion, i.e., 61%, of the total N-S base shear to be resisted 4
by the Control Building, as compared to 54% from the original analysis and 45% from the TABS analysis. This is mainly due to the increased torsional
- flexibility inherent in the STARDYNE model as reflected in the mode shape of the fundamental N-S mode described previously. This torsional flexibility appears to be due to the combined effects of floor flexibility
! and increased relative lateral rigidity of the Fuel Building as a result 1
j of the finite element modeling of the box-type Hold-Up Tank and Spent Fuel Pool shear wall systems.
I 4.
SUMMARY
The inertial forces are summarized in Table A-1. The STARDYNE flexible base analysis had undue conservatism due to the lack of radiation damping. The TABS i analysis was used to gain confidence ~ that the re-evaluation was conservative.
! The base-shear predicted by the original analysis is higher than that predicted by l otner analyses.
l i
1 p A-10 l
i I ,
i
U,rn The re-evaluation study results are within 6% of the STARDYNE fixed-base analyses for the total base shear. This is an excellent agreement considering the difference in mathematical sophistication that has evolved from 1971 to the present. The najor differences result from elastic load distribution which has led to a load increase in the Control Building end of the structure, as reflected in Table A-1, 1
Concrete structures only behave elastically at very low stress levels; beyond such levels, cracking develops and load redistribution occurs. The effects of load redistribution are exanined in detail in Section 7.
l l
A-11
REFERENCES l (1) Wilson, E.L., and Dovey, R.H. , " Static and Earthquake Analysis of Three Dimensional Frame and Shear Wall Buildings", EERC Report 72-1, U.C.
Berkeley, 1972.
[2] BC-TOP-4A, " Seismic Analysis of Structures and Equipment for Nuclear .
1 Power Plants", Topical Report, Bechtel Power Corporation, San Francisco, CA, Revision 3, November 1974.
1 A-12 i
s 9
J Table A-1 Summary Of Seismic Analyses Total Inertial Load N-S, SSE .25g, a = 5%
ANALYSIS CONTROL BUILDING FUEL BUILDING TOTAL DESCRIPTION SHEAR (KIPS) SHEAR (KIPS) SHEAR (KIPS)
Original 14,210 11 ,930 26,140 Re-Evaluation 9,890 8,590 18,480 TABS 6,830 8,260 15.090 STARDYNE Flexible-Base 8/17/78 12,200 7,870 20,070 STARDYNE Fixed-Base 8/24/78 11,910 7.680 19.590 i
l0 I
, APPENDIX B l l SHEAR CAPACITY CRITERIA
- 1. INTRODUCTION I l
This appendix describes the evaluation of fully grouted hollow concrete block shear wall test results and, based on these test results, the development of shear wall criteria used to evaluate the Trojan Control Building. Two sets 1
of test data are considered applicable to the capacity evaluation of the grouted concrete block shear walls (Refs.1 and 2] . Although each of these two test series has different test conditions and parameters, the results of both series can be used to derive realistic criteria for the determination of concrete block shear wall capacity.
P Empirical relationships arrived at by Schneider (Ref.1], and recent concrete ,
block shear wall test results-obtained by the University of California, Berkeley (Ref. 2], are used herein for the basis of criteria. Cyclic degradation is included in the criteria by comparison with cyclic tests. In addition, dowel capacity criteria are discussed and comparisons are made. Finally, the criteria are given for horizontal and vertical shear capacity.
l
)
- 2. DESCRIPTION OF TEST SET-UPS AND CONCLUSIONS !
l 1
1 2.1 Schneider's Tests 1 I
Tests were performed in 1969 at California Polytechnic College, Pcmona, California, by Prof. R.R. Schneider (Ref.1] . Various parameters were investipated in these series. These parameters can be cited as aspect ratio '
- i (H/W), vertical and horizontal reinforcement ratios (py and oh ), and grouting. !
.The testing scheme is shown in Figure B-1. ;
1 1
B-1 L
Based on all of the test results, Schneider developed the following empirical relationships:
H/W ULTIMATE SHEAR STRESS (PSI)
- l 0.2 < H/W < l .0 V/TW = 310-175H/W M U jlj 1.0 _ < H/W < 3.0 V/TW = 152.5-17. 5H/W
.a - e jjj 3.0 < H/W V/TW = 100
- 0. 2 < H/ W < l . 0 V/TW = 347.5-ll2.5H/W q@ 1. 0 < H/W < 3. 0 V/TW = 290-55H/W uv lj 3.0 < H/W < 4. 0 V/TW = 200-25H/W
.c - a jjj 4.0 < H/W V/TW = 100 where:
H = height of wall (in)
W = length of wall (in)
T = gross thickness of wall (in)
V = capacity of wall (lbs)
Schneider's test results used herein are tabulated in Table B-1.
For those tests with N = 0 (Figure B-1), the test report [Ref.1] infers some axial compression. Although the indus . exial force into the pier is not documented (Ref.1), several simplified linear analyses (with variable specimen modulus) were made by Bechtel with a finite element computer pro-gram. These showed that 25% to 40% of the vertical component of the diagonally applied load would be resisted by the two struts in compression, dep?nding on' the diameter and length of the struts as 'well as the stiffness !
assigned to the masonry portion of the test specimen.
B-2
6000 psi. In each wall, the reinforcement of the concrete blocks when considered over the gross area meets the minimum reinforcement requirements of the 1967 UBC. If only single wythe concrete block is considerede the reinforcement ratios in B-6 O- l i the Trojan walls for 7-5/8" thick block wythe ranges from 0.0024'(#6/24") l to 0.0048 (2#6/24") for horizontal steel and from 0.0017 (#5/24") to 0.0024(#6/24") for the vertical steel. These are within the ranges of j the Berkeley HCBL test specimens. j l Comparison of important material strengths of the Trojan walls with those ; f of Berkeley (HCBL) series and Schneider's test specimens is given in Table B-6. As shown in this table, the actual strength of Trojan wall blocks, mortar, cell grout and core fill concrete exceed those of the
- test specimens. The Berkeley single wythe specimens used blocks with net s
- q. block area equal to 58% of the gross. Schneider's specimens had 50%. The block to. gross area percentage of the blocks in Trojan walls is 50%. For the thinnest composite walls at Trojan which are 20" thick, the block area over the gross wall area is 40%. For composite walls in Trojan thicker than 20", the ratio of block to gross area decreases further. Therefore, the composite strength is higher in all the composite walls at Trojan than the strength of the blocks used in the Berkeley and Schneider tests. If the core reinforcement is considered as additional reinforcement other than at #
the discontinuities, the reinforcement ratios >then are higher than most of the test specimens considered for establishing the criteria. 4 HORIZONTAL SHEAR CAPACITY CRITERIA Comparison with both the Schneider and Berkeley tests has established that the ( modified Schneider criteria correlate very well in terms of resistance capacity. ' All specimens had vertical restraint; this primarily was to ensure the specimens failed in shear and were not limited by a bending failure. Bending failures are prevented in specimens with low aspect ratios (in the range of 2.0 or less) by either confining them by upper and lower stories or the application of dead load. B-7 l o O The major Trojan Control Building shear walls have significant dead load or , can mobilize it by small movements that would lead to unloading of the steel , columns and a transfer of load to the shear wall. Therefore, for these types of walls, the modified Schneider criteria are classified as the basic criteria. Since all the Trojan walls have more horizontal block reinforcement than vertical block reinforcement, the Schneider criteria with horizontal reinforcement are used. The basic criteria are obtained by reducing the criteria shown on page B-2 , by 25% and adding the axial load effects. Basic Criteria: V u = [261-84 h + ]WT .2 5hj1.0 V = [218-41 h + ]WT 1.0 $h1 3.0 V = [150-19 h + ]WT 3.0 3h3 4.0 V = [75 + }0 }WT 4. 0 3h p p h + #v i .002: y 1 0007 and ph 1 0013 for h 3 2 use f = .2 where: V u = allowable shear force (lbs) H = height of wall (in) W = width of wall (in) T = gross wall thickness (in) o n = axial stress (psi) (+ for compression; - for tension) ph = horizontal reinforcement ratio p y = vertical reinforcement ratio B-8 ,L( 4 i In order to qualify for the basic criteria, the walls must be subjected to dead load and have core reinforcing steel. Walls which have structural steel columns completely interrupting the core shall use an h ratio which limits W l to the distance between columns. i I i i Additional Criteria: ) a) Walls which are subjected to dead load, but no core reinforcing steel is l assumed..may be evaluated by the basic criteria; however, the stress is 4
- limited to 150 psi.
4 b) Walls which are not subjected to dead load, such as interior walls t between floor slabs, shall be evaluated using the following shear criteria: 1 Af j V = 4 [2 /f[ t,ff + )s y](.8W) u t where: $ = capacity reduction factor (0.85) ! f' = concrete compressive strength (6000 psi) t,ff= total wall thickness -8(in) f = yield strength of reinforcement (psi) j A, = minimum of horizontal or vertical reinforcement (in2/ft) l Other terms have same definition as given under the basic criteria, T i c) All walls must be checked for combined bending moment and dead load effects. c k B-9 .O: 1 These foregoing criteria are based on an empirical formulation which correlates , very well with recent. testing. They are conservative in. that:
- 1) The criteria are based on masonry testing and the actual structure is a composite of masonry and reinforced concrete.
- 2) .The concrete core and cell fill strength of the Control Building walls are much higher than the test specimens.
- 3) The added strength of the structural steel is not included.
- 5. VERTICAL SHEAR CAPACITY CRITERIA Vertical shear capaci ty is considered as resisted by the dowel action of the horizontal reinforcing steel and the steel beam-to-column connections.
] i 4 e D B-10 5.1 Rein forcing Dowel Cap 6::1ty Table B-5 shows a comparison of Schneider test results and dowel strength, assuming the dowel could develop a strength equal to its tensile strength of A su f . The Schneider test specimen had some induced compression and the .75 factor applied to the test data is a conservative estimate of what the maximum shear strength would be without the compression. Based on this information, the dowel strength ranges from .65A f to over u 1.0A3 f . However, as shown in Figure B-4, the test specimen reinforcing was located at the outer edges of the specimen and one of the dowels would pull out as shown in the figure before attaining f011 strength. l Test 84-3 from the PCA tests (Ref. 3] illustrates a situation where the l web of the test specimen carried the entire load since there was' no horizontal l reinforcement. In this test, the ultimate shear strength was v = 810 I psi. The test had: i 1 h = .5 p = .005 y p h =0 f y = 78 ksi f u = 120 ksi The strength based on dowel action at ultimate is: y =p y f. = (.005)(120,000) = 600 psi B-ll O Since the specimen reached 810 psi there is an increase of = 1.35. Thus, the test exhibited an ultimate shear strength 1.35 times greater than the calculated ultimate dowel action strength. The apparent reason for this strength increase is the f act that the reinforced concrete specimen had uniformly distributed reinforcing and there is most likely some shear friction. Nelson stud tests (Ref. 5) have shown shear strengths of 95% of the tensile ultimate strength for studs in the same size range as the reinforcing steel in the Control Building walls. Tests documented in (Ref. 6] have shown for steel embedments welded to plate that the shear strength for small steel areas :in the range of 1 in: is 70% of tensile ultimate strength. These tests are very conservative since they only had one-sided embedment and they also had axial movement. The-one-sided embedment results in higher bending moments than two-sided embedment and a slightly red'uced capacity. Based on the preceding, it can be concluded that the dowel shear strength of reinforcing bars in the range of 1 in2 and less are at least 90% of tensile ultimate strength. 5.2 Beam Connection Capacity For the strength of beam-to-column connections, twice the AISC - Part I allowables are used. This value is considered as a lower limit of the ultimate strength that can be obtained. B-12 a .~.. . . _ _ - . ._ . _ _ . _ _ _ ..(). s a ! 5,3 Vertical Shear Capacity Criteria . r The capacity of the vertical planes is determined by combining both ; i the reinforcing steel' dowel shear. strength and the beam-to-column {, , connections shear strength. These capacities are compared to the total l shear load in the vertical section under' consideration. These capacities - L j ' conservatively neglect any capacity of the concrete and masonry block. i i ! Total capacity = 2(AISC Connection Allowables)+ (.9)f [A, , i where: f u= horizontal reinforcing ultimate strength {A = summation of horizontal reinforcing steel areas i i ~ .I 1 I l l i B-13 1 l l I l l () REFERENCES 4
- - (1) Schneider, R.R., " Shear in Concrete Masonry Piers", California State Polytechnic College, Pomona, CA, 1969.
[2] Mayes, R.L., Clough, R.W., Hidalgo, P.A., and McNiven, H.D., " Seismic l Research on Multistory Masonry Buildings", North American Masonry i Conference, Boulder, CO, August 14-16, 1978. l [3] Barda, F. , Hanson, J.M. , and Corl ey, W.G. , " Shear Strength of Low-Rise a j Walls with Boundary Elements", ACI, SP53-8. [4] Froerer, 0.0. , "The Benefits of High-Strength Masonry" North American l l Masonry Conference, Boulder, CO, August 14-16, 1978. i j [5] " Design Data, Nelson Concrete Anchor Studs for Securing Steel to Concrete", i Gregory Industries, Inc., Lorain, OH.
- [6] " Containment Liner Plate Anchors and Steel Enbednent Test Results", P.L.
1 Chang-Lo, et al. , 4th International Conference on Structural Mechanics in Reactor Technology, San Francisco, CA, August 1977. 1 4 2 7 , B-14 ( . P iNo 21 i - /, . t / k k[ 2 , P .oiAces4Moto 5:4x (PIER DETAILS VAAY) Figure B-l Schneider's Test Scherre (Ref.1] bSP**eG3 [ ll *'9 MAM ACTuaf043 { I " LCAO CELLS I ii % ' - COLUMN 5 i ;l $P(Cwffe e i., i , ,, ' -~~ocs Ai, , . , , at , LOOP .svcL
- l. I
, Figure S-2 Single Pier Test Set-Up (Ref. 2] T
- 9
O l 4 Criteria: 300 - y u = 0.75v' u + 4 e
- Schneider's Tests
. + Berkeley (HCBL) Tests . e l + l $200-
- e 2'
=,, e a I . Me. s . +tNo reinforcement 100- , l 0 s s Y 100 200 Calculated (Criteria) - v (psi) u Figure B-3 Test Results Vs. Criteria : 'l f O i 1 l l 1 4 I g Shear Force y y e T l A Crack - h W d i Figure 8-4 Typical Reinforcing Bar Location At Cross-Section (Schneider's Tests) ] 1 1 S ~ ' . .v . ,: .Of t-4 .g. !. Table B-1 Schneider's Parameters And Test Results [Ref.1] i TEST VgTEST i
- A H- W u- c CAPACITY D O j PANEL (FT) (FT) H/W. (PSI) (IN2 ) y h (KIPS) l :
1 1.33 3.33 0.4 260 32 0 .0038 ----- 83 2 '5.33 2.00 2.7 1 31 1 92 .0063 ----- 25 I 6 5.33 2.67 2.0 124 256 .0048 ----- 32 7 4.00 4.00 1.0 140 384 .0032 - - - - - 54 i l 9 1.33 4.00 0.33 290 38 4 .0032 - - - - - 111 I i i I 12 4.00 4.00 1.0 243 384 .0031 .0048 93
- 13 5.33 2.67 2.0 178 256 .0048 .0048 46 15 6.00 2.00 3.0 128 1 92 .0063 .0048 25 28 5.33 2.67 2.0 137 256 .0047 .0010 35 31 4.00 4.00 1.0 158 384 .0047 -----
61 33 5.33 2.67 2.0 97 256 .0070 - - - - - 25 j l 8 2.67 5.33 0.5 203 512 .0024 ----- 104 l 4 l i a 2 I Table B-2 Schneider's Test Results Compared With His Criteria And The Basic Criteria SHEAR STRESS (PSI) : i PANEL H/W oy ph TEST v* u COMMENTS
- l 1 .40 .0038 -----
260 245 No horizontal reinforcing H/W < .5 9 .33 .0032 ----- 290 262 8 .50 .0024 ----- 203 216 7 1.0 .0032 ----- 140 136 No horizontal l reinforcing H/W = 1.0 l 31 1.0 .0047 ----- 158 141 12 1.0 .0031 .0048 243 237 Horizontal reinforcing ) H/W = 1.0 l 6 2.0' .0048 ----- 124 11 9 No horizontal reinforcing H/W = 2.0 ! 33 2.0 .0070 ----- 97 112 l 13 2.0 .0048 .0048 178 180 Horizontal reinforcing H/W = 2.0 167/ 28 2.0 .0047 .0010 137 1 21 ** 2 2.7 .0063 ----- 1 31 112 No horizontal reinforcing 15 3.0 .0063 .0048 127.5 126 Horizontal reinforcing Notes: *v u = 0.76vSchneider + ("n/4) where on"Vtest /4
- Second value considers the specimen has no horizontal ;
reinforcement. ' i I l ((/ l P Table B-3 Sin 91 e Pier Test Results [Ref. 21 l VERTICAL REBAR lt0RIZONTAL REBAR Mm ^ G "o E' @ t E L. O C C G #$ " ;E 8 m S SE E 2 E $ Cw uw ww 'd ood >- m m 5 de EW Eu M ep 5 E s o, E 12 o, - es W m w b!2 $5 i a CE ww x2 5 on ** $ on **
- G o$ "mG guG ggg ebG SPECIIlEN DESIGNATION gg go_
g gg gg gg gg gg gg . pg gg u -g pegg > wgg gag > gg gg ,r y rg g g g ,r ggg wgg gg3 l HCBL-ll-1 1.5 F 1330 None ---- ----- None ---- ----- ---- ----- 45.0 46.6 44.0 llCBL-ll-3 1.5 F 1833 2-#5 70.8 .0016 None - - - - ----- ---- .0016 45.0 49.1 25.1 !!CBL-ll-4 1.5 F 1833 2-#5 70.8 .0016 1-#5 47.9 .0008 :14.7 .0024 53.1 54.9 39.1 liCBL-11-6 1.5 F 1833 2-#5 70.8 .0016 4-#5 47.9 .0032 ,58.8 .0048 80.0 82.7 52.7 llCBL-ll-7 1.5 F 1905 2-#8 69.2 .0041 None ---- ----- ---- .0041 52.5 65.8 33.3 !!CBL-11-9 1.5 F 1905 2-#8 69.2 .0041 2-#5 47.9 .0016 29.4 .0057 53.8 56.9 41.9 IICBL-ll-ll 1.5 F 1330 2-#8 69.2 .0041 4-#6 73.9 .0046 130.6 .0086 84.4 87.7 50.8 (j/3 ' Table B-4 Berkeley Test Results Vs. Criteria BERKELEY MODIFIED SCHNEIDER N SPECIMEN Yu(TEST) 4WT CRITERI A DESIGNATION Py(%) P (%) (PSI) .75v' (PSI)1 v (PSI) HCBL-11-1 ---- --- 123 99 28 1 27 HCBL-ll-3 .16 --- 123 99 16 11 5 HCBL-il-4 .16 .08 145 99 25 124 HCBL-ll-6 .16 .32 21 9 170 34 204 HCBL-11-7 .41 --- 143 99 22 121 HCBL-11-9 .41 .16 147 9 92 27 126 HCBL-ll-ll .41 .46 230 170 33 . 203 Notes: 1The contribution of axial compressive load. 2This case has been interpreted as having no horizontal reinforcement. 1 i (~m V Table B-5 Comparison Of Schneider's Test Results And Dowel Capacity (No Horizontal Reinforcement, H/W < .05) SHEAR STRESS (PSI) .75 TEST TEST PANEL H/W p y TEST .75 TEST DOWEL DOWEL 00W$L l 1 .40 .0038 260 195 302 .65 .86 9 .33 .0032 290 218 254 .86 1.14 8 .50 .0024 203 152 202 .75 1.00 l l l l l l l l I l l l l --M .' V. Table B-6 ' Material Strengths TiROAN WALLS i-SPECIFIED AVEftAGE BERKELEY SOiHEIDER'S STRENGTH MINIPUM VALUE IICBL TEST-ITEM PARAMETER NOTATION VALUE (IN-PLANE) TESTS SPECIMENS Heavyweight Masonry f' 20001 4100 3000 3000 (Net) concrete blocks compressive " 1500 (Gross) strength (psi) Lightweight Masonry f' 20001 2700 ---- - - - - concrete blocks compressive " strength (psi) Mortar Mortar f 2000 3500 2500 3000 (Avg.) cm
- compressive ASTM strength Type M 1 (28 day psi)
Cell Grout Concrete f' 50002 >60003 40 00 3000 (Avg.) c ! compressive Grout: , , strength (psi) 1C:35:2G Core fill Concrete f' 50002 >6000~ ---- ---- concrete coinpressive c . j strength (psi) ! Reinforcing Yield stress f 40 >45 48 to 71 55 steel (ksi) Y I Notes: I Based on net cross-sectional area. 2 28-day compressive strength. t 3 Based on 90-day compressive strength. i __________m._ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ . _ _ _ _ _ _ _ _ _ _ _ _ _ _ _ , _ _ _ _ _ _ -- _--- - -w - --< w ~ , . o . \ APPENDIX C I STRAIN COMPATIBILITY This appendix addresses ~ the strain compatibility of the various components in the Trojan Control Building walls - The composite behavior requires that shear strains in the masonry and concrete are the same. The related stresses, however, will be quite different due to the different material properties. The walls are composed of masonry blocks and a concrete core. The cells in the blocks are filled with concrete. For a typical wall about 50% of the area is masonry and the other 50% is the concrete core. J l Both the Berkeley masonry tests (Reference 2, Appendix B)' and the PCA reinforced concrete tests (Reference 3, Appendix B) had shear strains ~in about the same range when the maximum capacities were achieved. l The PCA test showed that cracking occurs at a shear strain in the range of 350 u in/in to 400 u in/in for specimens without large amounts of reinforcement and for aspect ratios between .25 and 1.0. The shear strains at maximum capacity rarged between 5300 u in/in and 8500 u in/in. The Berkeley tests for specimens with horizontal reinforcing equal to or greater than the vertical reinforcing had shear strains of about 1250 p in/in at the start of nonlinearity and in the range of 6500 L in/in at maximum capacity. I Results of some representative test specimens are plotted in Figure C-1 for comparison. From this plot it is illustrated that maximum capacities are reached at about the same shear strain and that for a given strain shear stresses in concrete are substantially greater than the stresses in masonry. Therefore, C-1 s . t )- . calculations of shear capacities of composite walls t,ased on masonry strength values for both materials are conservative. C- 2 r I ? 800 - * = .5%, p = 0% PCA(84-3)o y h f' \ ( PCA(B6-4)p y = .25%, p h = .50% / / /'\ ~~,*N 600 / j / \ N N \ \. N s / / / 3 / \. '. o l 6 - y u m 400 - // / / / N.N,N e 8 */ rn - j,/ . I 200 - , Berkeley (Average 6 and 11) ,A
- Test 6.p y
= .16%, o = .32% h Test ll p = .41%, o = 46% y h .- , , , i . . . g e i e a i 5000 10,000 , Shear Strain p in/in j l 1 Figure C-1. Comparison Of Shear Stress Vs. Shear Strain l Of Test Specimes l l .,i r l o i)
- APPENDIX D DISPLACEMENT DETERMINATION In this appendix techniques are developed for determination of the displacement of the structure based on information presented in Appendix C. Figure C-1 is pre-sented again as Figure D-1 with the shear moduli shown for the Berkeley and the PCA test specimens. Also shown on Figure D-1 is another curve which is considered representative for the composite Trojan Control Building walls.
Since the walls consist of about equal portions of masonry and concrete, the composite curve for the system can be expressed as an average of both portions as shown in Figure 0-1. The composite curve was constructed as follows: The average of the initial slopes based on the lower PCA and the Berkeley tests is: l G 1 = (.8+.09)x106 s .45x106 psi 2 l From the Berkeley shear stress-lateral displacement curves, it is estimated that the 1 nonlinear behavior of these specimens started at about 100 psi. This was used i l as the limit for the initial shear modulus. For the second slope, again the average was used: G2 = (.07+.02)x106 = .045x106p sj l 2 l l l l The following relationships determine the shear strain (y) as a function of the ' shear stress (v): I 0-1 : O . O. Y" (.45x106) for 0 3y .5 100 psi y=(.04hx 6) + .00022 for v > 100 psi Based on the average shear stress per story obtained fran results of Case 4, Section 7 and the story height, the interstory displacements and the total building displacement can be calcualted. The following calculati,ons are fnr the West wall (Wall 1) for the N-S .25g SSE a t 8 = 5%. The displacement can be obtained by multiplying the story height l by the shear strain. Elevation 93'-117' H = 288"; y = 60 psi 6 3 =[4 . 10e ] 88 = = .04" 1 Elevation 77'-93' H = 192"; v = 115 psi 1 4 2
- I( . 5 = .11"
) + .00022]l92 l l Eleva tio:n 61 '-77 ' H = 192"; v = 135 psi 6 3 = [( jI sy + .00022]l92 = .19" Elevation 45'-61' H = 192"; V = 215 psi 6 4 = (( gj, 5) + .000221192 = .53" Total Displacement Elev.117' N-S [ = .87" t .9" D-2 . ._ __ ._. . . _ . _._ _ ~_ _ _ . _ . - . _ _ _ . _ . _ ~ . . _ 6- .Q ' Applying the'same technique to the North wall using the stresses obtained from ! Section.3 for the E-Wf.259 SSE"at a = 5% results in the' following: . El evation 93'-117'. H.= 288"; v = 25 psi l 25 = . 016"
- 6) = [(45x106)]288 l
Elevation 77'-93' H = 192"; y = 48 psi 48 6 = .020" 2
- I(.45x106)]192 1
l Elevation 61'-77.H = 192"; v = 56 psi l 56 6 3 = [(.45x106)ll92 = .024" Elevation 45'-61' H = 192"; v = 80 psi 80 6 = .034" 4 = [(.45x106[]l92 Total displacement Elev.117' E-W [ = .094"s .09" l The predicted displacements at the Control Building roof are .9" for the North-South direction and .09" for the East-West direction. D-3 ,~- r ,, + < . , . . . , _ _ , ____ _ _ ' _ _ _ _ _ _ _ _ _ O. - U 800 - g PCA (86-4) 600 / / / -~.'s 'N s / N / s 2 / 's ~ m / %, 0 / h / G = .07x106 / $ 400 - / a .c e ^ / / y Composite s G = .045x106 G = .8x106 lG, = .45x10e . 200 - y '/ * % - . G = .02x106 l j./'s Berkeley (Average 6 and 11) ! G = 09x106 0 ' ' . i i * * ' ' I . . i 5000 10,000 Shear Strain u in/in Figure D-1 Shear Stress Vs. Shear Strain Of The Test Specimens 1 4 APPENDIX E INELASTIC BEHAVIOR Section 7.0 of the report describes how the Control Building structure distributes maximum elastic loads when some of the members yield. Such yielding results in inelastic behavior and the seismic response forces based on elastic analysis' cannot develop. This section presents a brief discussion of structural inelastic behavior. The ability of a structure to resist the dynamic loadings due to an earthquake depends not only on the structural capacity but also on the inelastic deforma-bility and energy absorption capacity of the structure. Seismic forces computed on the basis of a conventional linear elastic design response spectrum are the forces which would develop in the structure if the structure responds to the earthquake linear-elastically. If the yield capacity of the structure is less than that required to resist the earthquake elastically, then inelastic defor-mation occurs. When inelastic deformation occurs, a load-limiting mechanism is developed, since there is energy absorption associated with the inelastic defor-ma tion. Thus, the inelastic response of the structure is reduced as compared to the elastic response, the amount of the reduction being directly related to the amount of inelastic deformation and thus energy absorption that occurs. If the structure is idealized, based on its fundamental mode, by an elasto-(perfectly)-plastic single degree-of-freedom system in which the " yield" point corresponds to the structural capacity, then the inelastic seismic response of -the structure to an earthquake can be expressed as a function of the " ductility i factor" of the elasto-plastic system. The ductility factor designated as u is defined to be the ratio of the maximum displacement (including the elastic and plastic components) to the displacement of the system at the yield point. E-1 o l h I i The allowable ductility factor of the system represents the maximum inelastic deformation permitted in the structure, which depends on the material, the structural type and the type of construction in the structure. The Unifom Building Code (UBC) relies on inelastic behavior to resist earthquake response forces. This is apparent when comparing the specified UBC earthquake forces with elastic response forces as determined from a response spectra, such as that used for the Trojan Power Plant. Based on the results of analytical studies on the seismic response of elasto-plastic systems, Newmark and Hall (Ref.1] developed the inelastic (elasto-- plastic) seismic response spectrum similar to the elastic response spectrum l 1 of a linear elastic system. The inelastic response spectrum values rcpresent l 1 the seismic responses of an elasto-plastic system to an earthquake for a specific damping value and ductility factor as a function of the (elastic) frequency of the system. For se smic design purposes, the studies concluded that the ratios of the elaste plastic to the elastic response spectrum va lues , expressed in terms of the ductility factor u, are as shown in the follt ing table [Ref. 21: RATIO 0F ELAST0-PLASTIC TO FREQUENCY RANGE CORRESPONDING ELASTIC RESPONSE SPECTRUM VALUES TO THE QUANTITY CONSERVED BETWEEN THE ELASTO-PLASTIC TOTAL RELATIVE ABSOLUTE AND ELASTIC SYSTEMS DISPLACEMENT ACCELERATION Displacement fu2 2.0 cps
- 1 f
Energy and Velocity " I 2.0 cps j f u < 9. 0 c p s* /2u -l G Force and Acceleration u 1 fu1 9.0 cps *
- Frequency values are approximate.
E-2 n ) ( ,- The Trojan Control Building has a frequency of about 7.0 cps in the North-South ' l direction based on elastic uncracked properties. After concrete cracking the frequency will be somewhat reduced, but still remain in the frequency range 1 where energy and velocity are conserved. The required resistance will be (1//27T) of the elastic value and the resulting displacement will be (u//2u-1) times the 1 elastic values. The following table illustrates the resistance required and the displacement for various low ductility factors. DUCTILITY INELASTIC DISPLACEMENT INELASTIC CAPACITY FACTOR (u ) ELASTIC DISPLACEMENT ELASTIC CAPACITY 1.25 1.02 .82 1.50 1.06 .71 2.00 1.15 .58 3.00 1.34 .45 5.00 1.67 .33 i The Trojan Control Building is not a single oscillator, but it is dominated primarily by one mode of response. An indication of its behavior in the elastic i range can be seen by making the extreme assumption that the structure had a yield i 1 capacity (or begins to go inelastic) only one-half of that required to resist a ' .25g earthquake in the linear elastic range. Based on the values given in the previous table, the required ductility is 2.5 and the displacement is only 25", greater than it would he to resist the earthquake in the linear-Mastic range. i 4 I E-3 ,s U REFEREllCES j l l [1] Newmark, N.M. , and Hall, W.J. , " Procedure and Criteria for Ea rthquake Resistant Design", National Bureau of Standards, Report No. NBS BS5 46, Building Practices for Disaster Mitigation, Building Science Series, No. 46, Vol. 1, pp. 209-236, February 1973. , [2] P!ewinark, N.M. , " Current Trends in the Seismic Analysis and Design of High Rise Structures", Chapter 16, Earthquake Engineering, edited by R. Wiegel, Prentice-Hall, NJ, pp. 403-424, 1970. l ) k l 1 l l l l 4 E-4 r ( PROFESSIONAL QUALIFICATIONS OF GEORGE KATANICS PRESENT Chief Civil / Structural Engineer POSITION Bechtel Power Corporation San Francisco Power Division PROFESSIONAL Registered Civil, Structural, and Mechanical Engineer DATA in California, Civil Engineer in Louisiana, and Professional Engineer in Arkansas; Member of of American Society of Civil Engineer and the Structural Engineer's Association of Northern California EDUCATION MS, Civil and Structural Engineering, Technical University of Budapest, Hungary; MBA, Golden Gate University, San Francisco.
SUMMARY
4 years: Chief Civil / Structural Engineer 4 years: Project Engineer 3 years: Civil / Structural Group Supervisor 1 year: Staff Assistant to the Chief Civil /
Structural Engineer 3 years: Senior Engineer 2 years: Job Engineer
. (All the above experience with Bechtel)
]
13 years: Structural Design Engineer and Senior ;
Structural Design Engineer 4 i
EXPERIENCE Mr. Katan'ics is presently the Chief Civil / Structural !
Engineer of Bechtel's San Francisco Power Division. l In this capacity he provides overall professional guidance and supervision to all Civil and Structurcl Engineers in the Engineering Department. His respon-sibilities include staffing of projects, administra-tion and technical development of personnel, technical review of project work, selection .of consultants, development of standards, and supporting projects
.in regulatory and other important technical matters. l Before his present assignment, he work as Group Supervisor and Project Engineer on the Arkansas Nuclear One, Unit 1 project. As Group Supervisor, .
he was in charge of the civil and structural engin~
eering design of that plant. As Project Engineer on the same project, he directed and supervised l all engineering activities (mechanical, civil, electrical, plant design, instrumentation) and worked with the client, construction, startup, and regulatory agencies, i
O EXPERIENCE (Cont'd.) .
Mr. Katanics served as Staff Assistant to the Chief Civil Engineer of the San Francisco Power Division where he worked on the development of criteria and analytical methods for the design of nuclear power plants. He also participated in civil / structural design and design reviews of nuclear power plant projects.
Previously, as a Senior Engineer and Job Engineer with Bechtel, Mr. Katanics worked on engineering studies and civil / structural design of nuclear and fossil-fired power plants, and transportation structures.
Before joining Bechtel, Mr. Katanics worked for various consulting engineering companies in Europe i and in the U.S. His experience included structural l design and construction supervision of power plants, mining facilities, high-rise buildings, port facil-ities, off-shore structures, steel, reinforced con-crete and prestressed concrete bridges.
4 e
0 4
'k
_ .- _ _ ~. . __
t .
PROFESSIONAL QUALIFICATIONS OF THE0DORE E. JOHNSON I
)
F POSITION Chief Civil / Structural Engineer ,
\
EDUCATION BS, Civil Engineering, Ohio University; MS, l 4
Applied Mechanics, Michigan State University 1 PROFESSIONAL Registered Professional Civil Engineer in California and a
DATA Minnesota; Member of the design working and subgroups, and the main committee of the ASME Section III Division 2, Code for Nuclear Concrete Reactor Vessels and Containments; j
Member of the Committee on Nuclear Structures and Materials, American Society of Civil Engineers; Division Coordinator j for the 4th and 5th International Conference on: Structural i Mechanics in Reactor Technology, Loading Conditions and Structural Analysis of Reactor Containment; Author or co- ;
- author of publications relating to Nuclear Concrete Containment Design. l l
l 4
SUMMARY
Recent Assignment: Chief Civil / Structural Engineer I 7 years: Group supervisor l
l 3 years: Lead engineer l
- 3 years
- . Engineer j 1 year: Structural dynamics engineer ;
EXPERIENCE As Chief Civil / Structural Engineer for the San Francisco Power Division, Ann Arbor Area Office, Mr. Johnson is responsible for I
- the following: establishes engineering design standards, guides and criteria; ensures technical development and training of personnel; monitors job performance of personnel; assigns personnel to specific projects; monitors projects for adherence to corporate and division standards; and provides guidance and instruction to achieve quality objectives.
4 l
As group supervisor of the Special Structures Group for Bechtel Power Corporation's San Francisco Power Division, Mr. Johnson was i responsible for development and standards for containment structural j
design, seismic analysis, structural analysis, and computer I applications. Mr. Johnson managed many test programs related to the following areas: containment liner plate anchors and steel embedments; large prestressing tendon end anchor regions; tornado missile impact; and pipe rupture restraints.
Before this assignment, Mr. Johnson participated in the design of the Arkansas Nuclear One reactor bu.'iding as lead engineer. He was responsible for the design, analysis, and preparation of specifi-cations and engineering drawings. As an engineer for the-containment group, he worked on the development of design techniques for pre-4~
stressed concrete containments with major emphasis on the steel liner plate. As an engineer with Bechtel Corporation's Mining and Metals Division, Mr. Johnson was responsible for determining noise and vibration control requirements and structural design for a Taconite project.
4 T11E0DORE E. J0l!NSON Resume Page 2 EXPERIENCE Before jcining Bechtel, Mr. Johnson .was with the Astronautics (Cont'd) Division of General Dynamics. As a structural dynamics engineer he participated in studies aimed at determining Atlas-Centaur missile response due to wind loadings.
l 1
l l
1
\ i f
... u u -.w. w - . i : m u . .. ~ . - ~
m - - + u a ..w . L .., . . . . . - _ . .
O PROFESSIONAL QUALIFICATI0tlS of RICHARD C. AtlDERS0tl PRESEtlT POSIT 10fl Engineering Manager, Bechtel Power Corporation EDUCATION BS, Mechanical Engineering, University of California, Berkeley PROFESS 10flAL Registered Professional Mechanical and Nuclear Engineer, DATA California; Member, American Nuclear Society SUMtiARY 1 year: Engineering manager 6 years: Chief nuclear / environmental engineer 1 year: Assistant project engineer 2 years: Project engineer 3 years: Mechanical supervisor 1 year: Senior engineer 1 7 years: Mechanical engineer l EXPERIENCE Mr. Anderson is the engineering manager for Bechtel's nuclear and fossil fuel projects in the North West and all project work and consulting services for Japanese clients.
Before being assigned to his present position, Mr. Anderson was the chief nuclear / environmental engineer responsible _ ;
for all nuclear and environmental engineering for the San Francisco Power Division of Bechtel Power Corporation.
Previously, he was assistant project engineer in charge of all technical work for the two 1100 MWe Mendocino nu-clear power plants for Pacific Gas & Electric Company.
He was also the mechanical supervisor, and later the pro-ject engineer on the 545-MWe nuclear generating plant at Monticello, Minnesota, for Northern States Power. This assignment included supervision and coordination of mechan-ical design, specifications, and procurement activities and later the supervision of all engineering activities includ-ing the quality control and quality assurance programs.
As a senior engineer and mechanical subgroup supervisor, he was responsible for " balance of plant" systems and equipment for the Tarapur nuclear-power station in India.
He was a mechanical engineer and the systems analysis leader on the FARET Fast Reactor Teit Facility for Arizona Nat-ional Laboratories; mechanical. engineer on the emergency cooling shield for the Peach Bottom HTGR nuclear reactor; systems engineer on the Hallam nuclear power facility and
-a mechanical engineer on the Dresden nuclear plant.
)
U PROFESSIONAL QUALIFICATIONS of WILLI Ali H. WHITE PRESENT POSITION Engineering Specialist, Bechtel Power Corporation EDUCATION BS, Civil Engineering, University of Idaho; MS, Civil Engineering, University of Colorado; PhD, Civil Engineer-ing, University of Colorado PROFESSIONAL Registered Professional Engineer, Oregon; Member, Ameri-DATA can Society of Civil Engineers
SUMMARY
2 years: Engineering specialist 2 years: Structural engineer 6 years: Assistant professor 1 year: Senior engineer EXPERIENCE Dr. White has been employed as an engineering specialist.
With Bechtel's San Francisco Power Division for almost two years. He is assigned to the chief civil engineer's staff where he is responsible for seismic analysis. l
[
Earlier, Dr. White was a structural engineer with the Ten- l nessee Valley Authority for two years where he was respon- l sible for seismic analysis of all Category I structures l for a twin-unit nuclear power plant, including the seismic i input for the design of the nuclear steam supply system. I 1
From 1968 to 1974, Dr. White was an assistant professor at Oregon State University where he taught undergraduate and I graduate courses in structural mechanics and analysis and computer applications.
Prior to joining Oregon State, Dr. White was employed at the Bettis Atomic Power Laboratory for a year. At Bettis, he was a senior engineer working on shock analysis of nu-clear reactors aboard submarines and was involved in pro-grams to assess the shock resistance of reactor internals.
s l
r^%
N ,) .
RESPONSE OF TROJAN NUCLEAR PO'w'ER PLANT CONTROL BUILDING TO SPECl?!ED SSE EVENT September 20, 1978 Myle J. Holley, Jr.*
and Boris Bresler**
- 1. INTRODUCTION The writers were engaged by the fir = of Lowenstein New=an, Reis
& Axelrad, outside counsel for Forciand General Electric Company, on August 4, 1978, to assis't in determining the capability of the Control Building of the Trojan Nuclear Power Plant to resist the specified SSE event withou: any structural consequences which could interfere with a safe shutdown.
In the course of studying this question over the past six weeks, the writers had many extensive discussions with engineers of the Bechtel Corporation, both a: their San Francisco offices and at the offices of the Portland General Electric Co=pany. PGI personnel participated in many of these discussions. Familiarization of the writers with the subjec:
st.1cture, with methods of analysis being used by the engineers to e c.d*aate its capacity, and with pertinent results obtained from those analyses, was facill:ated by the ce=plete cooperation extended to then by the engineers of both organizations.
- Principal, Hansen, Holley & Biggs, Inc. , Cambridge, Mass.
- Principal, Wiss, Janney, Eistner & Associates, Inc., Northbrook, Ill.
l
m-
~,) .
The writers visited the Trojan Plant and : cured the Control Building, personally observing its physical layout and construction features. This provided a useful supplement to information obtained by examination of the relevant design drawings and through discussions wi:h the engineers.
The writers have discussed the problem on =any occasions during the past six weeks, apart from the above-mentioned larger =eetings.
M. J. Holley also has discussed certain aspec:s of the problem with his professional associates, Dr. Robert J. Hansen and Mr. John M. Biggs, and B. Bresler discussed the problem of wall capaci:1es with his associate, Mr. Craig Comarrin.
It should be noted that the writers' efforts have not included any attempt to check the accuracy of the analyses executed by engineers of the Bechtel Corporation. Such checks could not have been accomplished in the time available. Rather, the writers focused on =ethods of analysis used in the evaluation and upon those results of analyses with which : hey were l 1
provided. 1
)
The report which follows is comprised of discussion of relevan:
1 aspec:s of the evaluation and, finally, the writers' Su==ary and Conclusions.
As noted above, in the course of this investiga: ion the writers discussed all aspects of the evaluation and are in full agreement on the results of the evaluation. In preparing this repor:, however, M. J. Holley is primarily responsible for sec: ions dealing with dyna =le analysis, slab-to-wall forces, non-linear response, da= ping consideration; and 2
f . .
' ' N_' '
displacement considerations; 3. 3resler is pri=arily responsible for the section dealing with shear wall capacities.
2, I ME* HODS OF DYNAMIC ANALYSIS OF THE CONTROL-Al'CLIARY-FUIL BUILDING COM?LE l l
Of the three different models used for dynamic analyses.(sticks, 1
{
TA3S, STARDYNE) only STARDYNE, based on finite element modeling of the l
structural components, accounts directly for all aspects of the distributed i stiffness and distributed masses. Therefore, the STARDYNE analysis provides l
the most reliable prediction of the principal mode shapes. Further, since the' STARDYNE analyses were for linear elastic response based on =ax1=u=
i possible (uncracked) stiffnesses, they predice useful upper-bcund esti=ates )
of internal forces for the principal structural component:. This follows from the fact that the fundamental period (0.15 sec) is in a region of the !
spectrum for which any increases in period (due to cracking-reduced stiffnesses) do not increase the accelerations.
- 3. SHEAR WALL CAPACITIES - -_ _
Cacaciev Criteria. --.
The earthquake resistant shear walls in the Control and Auxiliary buildings f all into three categories:
- a. Masonry block walls which " sandwich" structural steel framing I (columns) and a reinforced concrete core, I
- b. Masonry block walls which " sandwich" a reinforced cenerete core'only, and
/~N
\_) -
- c. Two-wythe concrete block walls withou: a concrete core.
Strue: ural beha"ior of " sandwich" valls (s:rength and load-defor=a-tion characteristics) can be deterrined approxi=ately from behavior of reinforced concrete or reinforced =asonry walls, as there is no experimental data on such walls.
Capacity criteria were established for the following principal modes of>failura:
- a. shear-compression (diagonal) failure
- b. flexure failure
- c. shear-friction or dowel action failure The lowest value is used as the capacity of a given wall.
The shear-comoression criterion has been derived by Bechtel engineers (1,2}
from their evaluation of tes:s on reinforced tasonry walls. While :he tes: .
specimens do no:
s1=ulate the ac:ual wall construe:1on encountered in the subjec:
building, there is no other data that would better si=ulate :his type of cons:ruction.
The interpretation ef the data in deriving the Bechtel "3asic Criteria" appears to be conservative, particularly as the test specimens use concrete block, =or:ar, and grou: having lower strength values than those used in the subject building.
Also, while data on the effec: of cyclic loading is li=1:ed, tes:
results reported by Mayes, et, al. show that cyclic loading which does not e::ceed a maximum shear strain of about 0.006, does not lead to redue:1on of strength below about 210 psi. The eaximum shear stresses based on calculated forces for the S-S valls all fall below these values.
_4 f
- 1. Schneider, R. R. , " Shear in Concrete Masonry Piers", California Sta:e Polytechnic College, Pomona, 1969.
- 2. Mayes, R. L., Clough, R. W., Hidalgo, P. L., and McNiven, H. D.,"Seis=1e l Research on Multistory Masonry Buildings", North American Masonry l Conference, Boulder, Co., August 14-16, 1978.
l
- /g . .
\.) '
i t
The flexure capacity criterion is based on generally accepted and documented formulations of ultimate moment capacity. The shear frictice or dowel action is used to evaluate vertical shear transfer capacity j'
'from the side walls to the-and walls. 1 1
Maximum Forces on Shear Walls.
i t As noted above, STARDYNE~ analyses predicted upper-bound forces t
- to be expected for linear-elastic response, assuming unli=dted shear wall capacities. In a few of the smaller shear walls, that is, walls which i !
contribute only a small portion of the - cotal story shear resistance, these !
predicted forces exceed the computed wall capacities. Alse, the predicted a maximum shear force in one of the masonry walls (the West wall) exceeds the
.+
computed (dowel) capacity. This implies that, for these particular shear i
! walls, maximum shear forces may equal the computed shear capacities. It i
further implies that the other walls may experience maximum shear forces a
which are slightly greater than the STARDYNE~ elastic response values, as j y
they accept the small excesses of predicted shear forces above the ca:acities 1
of the cited walls.
1 In sum =ary, the STARDYNE-predicted wall shear forces modified as a
described above, represent conservative maxima.
A
! Evaluation of Shear Walls. --
The critical shear walls in the Trojan Control and Auxiliary 4
i buildings under SSE event are the N-S walls at elevations 45 and ol. Therefore, j the following remarks are addressed to these walls, and while.some other 1
valls (at other elevations) may'also be highly stressed the considerations
, - - , . . , w , , ,, s, , - - -- - t
/~')
%/ -
described below would apply to the other walls as well.
In evaluating perfor=ance of these walls under a SSE event, the following approach was used:
For each wall element strength (capacity) has been calculated using criteria discussed above. These capacities were co= pared with the forces calculated using STARDYNZ linear elastic analysis. As noted above, redistribution of the forces, which would take place when some of the walls reach their capacities, results in so=ewhat increased shear forces in those walls which have substantial reserve capacities. The st= of the computed capacities of these shear walls is well above the sum of the predicted maximum shear forces. Consequently, the group of walls resisting a given story shear can withstand forces imposed during a SSE event without interfering with the safe shut down.
4 SLAB FORCES AND SLAB-TO *JALL FORCIS -
The focus of this report is pri=arily upon the behavior of the casonry shcar walls. It is obvious that the capacities of the floor and roof slabs to resist the in-plane forces i=pesed by the SSE condition is equally i=portant. However, these components are = ore conventional than the shear walls, and evaluation of their capacities is somewhat more straight forward.
The STARDYNE-predicted forces in the slabs represent reasonable upper-bound estimates. It is the writers' understanding that the slab capabilities for resisting these forces have been thoroughly investigated.
() . -
l l
It is further understood tha: these investigations have shown the capacities of the slabs, and the slab-co-wall connections, are adequate c resist the STARDYNE-predicted forces. '
r
- 5. NON-LINEAR SHEAR WALL RESPONSE As noted above, the STARDYNE linear analysis, based on maximum (uneracked concrete) stif fnesses-provides an adequate predic:Lon of the maximum force i= posed on the shear wall group and on the slabs by the SSE condition. However, some of the walls-are subjected to shear loads equal to their computed capaci:les, and the other walls are subjected'to shea icads equal to substantial fractions of their computed capacities. This implies that the walls will be distinctly non-linear in their response.
Under cycled loading their no-load stiffnesses will be less than the initial no-load '(uneracked)'stiffnesses used in the STARDYNE analyses. Moreever, within each leading cycle the stiffness must decrease with increasing load. 1 The slabs will experience some cracking and this cracking will reduce slab stiffnesses below (uneracked) stiffnesses assumed in the STARDYNZ i analyses. However, the effec:s of cracking on slab stiffnesses should be much less than the corresponding effects in the masonry walls. For this reason, the non-linearity of overall structural response is dominated by the wall non-linear responses.
The engineers have recognized : hat the STAROYNE analyses underestimate displacements, for the above described reasons. They have esti=a:ed a
_7 e
/T l
\) l
=aximum displacenen: (at roof level of the control building) of about 0 9",
in contrast to the corresponding STARDYNE-predicted displacement of 0.15".
Their es:imate is based on shear strains developed in shear wall test specimens, when such specimens reached their shear capaci:1es.
The writers have used a different me: hod :o estinate maxi =u:
displace =ents. This involved approxi=ating the non-linear force-displacement func:1on by a linear system of greatly-reduced stiffness. If the structure wi:h this reduced linear stiffness characteristic is assumed capable of developing the shear resisting forces predicted by STARDYNI :he corresponding increased displace =ents can readily be predicted from the specified SSI response spectru=.
By this approach, an assumed stiffness redue:1on to abou: one-fifth of the stiffness used in the STARDYNI analysis would lead to the same sir.-fold increase in displace =en:s i= plied by the engineers' d1splace=en: estimate.
This 80 percent reduction in linear stiffness appears to the vricers to be a conservative representation of the degraded non-linear s:iffness of the ac:ual walls subjected to cycled loading. It may be noted that for a 90 percen:
3:iffness redue:1on (1. e., effec:1ve stiffness only one-tenth of tha: used in the STARDYNI analysis) would lead to an eight-fold increase in the STARDYNI-predic:ed displacement, :ha: 1s, a maximum displace =en: of abou: 1.0 inch.
To further illustra:e the inherently-l1=i:ed nature of :he maxi =um displacement the writersmodified the above-described reduced-linear-stiffness analysis as follows. The reduced-linear stiffness model was assumed to yield at a force level only 5(1 percent of the STARDYNI-predic:ed i= posed force.
In the range of the response spee: rum of interes: this would lead :o only a 1
'\ I
%-]
I 25, percent additional increase in the maximu= displace =ent.
In su==ary, it is the writers' judg=en: : hat, based on the STARDYNE-predicted displace =ent of 0.15", the ac:ual =axi=u: displacement, at roof level, should be little, if any, in excess of 1.0 inch. It is l
i obvious that the corresponding =axi=u= floor-to-ficer relative displacemen:s would be less than this value.
i j
- 6. DAMPING CONSIDERATIONS In all their analyses of forces in the s: rue: ural componen:s, and displacenents, the engineers have assu=ed 5,percen: damping for responses to the SSE condition. The writers si=ilarly have assu=ed only 5 percent damping. l This is reasonable for the reinforced concre:e floor slabs for which it is reported that in-plane capacities have been found to be well in excess of the i= posed forces. However, for the casonry shear walls, subjec:ed to substantial fractions of their compu:ed capaci:1es, substantially larger da ping percen:ageswould be appropriate. Such larger damping would lead to s= aller predicted forces and displace =ents. The use of these, more realistic, larger damping percentages is not per=1::ed by the NRC, but they represent an un-accoun:ed-for conservatis: in all the analyses.
- 7. DISP 1.ACEMENT-DE?ENDENT CONSIDEEATIONS - _ - _ . - - _-
It should be emphasiced tha: judgments regarding the capaci:y of :he Control Building to resis: the SSE event without i=pairment of 1:s safety
_g.
i i
- . functions are almost entirely dependent on the anticipated max 1=um 1
displacements. Thus, magnitudes of anticipated internal forces are nec
- _ nearly as important in themselves as in the basis they previde fer evaluating 1
3 the maximum horiconcal displacements. This would not be che case in an 1
t l i
t evaluation of the capacity of the structure to resist a sustained, scacic, horitontal loading. For response to earthquake ground motion, a range of i
assumed strengths all may lead to displacements which are acceptable.
i
- It is the writers' understanding that the structural steel framework i
of the Control Building was designed to carry the entire gravity leading of l
(
j the building. This is of considerable importance. It implies that collapse a
of the building, or any of its slabs, could only occur if the anticipated heritoncal displacements were extremelv large; i. e.. large enough to totally
(
J !
eliminate lateral stability afforded by the walls, or large enough to fail a j beam-to-colu=n connection. Neither of these effects could be realited unless i i
che maximum horizontal displacements were many times as large as the maximum
- j. displacements that actually are anticipated. !
In short, collapse of the s building, or one of its slabs, does not appear to be a possible consequence
, of the specified SSE event.
More realistic displacement-related considerations which recuired
- address are
- a. capability of equipment to accommodate relative displacements (e. g. , floor-to-floor, or building-to-building) vichout loss of function.
l i
L()
- b. inplications of Con:rol Building displace =ents on the forces -
l transmi:ted to the hold-up tank enclosure and to the spen:
fuel pool structure.
I i
4
- c. possible contacts of the Con:rol Building with craneway columns of the Turbine Building as these two buildings 4
respond, independently, to the SSI event.
It is the writers ' unders:anding tha: the capabilities of all safety-rela:ed components ( e. g., equipment, piping, elec:rical runs) to accept large relative displace =ents have been deter =ined by an exhaustive
- on-site survey. This survey, executed by a tea = of engineers skilled in j the appropriate disciplines, and including supple =entary analyses where i l
1 required, is reported to have shown that relative displacenents in excess of those anticipated can be accocmodated by all safecy-related componen:s. l The survey is also reported to have disclosed tha: possible i=pingement of a piece of spalled =asonry block on any of the components, (however 1
unlikely), would not cause such component to fail to perform its intended function.
i The engineers' analyses show that the larges hori: ental displace =ents i
of the Control Building, in response :o the specified SSE event, would occur in the N-S direction. The STARDYNE analysis for this.cendicion leads to forces.in the reinforced concrete structural components of the hold-up tank i
enclosure and spent fuel pool (connected, by floor and roof slabs, through I I
the Auxiliary Building to the Control Building) which are well belov :he i I
capacities of these strue: ural ce=ponents. However, as has been noted, the 1
I l
. __ __ . -.a
i
?
- STARDYNE analyses do not account for the increased flexibility of the Control Building, arising from the reduced stiffness of the masonry walls under cycled loadings to large fractions of their calculated I
- capacities.
The engineers have evaluated :he maxi =um forces which can be imposed on the hold-up tank enclosure and spent fuel pool by the slabs.
- These imposed forces are limited by the in-plane bending moment capacities [
I of the slabs at their junctures with the fuel building. The slab momen:
s {
I capacities are limited by yielding of their E-W reinforcing steel. The results of these analyses are reported to demonstrate that the maxt=u=
forces that can be imposed on the hold-up cank anclosure and the spent fuel pool, regardless of the magnitude of N-S displacements of the Control Building, are well below the capacities of these fuel building structures.
t At roof eleva: ion of the Control Building, where the anticipated N-S hori: ental displacemen:s are maximum, a 2" clearance has been provided 3
1 between the Control Building and the craneway colu==s of the Turbine Building.
The engineers have executed an updated analysis of the Turbine Building response to the SSE event.
This analysis is reported to predict a maxinu=
5-S displacement of the Turbine Building colu=ns of 1.5". Since, as noted in an earlier Section of this report, the maxinum anticipated Centrol Building deflection is, in the writers' judgment, only about 1.0" (0.9" by the engineers '
estimate), the worst possible relative displacement of the two' buildings, 2.5",
is within the clearance, 3.0", which has been provided.' However, i: is conservative to base a prediction'of-the maximum building-to-building relative 3
(v displacement by direc: addition of their individual =aximum dis:lacements.
If, for example, the individual building displacemen:s were ce=bined on an SSRS basis, or if all the modal contributions :c the displacemen:s of the two buildings were co=bined on an SRSS basis the i=clied clearance =argin would be greater. In su==ary, it appears that contact between the rwo buildings is not to be anticipated.
- 8.
SUMMARY
AND CONCLUSIONS l
- a. Because of the presence of a steel frame designed to support all gravity loads and because of the small magnitude of l
displacements, nei:her collapse of :he Control Buildings nor I collapse of any of the structural members supper:ing its floors 1
can occur in the specified SSE even:.
l l
- b. In judging the ability of the Con:rol Building to resis che specified SSE event w1:hout censequences which eigh: interfere wi:h safe shutdown. predicted =ax1=u= displacements are the aspec: of response of paramount significance. Cocpu:ed internal forces are of significance only to the extent that they form a basis for judging the displacements.
- c. A necessary first step in the precess of computing displace =en:
maxima is a dynamic analysis of response in the linear elastic i
rN .
c q.) -
h range. The STARDYNE program is an effec:ive cool for this 4
purpose because it per=Ats a good representation of the = ass distributions and s:iffnesses throughouc the strue:ure. The designers properly utilized the STARDYNE analytical resul:s 4
rather than chose obtained from the stick-models or TABS analysis.
- d. Because of :he region of the response spec: rum in which the l
t significant natural frequencies are located. because of :he
-l linear-elastic behavior assu=ed, and because of the conservative damping ratio used. the STARDYNE determined story shears a
(for the group of resiscing walls) represent censervative, unper-bound, values of these group forces. Ic is significanc to noce that there is a conservative =argin between :hese group 1 j
l forces and the corresponding group streng:hs of the resisting l shear walls.
e.
The individual shear wall forces in some walls, particularly the smaller walls.have computed capaci:ies less than the STARDYNE predicted values of imposed forces. This simply means tha:
forces in these particular walls may reach their computed capacities, and that forces in other, less severely loaded, '
walls may develop forces larger than those predic:ed by the
-STARDYNE linear-elsstic analysis. Hevever, re-distribu-ion analyses by the Bechtel engineers show that the other-major walls would F ,
i T
_ _ d); . , - - +-
i.
_ , ,_ _l. i . . _ .
2--
I s\ .
% .) -
develop forces which are st:11 well below their capacities.
Consequently, the group of walls resisting a given story shear can withstand forces imposed during a SSE event without interfering with safe shut-down.
- f. Based upon wall shear strains at their capacity loads, reflecting the non-linear response of shear wall test specimens under cycled leading, the engineers have estime:ed a maximum horizontal deflection of 0.9". This deflection (at roof level, at the west end of the Control Building) is six ci=es the
, displacement outputted by the STARDY3E analysis. The latter estimate assumed stiffnesses based en uncracked concrete, 1. e., stiffnesses which are representative at very low levels of loading.
I 8 By an independent approach, approximating the non-linear 1
characteristics of shear walls with reduced-stiffness linear characteristics, the writers satisfied themselves that 3echtel engineers' estimate of maximum horicental displacement is conservative. However, for purposes of evaluating the consecuences of displacements, they prefer the slightly more --
conservative conclusion that the maximum deflection vill not exceed about 1.0".
l
1 l
- h. Relative displacements (c. g. floor-to-floor) will, of course, be less than the maximum value cited above. However, a survey carried out by 3echtel engineers has disclosed tha: all safety-related components can acco=modate. relative displacements well in excess of 1.0" without loss of fune:1on. The floor slabs which connect Control Building, Auxiliary Buildings, and Tuel Building are limited in their capacity to impose forces on the Hold-Up Tank enclosure and Spen: Fuel strue:ures.
Analyses by the Bechtel engineers show that these limi:ed forces, which are independent of the magnitude of Control Building displacements, are well within the capaci:ies of the structures in :he Fuel Building.
- 1. Up-dated dynamic analyses of the Turbine Building response :o the SSE event show that the displacements of this structure combined with Control Building displacements are not sufficient to cause contact between the two buildings.
j.
From all the above 1: is the writers ' j ud gment that the Trojan Control Building, in 1:s as-built condition, can withstand the specified SSE event with no consequences that could interfere with safe shutdown, i
1 l
j
(O NAME: Boris Bresler ADDRESS: Bay Bridge Office Plaza, Suite 485 5801 Christie Avenue Emeryville, California 94608 4
TITLE: Senior Consultant and Manager of California Office, Wiss, Janney, Elstner and Associates, Inc.
L I
EDUCATION: BS in Civil Engineering from the University of California, Berkeley in 1941, . and a MS in Aeronauti-cal Engineering in 1946 from the California Institute ;
of Technology, Pasadena. I PROFESSIONAL AWARDS AND AFFILIATIONS: In 1959 Mr. Bresler was awarded the Wason Medal for Research by the American Concrete Institute; in 1961 l
he held a National Science Postdoctoral Fellowship; !
and in 1962, a Cuggenheim Fellowship. In 1968 he was '
awarded State-of-the-Art of Civil Engineering Award by the American Society of Civil Engineering. He is a Fellow of the American Concrete Institute, served on .
numerous technical committees during the past 30 years, l and was a member of the Board of Directors from 1970- !
1973. Also, he is a Fellow of the American Society i of Civil Engineers, and from 1972-1976 served on the l Executive Committee of the Structural Division. He !
is a member of the American Society for Testing and Materials, Structural Engineers Association of Northern !
California, Reinforced Concrete Research Council, and l International Association of Bridge and Structural Engineers (member of Working Commission III). He is l 8 Registered Engineer in the State of California.
l EXPERIENCE: !
Boris Bresler has been engaged in teaching, research and consulting in structural engineering since 1947. His l primary field of interest has been behavior and distress l in structures exposed to complex loading conditions and various special environments. In his recent research, he has developed advanced analytical methods for analyz-ing response of reinforced concrete structures to fire and other extreme temperature .ovironments. He has also been in the forefront of developing systematic procedures for evaluation of seismic hazards in existing buildings.
Mr. Bresler has been a member of the faculty of the Department of Civil Engineering at the University of California, Berkeley since 1946, and is currently on leave of absence from the University. He is author of more than forty technical papers, is co-author of
" Design of Steel Structures" published by John Wiley and Son, and is co-author and editor of " Reinforced Concrete Engineering" published by Wiley-Interscience.
i l
I.
p
- b Sept. 1978 PROFESSION R QUALIFICATIONS OF MYLE J. HOLLEY, JR.
Myle J. Holley, Jr. , received the S.B. and S.M. degrees from MIT in 1939 and 1947 respectively. He joined the MIT Faculty in 1946 and retired on Sept. 30, 1974. While on the faculty he taught subjects in structural analysis and design, and supervised structural research projects. The latter j included work in the fields of massive reinforced concrete structures, pre- l stressed concrete, st'uctural r applications of granite, high-strength rein- l forced concrete beams, and the performance of thin arch concrete dams. For l several years he was in charge of the Structural Division of the Civil Engi-neering Department.
Mr. Holley is a registered Professional Engineer in the Commonwealth of Massachusetts. He is a member of the American Society of Civil Engineers, American Concrete Institute, and the American Society for Engineering Educa-tion. He has served on numerous professional committees, and currently he is a member of ASME/ACI Committee on Concrete Components for Nuclear Containment l Structures, and of the Seismic Advisory Committee of the Massachusetts Building l Code Commission. He also served far several years on ACI 348, Structural Safety, and ACI 349, Concrete Nuclear Structures, and currently is an Associate of each of those committees.
Prior to joining the MIT faculty Mr. Holley acquired several years ex- !
perience in the analysis and design of heavy machinery. From 1941 to 1946 he was employed by the S. Morgan Smith Co. (now the York, Pa. Division of Allis- !
Chalmers Manufacturing Co.). In their employ he was a principal participant in the Smith-Putnam Turbine Project, involving the design, construction, and operation of the world's largest aero-electric plant. While employed by the S. Morgan Smith Co., he also was a principal stress analyst in the design of a 100,000 HP compressor for a major wind-tunnel facility for NACA, as well as the stress analyst in the design of a large human centrifuge. His responsibil-ities also included the stress and dynamic analysis of large hydraulic turbines.
In the period from 1947 to 1952 he was employed by Fay, Spofford, and Thorndike as a designer, and later as a structural consultant. As such he assisted in the design of numerous structures including a very large aircraft hangar, steam power plant, elevated highways, etc. Since 1952 he has served as struc-tural consultant to a number of engineering firms, insurance companies, the Commonwealth of Massachusetts, and others.
Mr. Holley has served as a structural Consultant to Stone and Webster Engineering Corporation in their designs of several nuclear power plants, including the plants at Rowe, Massachusetts Haddam, Connecticut (Connecticut Yankee)
Wicasset, Maine (Maine, Yankee)
In addition he serves as a member of Stone and Webster Engineering Corpora-tion's Design Review Boards on their Reference Nuclear Power Plant, the Jamesport Nuclear Power Plant, and Millstone No. 3 Nuclear Power Plant.
Mr. Holley has served as a structural consultant to American Electric Power Co. in their design of the Cook Nuclear Power Plant #1, and has pro-vided advice in relation to problems crising in the construction of that plant.
g .
U' _2 Mr. Holley served as a structural consultant to Rochester Gas and Electric Company during the design of their Ginna Nuclear Power Plant.
Mr. Holley served as a structural consultant to United Engineers and Constructors on their designs of the Brunswick, Indian Point #3, and WPPSS Nuclear Power Plants.
Mr. Holley served as a structural consultant to Gibbs and Hill in their design of the Aguirre Nuclear Power Plant, Unit No.1 for ENEL.
I 1
4 J
(~
\~/
1 UNITED STATES OF AMERICA l NUCLEAR REGULATORY COMMISSION l
l l
BEFORE THE ATOMIC SAFETY AND LICENSING BOARD In the Matter of )
) Docket 50-344 PORTLAND CENERAL ELECTRIC COMPANY )
et al ) (Control Building Proceeding)
) ,
, (Trojan Nuclear Plant) ) l l
TESTIMONY OF CLEN E. BREDEMEIER My name is Glen E. Bredemeier. My title is Vice President, Power l 1
Operations. A statement of my qualifications is attached. The follow- l l
ing discussion describes, from a Power Operations standpoint, the conse- '
quences of Trojan not being allowed to operate.
l The Trojan Nuclear Plant has a demonstrated capacity of 1080 MW. As a 67.5 percent owner of Trojan, Portland General Electric Company (PGE) l is entitled to 729 MW of ent;gy when the Plant is operating at its peak. l Taking such matters as temporary outages into consideration, PGE normally l expects to obtain'an operating average of 620 MW of energy production 1
from Trojan. The current incremental cost of such energy is approxi- l mately 3 mills /kWh.
4 Our initial analysis of the power supply situation (June 1978), consider-ing that Trojan was shut down, anticipated a termination of the supply of surplus hydroelectric energy during July as the annual spring runoff subsided in the region. Replacing Trojan's energy supply was expected to result in added costs which would soon escalate to $7 million per month and which would total over $40 million by the end of 1978. l l
The situation as it has developed to date has indeed brought excess power costs, but for ceveral reasons, those costs have not been as great as at first anticipated. First, the spring runoff situation persisted substan-tially longer than predicted, resulting in the availability of significant
q Y./
quantities of surplus hydro energy into early October. At 3 mills /kWh through Au: gust 31 and 3.5 mills /kWh thereaf ter, this surplus hydro energy has not cost much more than the approximately 3 mills /kWh incremental cost of Trojan production.
In addition to the relief provided by the availability of surplus hydro energy, PCE has been able to obtain several commitments for supply of power which, while significantly more costly than Trojan's output, are nevertheless less costly than original estimates of probable power supply cost. We have entered into two contracts for purchase of energy at a cost of 18 to 22 mills /kWh delivered. We have also arranged to borrow, or for the option to borrow, substantial quantities of energy with repay-ment in kind scheduled for dates as far in the future as the 1982-83 operating season. The cost of energy to repay these oblig.itions would appear to vary from about 6 to 18 mills /kWh. These several arrangements along with the probable continuing availability of some " spot" purchases should greatly minimize any expected operation of oil-fired combustion turbine generating units where incremental generating costs would be in the range of 28 to 35 mills /kWh.
Despite the improved situation described above, PGE's actual excess power costs to meet load requirements were $3 million in August and about $2 million in September. We anticipate additional excess power costs of $18 million for the period October through December. A deteri-oration in the present situation could result in greater costs. Added costs will persist into 1979 if Trojan is not restored to service.
i
. . . .--- . . . . . -. - . - . _ . . - . .~ -
{i f NAME: Glen E. Bredemeier TITLE AND DUTIES: Vice President, Power Operations - Responsible for '
l System Load Dispatching, scheduling of generation and j' power interchanges, pooling and coordination, power contracts, fuel planning and procurement, mining
) and exploration, hydrogeneration and communication.
- EDUCATION
- OSU - BS in EE (1942); Stanford U. - Executive i
Development Program (1966); NF - OSU (1968);
INP - NUS (1969).
. PROFESSIONAL AFFILIATIONS: PE - Oregon; PE0; NSPE; lEEE - Senior Member; NELPA; Tau Beta Pi, Eta Kappa Nu, Sigma Tau Engineering Societies.
EXPERIENCE: General Electric Company - Test Engineer (1942-1943);
U. S. Navy - Radar Countermeasures Officer (1943-1946);
Portland General Electric Company - Electrical Drafts-man (1946-1948); Assistant to Superintendent of Production (1948-1953); Intercompany Pool Engineer -
Power Pool Operations (1953-1961); Manager, Power i; Operations Department (1961-1975); Vice President, Power Operations (1975- ).-
Y\
.. ]
O l
UNITED STATES OF AMERICA NUCLEAR REGULATORY COMMISSION BEFORE THE ATOMIC SAFETY AND LICENSING BOARD In the lhtter of )
) l PORTLAND GENERAL ELECTRIC COMPANY, ) Docket No. 50-344 ET AL. ) (Control Building)
)
(Trojan Nuclear Plant) )
TESTIMONY OF HECTOR J. DUROCHER I am the' Assistant Administrator for Power Management of the Bonneville l Power Administration (BPA). In this position, I am responsible for :
administration of BPA functions respecting the marketing, exchange, and delivery of electric power.
BP/ is a regional marketing agency within the Federal Department of Energy. It markets energy produced by 30 Federal hydroelectric projects constructed and operat.ed by the Corps of Engineers and the Bureau of Reclamation in the Pacific Northwest. In addition, BPA has acquired rights to and markets output from the City of Eugene, Oregon's 30 per-cent share of the capability of the Trojan Nuclear Plant, the Hanford generating plant of the Washington Public Power Supply System (WPPSS) and, when completed, will market output from all or parts of three additional nuclear plants under construction by WPPSS.
i Page. l'- Testimony'of Hector J. Durocher a .- i
)
OL With Trojan not operating, utilities in the Pacific Northwest will have '
difficulty in. satisfying their total power loads. BPA has determined that it cannot satisfy its total loads.
l To satisy its firml/ power commitments, BPA will use power from the !
Hanford Nuclear Plant. Due to the closure of Trojan, BPA has withdrawn from its industrial customers the power produced by the Hanford Nuclear Plant formerly contracted to industries on a withdrawal basis. The i incremental cost of power produced by the Trojan Nuclear Plant is about
! 3 mills.per kilowatt-hour while the cost to recipients of energy produced
- by the Hanford Nuclear Plant is expected to be 12-14 mills per kilowatt-I hour. This means that BPA will incur an additional cost of 9-11 mills per kilowatt-hour to supply the 2.0 billion kilowatt-hours to purchasers of firm power the could have been produced by BPA's share of the Trojan l Nuclear Plant during the July 1, 1978 - June 30, 1979 year. This represents an additional cost-to BPA of $18-22 million, which BPA will have to pass on to its customers.
t t
1/ BPA must satisfy iti~ firm power commitments without regard to streamflow conditions on its system, the operating capabilities of its generatina P ants, l or other factors except major natural forces. BPA also supplies nonfirm power, but only when it.is assured of power in amounts greater than that needed to supply its firm power obligations..
LPage 2 Testimony of Hector J.' Durocher r- . - - - ,.
e- - - ,w,.- , ,, , nwy
('~\
l V .
l The notice of withdrawal of the Hanford power has compelled BPA industrial customers to seek and acquire power from other sources at costs higher ,
than the cost of power that could have been supplied to them by the Hanford Nuclear Plant. Energy produced by the Hanford Nuclear Plant would have cost these industries 12-14 mills per kilowatt-hour. To replace power from Hanford, industries to date have arranged to obtain energy from the following sources:
SOURCE MEGAWATT-HOURS MILLS PER KWH TOTAL COST Montana Power Company 100,000 21.84 $2,184,000
-West Kootenay P&L Co. 418,700 17.63 $7,381,681 This power represents only a portion of the 2.25 billion kilowatt-hours which could have been supplied by Hanford. BPA expects that the additional power, if available, for the industries will have to be purchased at a higher cost'than the power they have already obtained. B.C. Hydro, which sold surplus power last year at rates varying between 20 and
-30 mills per kilowatt-hour, has no such power available this year.
Utilities in the Pacific Southwest, if they have power available for 1 sale, will sell it at at least their cost, which is 23-25 mills per '
kilowatt-hour, plus losses. BPA therefore estimates that the average cost of the kilouatt-hours that industries would have to purchase to replace that not provided by the Hanford Nuclear Plant would be 25 mills per kilowatt-hour. .This price is 11-13 mills per kilowatt-hour more than the cost of power from Hanford, so that the industrial customers will have to pay.an additional. amount of $25-29 million for energy in
-1978-79.
1
'Page 3 - Testimony of Hector J. Durocher
- __j
hi l
i If energy is simply not available or if these industries are unable
- economically to pay the price necessary to acquire additional energy to i
l totally replace that not produced by Hanford, they will have to curtail l
l production approximately 131/2 weeks earlier than would otherwic- '
l
! necessary unless heavy early fall rains occur. Our estimates show i l l that this would result in the loss of approximately 600-800 jobs in l
l direct employment and perhaps up to 2,000 jobs in related services and industries. In addition, the loss of production and shutdown and startup costs would result in further cost and losses in the tens of millions of dollars.
If the Trojan plant is restarted this fall, a major part of the Hanford
)
ergy would be reassigned to BPA's industrial customers.
This would l rkedly reduce the excess costs which BPA would incur and pass on to l l
ts customers. It would also limit the industries' exposure to added ,
costs to the purchase commitments they have already made.
. l 1
i i.
1 Page 4 - Testimony of Hector J. Durocher
_. ._ .. - - - .