ML19241A969

Forwards Bechtel Responses to NRC Re Bolt Assembly Dead Loads & Block & Beam to Column Connection Capacities.Balance of Responses Will Be Submitted 790710. Certificate of Svc Encl

ML19241A969
Person / Time
Site:	Trojan File:Portland General Electric icon.png
Issue date:	07/06/1979
From:	Rachel Johnson PORTLAND GENERAL ELECTRIC CO.
To:	Schwencer A Office of Nuclear Reactor Regulation
References
TAC-07551, TAC-11299, TAC-7551, NUDOCS 7907110479
Download: ML19241A969 (55)

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1 58 _JTi July 6, 1979 Trojan Nuclear 'lant Docket 50-344 License NPF-1 Director of Nuclear Reactor Regulation ATTN:

Mr. A.

Schwencer, Chief Operating Reactors Branch di Division of Operating Reactors U.S. Nuclear Regulatory Conmission

'4a s hi ng t o n,

D. C. 20555

Dear Sir:

Enclosed are additional responses prepared by Becntel Power Corpora-tion covering most of the outstanding quest'.ons from your letter of May 18, 1979. As discussed with your staff, answers to the balance of these questions will be submitted Tuesday, July 10.

The scheduled revision of PCE-1020 will follow the responses to the re=aining ques-tions by one weck as committed in the June 15 ceeting with your staff, Bechtel Power Corporation and PCE in San Francisco.

Sincerely, h

/

I R. W. Johnson Corporate At torney Portland General Electric Company R'JJ/JCL/4sb6 A2 5 Enclosure

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A, c>ill 790711 6'/

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UNITED STATES OF AMERICA NUCLEAP REGULATORY COMMISSION BEFORE THE ATOMIC SAFETY AND LICENSING BOARD In the Matter of

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)

Do cke t 50-344 PORTLAND GENERAL ELECTRIC COMPANY,

)

et e.1

)

(Control Building Proceeding)

)

(Trojan Nuclear Plant)

)

CERTIFICATE OF SERVICE I hereby certify that on July 6, 1979, Licensee's letter to the Director of Nuclear Reactor Regulation dated July 6, 1979 and an attachment entitled " Request for Additional Information, Trojan Nuclear Plant, Preposed Control Building Design", have been served upon the persons listed below by depositing copies thereof in the United States = ail with proper postage af fixed for first class mail.

Ma rshall E. Mille r, Esq., Chairman Joseph R. Gray, Esq.

Atomic Safety and Licensing Board Counsel for NRC Staf f U.

S.

Nuclear Regulatory Co==ission U. S. Nuclear Regulatory Coccission Washington, D.

C.

20555 Washington, D.

C.

20555 Dr. Kenneth A. McColloc, Dean Lowe ns tein, Newman, Reis, Axelrad &

Division of Engineering, Toll Architecture and Technology 1025 Connecticut Avenue, N.

b.

Oklahoma State University Suite 1214 St illwater, Oklaho=a 74074 Washington, D. C.

20036 Dr. Hugh C. Paxton Richard M.

Sandvik, Esq.

1229 - 41st Street Assistant Attorney General Los Alamos, New Mexico 87544 State of Oregon Department of Justice Atomic Safety and Licensing Board 500 Pacific Building Panel 520 S. W. Yamhill U.

S. Nuclear Regulatory Coenission Portland, Oregon 97204 Washington, D. C.

20555 William Kinsey, Esq.

Atomic Safety and Licensing Appeal Bonneville Powe r Administration Panel P. O.

Box 3621 U. S. Nuclear Regulatory Commission Portland, Oregon 97208 Washington, D. C.

20555 Docketing and Service Section Office of the Secretary U.

S.

Nuclear Regulatory Commisei n Washington, D.

C.

20555 as22:

n. ). 7 e

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CERTIFICATE OF SERVICE Ms. Nina Bell Mr. Eugene Rosolle 728 S. E. 26th Avenue Coalition for Safe Power Portland, Oregon 97214 215 S. E. 9th Avenue Portland, Oregon 97214 Mr. John A. Kullberg Route 1, Box 2 50Q Columbia County Courthouse Sauvie Island, Oregon 97231 Law Library Circuit Court Room Mr. David B. McCoy St. Helens, Oregon 97051 348 Hussey Lane Grants Pass, Oregon 97526 Dr. Harold I. Laursen 1520 N.

W.

13th Ms. C. Call Parson Corvallis, Oregon 97130 P. O.

Box 2992 Kodiak, Alaska 99615 ghf

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RonaldW.finnson Corporate Attorney Portland General Electric Company Dated: July 6, 1979 c

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6.

(a)

For the " Criteria for Bolts", provide the following:

A clear description of the bolt assembly and hardware arrangement.

Answer:

The bolt assembly consists of a 1-3/4" diameter ASTM A490 rod threaded on both ends, a 2-3/8 in. thick, 18-1/4 in. by 18-1/4 in. ASTM A36 bearing plate on the inside of the Control Build-ing, the 3" thick ASTM A36 steel plate on the outside of the Control Building, and a heavy hexagon nut and hardened washer on each end of the rod.

One inch of grout will be placed be-tween tne 3" steel plate and the existing wall, and between the bearing plate and any concrete or block surface.

The bolt assembly and hardware arrangement are shown in the attached figure.

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N$55 UU'

t I" GROUT (GRCUr IS OMI1*/'ED WHEN NEW CONCRErG /6 PLACED AGA/NST 5'th) 3 "A STM, A 3 GPLAT61,

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BOL T' ASSEMBLY V78 SEE RESPONSE No G t'a.)

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6.

(b)

Page 1 of 3 pages For the " Criteria for Bolts", provide the following:

1 The bKsis for the formula to eclculate the allowable shear force for the bolt including the contact area between the wall and the stool, the stress distribution at the wall / steel inter-face and the maximum compressive stress induced in the wall at this interface,long with justification for the value.

Answer:

The bolted conection of the steel plate to the concrete wall provides a means of transferring forces from one struc-tural element to the other.

The transfer of forces is based on the principle that a clamping force between two potential sliding surfaces provides resistance to sliding.

The magni-tude of the sliding resistance is equal to the clamping force times the coefficient of friction and is expressed by the following formula:

Y,$N ? j YY $$ $$

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Q. 6.

(b)

Page 2 of 3 pages V - pF where7 V = resistance to sliding coefficient of friction

=

a F = clamping force Tha clamping force, F, is supplied by the tension in the bolts.

The loss factor, (L), is added to take into consideration the losses in the bolt tension.

Adding a factor of safety (P.S.)

the resistance of one bolt becomes, V=gLF F.S.

It is seen from the above relation that the contact area between the wall and the steel is irrelevant to the magnitude of the shear resistance.

The size of the bearing plates was determined by dividing the initial bolt tensile load of 200 kips by the allowable masonry bearing stress of 600 psi which was calculated f rom Table No.

24-H of UBC 1976.

Since the allowable bearing stress for masonry is lower than that for concrete, the naso'ry controls in the design of the bearing plates.

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6.

(b)

Page 3 of 3 pages Using formulas for plates on elastic foundation, the press-ures qnder the 3" plate and the 2-3/8" thick bearing plate have been calculated.

For the 3" plate, formulas for a point load on an infinite plate were used.

The maximum compressive stress under the 3" plate is 1120 psi.

The stress reduces to 600 psi at 6-1/2" and to 150 psi at 12" away from the center of the bolt.

The pressure under the 2-3/8 " plate was calcu-lated assuming a point load on a plate with a diameter of 18-1/4".

The maximum compressive stress is 1760 psi.

These local stresses are justified because the average streso under the bearing plate will be equal to the code allowable bearing stress and at no point under the plates will the stress exceed the compressive strength of the block, r

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6.

(c)

Page 1 of 4 pages For the "Cciteria for Bolts", provide the following:

The bdsis for tte assumed loss factor.

Answer:

The loss factor, L,

is the percentage of tension remaining in the bolts after all losses (shrinkage, creep, bolt relaxation and temperature effects) have been subtracted from the initial tension force.

The losses considered and values used to de-velop a conservative loss factor for the bolts are as f ollows:

1.

Shrinkage:

The following upper limit values have been used in deter-mining the loss due to shrinkage:

New concrete, 355 x 10-6

-6 Existing block, 200 x 10 Shrinkage in the existing block occurs where new concrete is placed against it.

This is due to cwelling caused by water migrating from the new concrete.

2.

Creep Creep in both the new concrete and existing block is taken as 1.6 times the elastic deformation.

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(c)

Page 2 of 4 pages Creep in the 1"

thickness of grout under the base plates afjd 3 inch plate is taken as 2.0 times the elastic d4 formation.

The creep losses are conservatively calculated using the maximum compressive stresses given in response to Question No. 6 (b).

3.

Bolt relaxation:

The long term bolt relaxation is taken as 2.5% of the initial bolt tension.

Initial bolt relaxation will not contribute to the bolt losses because of the two-pass tensioning described below.

r The bolts will be tensioned with a stod tensioner in a two pass progra n.

The bolts will be tensioned to the required tension in the first pass.

A second pass will be made 24 hours2.777778e-4 days <br />0.00667 hours <br />3.968254e-5 weeks <br />9.132e-6 months <br /> or more after the first pass to compensate for the 'osses due to initial bolt relaxation, elastic de f o rn._ t ivn, and the influence of adjacent bolts.

The tension in the bolts will be monitored and checked.

4.

Losses due__to temperature:

Losses in the bolt tension due to a temperature rise of 50' have been included.

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(c)

Page 3 of 4 pages The values for bolt losses due to creep and shrinkage have been determined by considering the properties of the low shriniage and creep concrete specifically designed for the Corplex modifications.

The value for long term bolt relax-ation has been obtained frrm " Guide for Design Criteria for Bolted and Riveted Joints", by J.

W.

Fisher and J. Struik, 1974.

The design value losaes were determined for the three differ-ent attachment conditions: 1) bolts attaching the plate to new concrete; 2) bolts attaching the plate to new concrete and existing olock; 3) bolts attaching plate to existing block.

The maximun decrease in tension for the three conditions is less than 25% of the initial tension.

A tabulation of the i

losses for the three conditions is given in the attached table.

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shrinkage i

i l

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Tbtal l

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5.8%

l l

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2.5%

i 1.7%

i 21.3%

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cmcrete I

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l 1

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l 1.4%

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See response M>. 6(c) cx,'

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0. 8.

Verify that all resistances and stifiaesses based upon dead loadiconsiderations considers the dead load to be reduced by the vertical earthquake component.

Answer:

The shear capacity of a wall panel, as given by the flexural equation described in Section 3.4.2.2 of PGE-1020, is in-creased by a compressive axial load.

The axial load, derived from the dead load, hac been decreased to account for the ver-cical component of the earthquake.

As indicated by the test results, the stif fness of the walla depends on the axial load.

During the earthquake, the axial load on the walls oscilates about the mean value which results from the static dead loadi In determining the mean frecuency, the mean value of the axial i.c a d is derived frem the direct dead load and the influence of the deviation from the mean value is discussed in response to question 47.

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12.

Page 1 of 4 pages Justify the ductility limit of 4 for the outer rebar in the flextral calculations.

Also, considering displacement com-patidility for the entire structure using the stiffnesses indicated by the test results, what are the strains predicted in the outer rebar.

Justify their acceptability in light of your assumptions.

Additionally, for the flexural analysis equations Justify the use of a comp ~ession zone length of 10%

r of the total effective length, and cupply the maximum values of E and Justify the use of a linear stress-strain relation-c ship for the concrete in compression.

Answer:

Both Sec tion 10. 2 of ACI 318-71 and Section 2610 of UBC-76 i

allow fe,r certain assumptione in determination of the ulti-mate strengths of members subjected to flexure and axial loads, when the members satisfy tne applicable conditions of equilibritm and compatibility of strains.

The referenced sections of tie above codes provide that, for strains in reinforcement greater than those corresponding to the speci-fied yield strength, f, the stress in the reinforcement y

shall be considered independent of strain and equal to f.

y The codes, therefore, impose a limitation on the stress in the reinforcement and do not have any restriction on the strain level.

The maximum usable strain in the concrete at the extreme compression fiber, however, shall be limited to 0.003.

Therefore a ductility ratio, used for purpaces of illustration iri PGE-10 2 0, is consistent with code provisions provided the maximum concrete strain is less than 0.003.

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c F, Q. 12.

Page 3 of 4 pages the displacoment compatibility of the entire structure, dispgocements associated with factored OBE loads will in-creasi by a factor of approximately 4.3 over those associated with the unfactored OBE loads.

Based on the above, the computed ultimate interstorey displacement can be esti-mated to be between 0.03 in. and 0.04 in. for a typical panel of comparable dimensions as above.

Also, the response to Question No. 43 indicates that the double curvature assump-tion provides for conservative evaluation of capacity and a more flexible structure is expected.

A more realistic assess-ment of the behavior of the walls will provide for a displace-ment closer to the ones above.

The relationship between the concrete stress distribution and I

tne concrete strain, according to the code provisions, may be assumed to be a rectangle, trapezoid, parabola, or other shape which results in prediction of strength in substantial agree-ment with the results of comprehensive tests.

Many research-ers in the past have obtained good correlation with experi-rae n t al results by assuming a triangular stress distribution.

Reference may be made to the paper, " Strength of High-Rise Shear Walls - Rectangular CrossSection" by Cardenas and Magura in publication No. SP-36 of ACI.

The theoretical flexural equation described in Section 3.4.2.2 of PGE-1020 has been analyzed by computer for a 24 inch thick wall by assuming a value for the modulus of elasticity of concrete, Ec, equal to 6

3.92 x 10 psi, as given by the procedure described in Appen-dix B of PGE-1020.

In calculating E the compressive c,

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o. 12.

Page 4 of 4 pageG strength of concrete is taken as 5000 psi.

The results of this, analysis showed that both for the reinforced and unrein-forced cores and for a range of vertical etress as exists in Complex walls, the calculated values of shcar stress capacity could be obtained within a bound of +5% if the assumption was made that the compressive zone was 10% of the wall's effective length.

Hence the asumption of 10% is made to simplify the flexural equation and facilitate hand calculation without any loss of accuracy.

The analysis also showed that the maximum concrete strain was 0.0016.

For strains in this range, stress-strain relationships may be considered linear.

47F A

Q. 16.

Page 1 of 8 pages Provide the basis for your calculation for the block and the geam to column connection capacities.

Include a dis-cussfbn of the strain compatibility of the two, and the basis for the 100 psi allowable vertical shear on the block at corners which seems to include a 1/3 increase in UBC allowable stresses which would not be appropriate nor in line with current practice.

Answer:

Section 3.5 of pGE-1020 describes the procedure followed to examine the mechanism of vertical shear transfer from side-walls to end walls of the Compler.

An alternative approach is discussed below where the capacity of the corners of the i

walls to transfer the vertical shear is shcwn to be a combi-nation of the contribution from the beam-column connection and the shear friction developed by the continuous horizon-tal reinforcing steel in the concrete block masonry.

In order to determine the ability of the beam-column connec-tion to transfer the vertical shear, it is necessary to con-sider both the ultimate strength and the load-deformation characteristics.

These aspects will be examined in light of the failure plane envisaged and the type of connection.

The beam-column connections of the structural framing system embedded in the Complex walls have been designed as simple bearing type connections according to the working stress ve,--

03

Q.

16.

Page 2 of 8 pages

<~

k method of AISC.

They consist of connection angles bolted to e

the Rolumn and the web of the beam.

In order for a f ailure to develop at the boundary between a wall panel and an em-bedded steel column, a crack plane would have to form through the joint, thus generating a potential for a verti-cal slip between the connection angle and the column.

If it is conservatively assumed that no interaction take

]

place between the wall panel and the embedded column, this T

potential vertical slip will be resisted by two mechanisms:

1 1)

The resistance of the beam-column connection 2)

The shear friction and dowel action developed by the horizontal reinforcing steel of the concrete block masonry crossing the crack plane.

In order for the effects of the two resistance mechanisms to be additive, both the beam-column connection and the shear-friction mechanisms must carry their ultimate loads at a comparable value of slip.

1.

Consistency of deformation a.

Beam-column connection The experimental load-slip data for the cteel con-nections are obtained from References 1 and 2.

The load-slip curve for a six-rivet joint loaded e?"'"

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O. 16.

Page 3 of 8 pages in tension presented by Davis and Woodruf f in Ref. 1 indicates that che ultimate load for the connection is attained in the range of 0.02 inch to 0.03 inch.

The load-slip characteristic by Bendigo et al.,

in Ref. 2 has shown the following:

"It should be noted that riveted joints often experience slip despite the customary assump-tion that hot driven rivets fill the holes completely.

In these tests, slip amounted to approximately 0.02 inch, which is about one-third to one-quarter of the slip experienced by the bolted Joints".

f The conclusion is, therefore, that the bolted connection will attain a slip at ultimate load equal to 0.06 inch to 0.08 inch.

Applying the "one-third to one-quarter" relationship to the Davis and Woodruff results would, however, indi-cate an ultimate slip of about 0.06 inch to 0.12 inch.

A reasonable value of slip at ultimate load for high strength bolt in a simple connection is considered to be 0.08 inch.

b.

Shear-friction of reinforcing steel Survey of the existing literature has not provi-ded any experimental data on load-slip character-istic of masonry block with in-fill concrete.

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Q. 16.

Page 4 of 8 pages However, results of tests reported i n Re f. 3 can be

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used to predict the behavior.

The concrete strengths of the test specimens used to develop the load-slip curves in Ref. 3 varied from 3930 psi to 4090 psi.

These strengths are compar-able to the strength of the concrete filled masonry blocks of the Complex walls.

The paper shows that the maximum shear resistance for all specimens with prepared joints reached their peak at a slip between about 0.02 inch and 0.03 inch.

However, at a slip between 0.08 inch and 0.10 inch, the resistance was still maintained without any appreciable loss in the capacity.

The extent of the slip is further sup-k ported by the shear wall testing program of Appendix A,

PCE-1020.

Specimen D, which had no core concrete, exhibited a slip of about 0.10 inch and the specimen maintained the shear resisting mechanism.

It can, therefore, be concluded that the shear resistance mechanisms, as provided by the beam-column connections of the fully embedded structural framing system and the shear-friction of the continuous horizontal reinforcing steel in the mono-lithic concrete block masonry units, will develop their ultimate resistance at about a slip of 0.08 inch.

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Q. 16.

Page 5 of 8 pages 2.

Ultitate Resistance a.

Beam-column connection In order for the vertical shear crack plane to form through the connection, the u.ti:ste strength of the connection will be governed by the shear strength of the A-325X bolts and by the bearing on A-36 material under te bolts.

Concrete encasement of the fully embedue. Joints will preclude any prying action in-duced by bending moment.

This suggests that the evidence derived from experiments on lap joints and butt splices should apply to this type of connection also, with the exception that the tansile failure on I

the net section experienced by butt and lap splices would not be expected, due to the arrangement of the connected pieces.

Wallaert and Fisher (Ref. 4) gave a conservative estimate of ultimate shear strength of A325 bolts as 75 ksi.

Since AISC specifications permit a design shear stress of 22 ksi, this indicates a 75/22 = 3.41 factor of safety with respect to ultimate.

The facter of safety for bearing may be obtained as follows:

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Page 6 of 8 pages Jones (Ref. 5) has shown that the tensile strength of lap and butt joints would not be impaired if the

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nominal bearing stress was less than 2.25 times the nominal tensile stress on the net section.

This ratio has been adopted by AISC, where the allowable bearing stress is restricted to 1.35f.

It implies, y

therefore, that the factor of safety for bearing is at least equal to the factor of safety for tension on the net section of the connected materials.

The tests performed on the structural cteel material used in the Complex indicated an ultimate tensile strength of approximately 62 kai.

Taking the A TSC allowable tensile stress of 0.60f 21.6 ksi on the net

=

y section will, therefort, provide a factor of safety i

for tension of 62/21.6 = 2.87, A factor of 2.8 will, therefore, be taken as the factor of safety against bearing type connection designed by AISC working stress methods.

This value is conservative since the type of tension failure which the bearing stress criteria is intended to prevent would not occur in a beam-column goint in the Complex walls.

b.

Shear friction of reinforcing steel Section 11. 5 of the ACI Building Code, ACI 318-77, allows design for shear transfer to be based on the

" shear friction" hypothesis proposed by Birkeland (Ref. 6) and Mast (Ref. 7).

In this approach, it

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Q. 16.

Page 7 of 8 pages is assumed that for some unspecified reason a crack exists in the shear plane.

The shear resistance is

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then assumed to be developed entirely by the fric-tional resistance to sliding of one crack face over the other, when acted on by a normal force equal to the yield strength of the reinforcement crossing the shear plane.

A value of the apparent coefficient of friction, is used to qualify this behavior.

y, For a crack in monolithic construction, is taken u

as 1.4.

This value, as suggested 4/ ACI 318-77, provides a conservative estimate of the ahear trans-fer strength of concrete cracked along the shear plane (See Ref. 8).

Pauley et al. (Ref. 3) obtained values of the apparent coefficient of friction for f

specially prepared rough surfaces equal to 3.4, 2.0 and 1.6 for reinforcement ratios of 0.31, 0.69 and 1.23 respectively.

As can be seen, for this well prepared joint, the value of y increases as rein-forcement ratios decrease.

Results of tests in Ref.

3 also indicate that, especially for specimens with smaller reinforcement percentage, the capacities as given by shear-f riction were maintained during a large number of alternating load cycles.

In the Complex walls, the continuous horizontal reinforcing steel in the concrete block masonry constitutes a percentage of 0.12, and therefore, based on the above data, an apparent coefficient of friction, p,

can be taken as 1.4.

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Q. 16.

Page 8 of 8 pages The ultimate shear resistance will, therefore, be taken as vu "[2.8 x (Beam connection value of Table I,AISC)

+ ( 1. 4 x As*f x h) y

where, A

= Area of continuous horizontal reinforcing s

2 steel in masonry (in /ft) f

= Yield strength of reinforcing steel = 40 kai y

h

= Height of wall (ft.)

The vertical shear resistance corresponding to the unfac-tared CBE condition is obtained by dividing the ultimate capacity by the load factor of 1.4 after reducing the con-f tribution of shear-friction of reinforcing steel by the capacity reduction factor of 0.85.

Based on the above, the values given in Section 3.5 of PGE-1020, will be revised as folicws:

Corner Verticcl Shear Force (kips)

Capacitv (kips)

R-55 2357 2742 N-55 1260 1763 BID @? I'\\ ?"..i

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References for Response No. 16

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1.

Da v i s,

R.C.,

and Woodruff, G.

B.,

" Tension Tests of Large Riveted Joints", Transactions of the ASCE, 1940 2.

Dendigo, R.A.,

Hansen, R.M.

and Rumpf, J.L.,

"Long Bolted Joints",

Journal of the Structura) Division, ASCE, De cembe r, 1963 3.

Pauley, T.,

t'a r k,

R.,

and Phillips, M.H.,

" Horizontal Construction Joints in Cast-in-Place Reinf orced Concrete", ACI Publication No. SP42-27

.4.

Wallaert, J.J.

and Fisher, J.W.,

" Shear Strength of High-Strength Bolts", Journal of the Structural Division, ASCE, June, 1965 5.

Jones, J., " Bearing Ratio Effect on Strength of Riveted Joints",

Transactions of the ASCE, Novembe r, 1956 6.

Birkeland, P.W.

and Birkeland, H.W.

" Connections in Precast Concrete Construction", Journal of the American Concrete Institute, March, 1966 7.

Mast, R.P., " Auxiliary Reinf orcement in Concrete Connections",

Proceedings, ASCE, June, 1968 8.

Mattock, A.H.

and Hawkins, N.M., " Shear Transfer in Reinforced Concrete-Recent Research". PCI Journal, March /A p ril, 1972 p;,',,,,;c. p n

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Page 1 of 3 pages Discuss in detail the effects on the in-plane wall shear cap (city of any tension induced in the walls by the gross over' turning moments and the " plate bending" of the walls generated by the earthquake component perpendicular to these walls.

Answer:

In general, the tension forces developed from seismic res-ponses on one side of a structure due to gross bending are offset or balanced by an equal amount of compression on the other side of the structure.

On a smaller scale, this type of behavior developed i-the test specimens and is reflected in the observed capacities.

The criteria for capacity as given in PGE-1020 is ::inear with respect to axial load, therefore, no major influence on the capacity is expected frem tension induced by gross bending.

This agrees with current shear wall design practice since there is no special consideration for this effect.

The plate bending effect in the shear walls which are carry-ing 'he primary shear loads is due to the component of earth-quake perpandicular to the component causing the primary shear.

The Tro]an FSAR does not require the effects of the two hori-zontal components to be combined; but, if it were considered, (f)

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O. 17.

Page 2 of 3 pages the effect would be small.

Taking into acceent that the paa( response due to ts, horizontal components of the earth-quake do not occur simultaneously, the effects can be combined I

using the recommendation of Newmark wh i ch is" -- take the combined ef fects as 100 percent of the effect in one particu-lar direction and 40 percent of the ef f ects corresponding to the two directions of motion at right angles to the principal motion considervd".

Under these conditions, the plate bending effect is small.

If 40 percent of the transverse inertia loads were considered as a static load, the load could be resisted by the vertical reinforcing steel in the block of the composite walls only without yielding.

Considering the ultimate case of the effect of the full longitudinal shear force combined wi th the dead load, some vertical reinforcing steel may yield.

With the imposition of 40 percent of the transverse inertia loads and considering the transient nature of the load, the load could cause slight additional yielding as the energy associated with the transverse inertia loads is absorbed.

The amount of additional yielding can be bounded by considering the deformation during a static application of the loads.

As indicated above, the load can be resisted with the vertical blCCk reinforcing steel only ard the steel does not yield.

The strain energy in the steel for this loading condi tion is less than one half the yield strain times the yield stress (0.5 e fy y) since the strain energy is the area t/ ). E,. "3

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1

Q. 17.

Page 3 of 3 pages under the stress-strain diagram and the stress increases linearly wi th strain up to the yield strain.

The additional strain that will develop is that required to develop enough strain energy to balance that fr'om the transverse inertia loads.

If the reinforcing steel in the zone at the wall under consideration is at yield or beyond, the strain in the reinforcing steel increases without an increase in stress.

Since the area under the stress-strain diagram is the strain energy and the energy to be absorbed is less than 0.5 e f yy, the additional strain would be only one half the yield strain.

This amount of additional strain will not reduce the in-plane shear capacities.

Reference 1.

Newmark, N.

M.,

W.

J. Hall, Comments on Inelastic Seismic Capacity of Nuclear Reactor Structures, Civil Engineers and Nuclear Power, Vol 2, ASCE Convention, Boston, April 2-6, 1979.

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O. 29.

Page 1 of 11 pages De scribe in detail the modifications necessary to ensure the seismic qualification of the complex as a result of the strebsthening or stiffening of the structure and sequence in which they will be performed.

Answer:

The review of safety-related equipment, c om po ne n t s, and pip-ing within the Complex has been made in accordance with Ap-pendix B of PGE -10 2 0.

The review of the safety-related equipment and instrumentation has not been completed.

How-ever, results to date indicate that no modifications are re-quired by the revised response spectra.

The results of this review to date have revealed that some modifications to the f

cable tray supports and piping supports are required.

The cable tray supports which require modification to remain seismically qualified to the revised response spectra are indicated on the attached Table 29-1.

The modifications of these cable tray supports will be made prior to the structural modifications to the Complex.

The piping system supports which require modifications are indicated in Table 29-2.

This modification work will be per-formed in a sequence for those piping systems required for safe-shutdown, ECCS, or to mitigate...itigate consequences of accidents which could result in releases exceeding 10CFR100 guideline limits, the change to the support to meet SSE f--

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Page 2 of 11 pages requirements will be implemented prior to the structural modification tha t necessitated it.

These items are ident-ifie'd by asterisk in Table 29-2.

Evaluations have been performed to confirm that all safety related piping, cable trays and equipment attached to non-shear walls in the Complex would retain their support cap-abilities when subjected to a seismic event.

The walls in question are identified in Tabic 29-3.

Evaluations have confirmed that the stress levels in these walls are low and that non-linear behaviour which could reduce this support function will not occur.

The safety related equipment, cabling and piping dependent upon these walls for support are identified in Table 29-4.

Their support location and configurations have been con-firmed by plant survey.

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Page 3 of 11 pages TABLE 29-1 CABLE TRAY SUPPORT $i REQUIRING hCCIFICATION Builcing Elevation Support Number Aux 11iar' 77' Area 3

17,16,21,22,23,24,35 Control 61' Area 6

6C 77' Area 6

119,121,122 93' Area 13 7,13,15 t

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0. 29.

Page 4 of 11 pages TABLE 29-2 PIPING SUPPORTS TO BE ADDED Large Pipe Small Pipe (2 1/2">0.D) 10" Cid-9-2 SR-300

• 3/4" HKD-1-53 SR-317

• 3" CS-151R-9-3 SR-300 1-1/2" HKD-3-58 SR-322

• 4" CS-IS1R-12-2 SR-301 1-1/2" HKD-1-6 0 SR-3 20

• 4" CS-151R-26-1 SR-300 3/4" HKO-1-60 SR-325

• 4" HS D-91 -4 SR-300 2"

HKD-1-69 SR-318

• 4" HBD-91-4 SR-301 2"

HKD-1-69 SR-319

• HCC-62-1 SR-300 3/4" HKD-1-69 SR-321

• HCD-2-7 SA-300 1"

HKD-3-76 SR-323

• 4" RC-151R-1-2 SR-302 1"

HKD-1-76 SR-324

• 4" RC-151R-1-2 SR-303 2"

HKD-2-64 SR-303

• 3" RC-151R-19-1 SR-300 2"

HKD-2-64 SR-304

• 8" HBD-91-2 SR-302 1"

HKD-1-69 SR-320

• 3" HCD-19-1 SR-300 10" GCB-7-1 SS-300

• f 10" GCB-9-1 55-304 12" HBD-28-2 55-301 6" HBD-33-1 SS-302

• 6" HDD-33-2 SS-303

• 4" HCC-23-2 SS-300

• HBD-22-3 SR-301 HBD-22-3 SR-302

• 4" SI-150lR-1-1 SR-301

• 3" SI-15018-1-1 SR-302 10" HCC-49-1 SR-300 6" HFD-3-6 SR-300

• 4" CS-151R-6-1 SS-300 4" CS-151R-6-1 55-301 4" CS-151R-6-1 SS-302 4" CS-151R-6-3 S5-300 4" CS-151R-6-3 SS-301 3" CS-151R-9-2 SS-300 3" CS-151R-9-2 SS-301 3" CS-151R-9-2 SS-302 3" RC-151R-19-1 SS-301 3" HCC-29-3 SS-300

• Installation required prior to modifications strengthening the Complex

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C. 29.

Page 5 of 11 pages TABLE 29-2 PIPINC SUPPORTS TC BE ADDED 10" HCC-48-1 SR-301 1"

JBD-31-61 SA-309 10" 1:CC-49-2 SR-300 1-1/2" JBD-31-67 SA-312 3" HCC-65-2 SR-300 1"

JBD-31-70 SA-311 4" JBD-35-1 SA-301 1"

JBD-31-96 SA-310 4 " JBD-3 6-1 SA-300 3" hCB-3-1 SR-300 10" HCC-48-1 SP-302 8"

hCC-46-1 SR-3C3 10" HCC-48-1 SP-304 HCC-48-4 SR-300 w

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0. 29.

Page 6 of 11 pages ATTACHMENT TABLE 29-2 PIPING SUPPORTS TO BE ADDED Large Pipe Small Pipe (2 1/2">0.D) 3" CS-151R-30-1 H-27 3" CS-151R-30-1 H-28 3" CS-2501R-28-2 SR-78 3" CS-250lR-28-2 SR-81 8" RH-60lR-7-1 SA-19 10" GCB-9-2 SP-31 10" GCB-9-2 SR-41 12" HBD-28-1 SA-213

• 12" HBD-30-1 SA-241

• 14" HBD-31-2 SA-807 f

4" HCC-23-2 SA-129

• 8" CS-151R-5-1 SR-26 14" SI-151R-10-1 SR-76 10" GCB-9-2 SR-38 24" HBD-27-3 SR-300

• 24" HBD-27-4 SR-815 14" HBD-31-2 SR-111

• 6" HBD-33-1 SR-193

• 4" CS-151R-6-1 SR-17 4" CS-151R-6-3 SR-24 3" CS-151R-9-1 SR-97 3" CS-151R-9-3 SR-87 14" SI-151R-10-1 SA-81 3" CS-151R-12-2 SR-157 4" CS-151R-12-3 SR-110 4" CS-151R-12-3 SR-lll 4" CS-151R-12-3 SR-ll2 4" CS-151R-12-3 SR-113 4" CS-151R-12-3 SR-ll7 4" CS-151R-12-5 SA-149 3" CS-151R-16-1 SA-159

- -* 3 c ppo tr ysuG. I. eu. v u. m i. c d pm tc; 7 nodi f ica t ior.s s t t engetr --- --

enino the Complex L} h b

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Q 29.

Page 7 of 11 pages ATTACHMENT TABLE 29-2 PIPING SUPPORTS TO BE MODIFIED Large Pipe Small Pipe (2 1/2">0.D) 4" CS-151R-26-1 SR-41 4" CS-151R-26-1 SR-155 3" CS-250lR-5-4 SR-75 14" SI-601R-5-1 SR-41 4"

SI-2501R-3-3 SR-46 4" SI 2501R-3-3 SR-50 8" GBD-18-4 SA-1 8" GBD-18-4 SA-6 8" HBD-91-3 SR-70 8" HBD-91-3 SR-73 8" HBD-91-3 SA-72 8" HBD-91-3 SA-75 3" HCC-27-1 H-2 3" HCC-27-1 H-4 f

3" HCC-27-1 H-6 3" HCC-39-3 SA-156 3" HCC-62-1 SR-23 4" CS-151R-5-3 SR-11 3" CS-151R-9-3 SR-83 4" CS-151R-12-3 SR-107 4" CS-1518-12-3 SR-ll5 3" CS-151R-12-5 SR-138 3" CS-2501R-5-4 SR-78 8" HBD-91-3 SR-53 8" HBD-91-1 SR-54 8" HBD-91-1 SR-69 3" HCC-12-2 SR-27 3" HCC-62-1 SR-1 12" RH-60lR-7-1 SR-63 4" SI-151R-10-7 SR-2 4" SI-150lR-1-1 SR-65

• 4" SI-1501R-1-1 SR-69

• 10" GCB-9-1 H-21 *

• Support modification required prior to modifications s tre ng the ning the Complex h

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Q. 29.

Pace 8 of 11 pages ATTACHMENT TABLE 29-2 PIPING SUPPORTS TO BE MODIFIED

~;

Large Pipe Small Pipe (2 1/2">0.D) 6" HBD-33-2 SR-105*

14" HBD-34-3 SA-242*

6" HFD-2-2 SR-68

• 12" RH-60lR-7-1 SR-62

• 4" SI-151R-10-7 SR-5 14" HBD-27-1 SR-225*

14" HBD-27-1 SR-227*

30" HFD-1-1 SR-216*

30" HFD-1-1 SR-217*

6' HFD-2-6 SR-75

• 6" HFD-3-6 SR-15

• 10" HCC-48-1 SS-A4 8" GCB-7-1 SR-61 4" HBE-31-1 SR-6

}

10" HCC-48-2 SR-70 10" HCC-49-2 SA-103 3" HKD-2-1 SR-60 8" GCB-7-1 SR-63 10" HCC-48-J SR-37 10" HCC-48-2 SR-82 10" HCC-48-2 SR-86

• Support modification required prior to modifications strengthening the Complex V'

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Q. 29.

Page 9 of 11 paqes TABLE 29-3 LIST OF NON-SHEAR WALLS 7

Floor Building Elevation Wall Location Control Bldg.

65'-77' Walls surrounding the Battery room 61'-77' Col. line 51 E-W 61'-77' Col. line 53 E-W (check) 61'-77' West of N 77'-93' South of 51 E-W and N-S walls 77'-93' Computer room North I

and East walls.

93' -105' Wall on 51 N-S 105'-117' Wall on 51 E-W Wall on "0" N-S Auxiliarv 45'-61' Wall H between 54 Bldg.

and 55 N-S 45'-61' East of line H all the 8'-0" high small walls 61'-77' West of line E between 60 and 61 N-S 77'-93' Line E and West of Eenclosing Valve Compartment; 8'-0" high walls h

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O. 29.

Page 10 of 11 pages TABLE 29-4 SAFETY-RELATED COMPONENTS ON NON-SHEAR WALLS Piping Area /

or Pull Control Elevation Restraint Conduit Box Panel Others Aux 111ary/45' SI-151R-10 BB469X BPB457 None Tray suppote HCC-71-50 BB470X APB442 BIR 121 AB468X BI-4909 BBR 411 BB434X BI-4909 BB435X BB436X BB490X AB462X BB4153 Auxiliary /61' CS-151R-6-3 BB-4107 None None SI-60lR-9-50 BB-4050 HCB-9-1 AB-4992 CS-60lR-4-51 AB-4969 SI-601R-9 Valve leak-off line for for MOV-8809B Auxiliary /77' CS-250lR-28-50 CS-250lR-28-51 CS-151R-9-58 CS-151R-9-4 Control /61' HKD-2-51 AB-1011 BPB-108 C-182 Roon cooler BB-1010 BPB-116 C-181 supply &

CB-1012 BPB-109 C-180 return BI-1033 L-12 (V-1458 & C)

BI-1034 L 28 BI-1035 0-23 BI-1036 D-62 L-05 L-27 C-262 L-29 D-09 2

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0. 29.

Page 11 of 11 pages TABLE 29-4 (Continued)

~_ SAFETY-RELATED COMPONENTS ON NON-SHEAR WALLS Piping Area /

or Pull Control Elevation Restraint Conduit Box Panel Others Bettery room None BP-1013 None None Space Heater BP-1091 Circuit BB-1072 breaker BB-1185 Q-28 BB-1187 Q-22

~

AB-1021 BB-1072 BD-1286 AP-1005 Control /77' None AB-10ll APB-136 C-249 AB-1045 CPB-165 C-243 I

BB-1138 C-179 AI-1044 AI-1050 AI-1051 BI-1051 AI-1057 AI-1044 AB-1045 AB-1069 BB-1691 Control /93' DI-1901 CPB-158 1

DI-1902 DPB-182 Tray CIA-207 BB-1168 Duct work BB-1022 AB-1081 AB-1017 Control /103' HTD-1-53 APBV-09 C-254 BPBV-09 C-255 APB-125 C-178 BTB-104 C-259-1 APBV-17 C-259-2 BPBV-17 C-260-1 tav-007 C-260-2 BBV-007 et us

Q. 31.

Page 1 of 2 prgo1 Summarize the details of your evaluations which determined that placement of the reinforcing steel, the forms and the concrete will not significantly degrade the seismic capa-bility of the Complex.

Include a definition of significant.

Answer:

The evaluation considered the effect on the existing Complex of the forms, reinforcement and the placing of concrete.

The columns that will be exposed during the modification work were investigated and they were found to be capable of resist-ing the loads induced by the fluid concrete in combination with an earthquaxe, ir addition to the loads in the columns due to dead load, live load and. earthquaxe loads.

The caist-ing block walls were invest. gated for the effects of fluid i

concrete in combination wich an earthquake and were found to be adequate.

The 3-inch thick plate will be used as the outside form where new concrete is placed on the R-line wall up to Eleva-tion 76'-3".

Before concrete is placed, the bolts will be installed through the plate and the existing block walls.

They will be tightened, but not tensioned.

The bolts will prevent the steel plate from moving during concrete place-ment and in the event on an earthquake.

The existing block walls are adequate to withstand the loads induced in them by the plate and uncured concrete during an earthquake.

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Q. 31.

Page 2 of 2 pages Another investigation considered the increase in strength of the new walls as a function of time and the associated wall loads and capacities.

The loads on the new concrete walls were obtained by proportioning the total shear on the wall between the new and existing concrete according to their relative stiffnesses.

Tne stiffnesses and shear capacities of the new walls at various time intervals were based on the increase of f'c with time.

Upon comparing the shear capacity of the new walls with their load demand, it was found that the capacity always exceeded the demand.

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(a)

Provide the correlation, wall for wall, between the test specimens and the actual walls, and Justification for the appl 3cability of the test specimen results to the actual wall including a discussion of the similarities of such items as reinforcing steel ratio and continuity, encasement, material strengths, joint preparation (especially where drypack was used), etc.

Answer:

The results from a particular test specimen were not applied to a specific wall.

In the case of wall capacity, the test specimens formed the basis f or trie use of analytical flexure equation but the results from the test specimens were not used directly.

The applicability of the flexure equation is dis-cussed in response to Question No. 43.

In the case of wall stiffness, the results from a group of test specimens, which represent the range of conditions existing in the actual walls, were used to develop the non-dimensional stiffness re-duction factors.

The applicability of this information is discussed in res po nse to Question No. 46.

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0 37.

(b)

Page 1 of 2 pages With regard to the drypack, refer to the article by Kahn and Hanson entitled, "Infilleo Walls for Earthquake Strengthening" in the February 1979 ASCE Journal of the Structural Division.

This article describes a " brittle" failure of a test specimen with a drypack goint.

Discuss the implications of this with respect to the walls in the Tro]an compicx with the drypack Joints and the applicability of the test results from speci-mens without drypack Joints.

Answer:

The article in question was based upon a thesis report of the same ti tle written at the University of Michigan under NSF Grar.t No. GI-39123.

A review of this report indicates that there are s lificant differences between the drypack zones 4

in the Mict..gan test specimens and the drypack zones in the Complex walls.

The following differences were noted:

1.

The dimensions of the drypack are quite different.

In the Michigan test specimen the horizontal drypack region is 3 inches nigh and only 3 inches thick with the ver-tical reinforcing bars running down the centerline.

This arrangement, with little cover over the reinforcing steel, would seem to promote spalling of the drypeck (after the drypack-concrete bond is broken), with the reinforcing steel acting as wedges to cause longitudinal splitting and eventual spalling.

In the Complex the drypack regions are at least 14 inchen thick with two rows of vertical robars.

These two rows are external to tp[D '% f) in a',

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(b)

Page 2 of 2 pages a large portion of the drypack, and would act to confine the drypack, and inhibit spalling.

Spalling would also be in-hibited by the greater concrete cover in the Complex walls (approx.

4" vs 1-1/2").

The delay of spalling would delay the deterioration of the shear friction mechanism.

2.

The Michis - test specimens were deliberately de. signed as heavily reinte;'ced shear walls inside a nrn-ductile frame.

The " brittle" failure described by the authors appears to be due to the interaction of the infill panel and the frame.

Because of the relative stiffnesses, the columns of the Michigan test specimens carry little shear until the shear friction transfer between the drypack and surrounding con-crete has been extensively degraded.

At this point the shear load is largely transferred to the columns, which then fail in shear in a brittle fashion (since they were designed to be non-ductile).

This situation would not be present in the Complex walls.

The major walls in the Complex do not have any drypack.

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Page 1 of 5 pages Describe in detail how the constant bending moment applied to the test specimens via the auxiliary loading system in conjunction with the main loading system compares to that which would exist due to end restraint in the actual Trojan walls, to justify the applicability to the test specimens results directly to the actual walls.

Answer:

In the shear wall testing program the test specimen.a without steel struts or embedded steel columns had free vertical boundaries.

This condition, therefore, was not an exact representation of the actual Complex wall panels where the behavior of a panel is dependent upon its interaction with the adjacent panels.

Since it was not possible to simulate the interaction effect in the testing program, the test specimens were subjected to a loading condition where the auxiliary loading system in conjunction with the main leading system moved the point of contraflexure near the mid height of the specimen.

As will be shown in the following analysis, the shear capacity of the wall panels obtained from this test set up is a conservative assessment of the actual panel capacity.

In developing the shear capacity of individual wall panels by application of the flexural analysis equation, credit was taken only for the fully embedded vertical reinforcing steel, whicn provided moment resistance at top and bottom of the panel.

The vertical faces of the panel were considered to be totally free.

In actual Complex walls a significant amount

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Q. 43.

Page 2 of 5 pages of vertical shear resistance is generated at the panel vertical interfaces.

Moreover, no credit was taken for the bond between the embedded steel columns and the surrounding concrete core, and no interaction has been assumed to have taken place at this interface.

The vertical shear resistance is given solely by the continuous horizontal block reinforcing steel through the mechanism of shear f riction and the beam-column connection activated by the panel when the panel tends to rotate and pushes against the beam flange.

The following analysis evaluates the shear capacities of wall panels by considering the vertical shear resistance at the side boundaries and a conservative assumption of the single curvature cantilever action of the panel.

In evaluating the vertical resistance of horizontal block rebars, the coeffi-cient of friction, will be taken as 1.4 for the contin-u, uous masonry construction.

The ultimate strength of the beam-column connection will be taken as 2.8 times the working stress capacities given in Table I of AISC for ASTM A325-X bolts in bearing.

The factpr of safety of 2.8 and the apparent coefficient of friction of 1.4 are established in response to Question No. 16.

The compatibility of deformation f

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Q. 43.

Page 3 of S pages between masonry and the slip in the beam-column connection is alsc dita.ussed fn the same response.

Figure 443-1 below shows the free body diagram of a typical wall panel.

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3 Figure 43-1 A,= Horizontal continuous block reinforcing steel Vertical continuous reinforcing steel A

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y For other definitions see Section 3.4.2.2 of PGE-1020 Vertical shear resistance V=V1+V2 Shear friction developed by block reinforcing Vy:

steel, 4-95 9 24" o.c.,

typically 2

A,

= 0.62 in / foot height of wall V3 = U. A,. fy

= (1.4 x 0.62 x 40 x h)/12

= 2.9 h kips, where h is in incnes

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Q. 43.

Page 4 of 5 pages 2:

Shear resistance through beam-column connection:

V For illustration purpose a typical floor beam, W24 x 68 will be considered.

The allowable working stress bearing values of the connection f rom AISC table I-87 = 126 kips V2 = 2.8 x 126

= 353 kips The moment resistance given by the continuous vertical rein-forcing steel and vertical load N (See Section 3.4.2.2 of PGE-1020) is:

M=

0.4 65 A f l 2 + 0.467 N1, g yy Assuming single curvature, P.h = (V1 +V2) lw + M or 2

lw l

y w

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(2,900 r. + 353,000) + (0.465 A f

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blw 2,900 353,000 1,

t=

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+ 0.467 o I y y n

b bh h

is the ultimate shear stress ir. psi. Figure 43-2 Where r shows the values of chear stress given by the above formula for values of p equal to.0021 and.0012 respectively for y

various values of a and for a panel which is 31 ft. long, n

16 ft. high and 27 inches wide.

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Q. 43. Page 5 of 5 pages The shear stress as calculated by the formula given in Section 3.4.2.2 of PGE-1020, assuming double curvature, is also plotted for comparison. As can be seen f rom the figure, the shear stresses based on the double curvature assumption are very low for a range of on between 0 and 100 psi when compared to those obtained by assuming single curvature and shear resistance along vertical faces. Since all Complex walls are within that range of vertical stress it can be concluded that application of the double curvature principle provides a conservative assessment of the shear capacity of the Complex walls. p' ' g' c. 2, c rm, .,,,,,r

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Q. 4 5. Considering the attength of the column connections for the actcal walls, demonstrate that they are capable of resist-ing the axial forces indicated by those results for the columns in specimens L1 and L2. Justify any exceedances of the beam / column connection capacity. Answer: The response to Question No. 39 explains that the specimens L1 and L2 were tested to obtain a knowledge of the behavior of the Complex walls where the structural steel columns are continuous through the floors and where bond is assumed to exist between the embedded columns and the surrounding con-crete. The stiffness obtained from these test data was used in the STARDYNE analysis since an upper bound stif fness provided for a more critical condition for the floor response spectra wi thout causing significant changes in the level of shear forces in the Complex walls. The shear resistancos of l the specimens L1 and L2, however, were not used for evalua-tion of the shear capacity of the walls. Furthermore, as explained in response to Question No. 43, even if the bond and interaction between the steel columns and the core concrete is conservatively ignored, the beam-column connec-tion resistance combined with the shea r. f rictier provided by the continuous horizontal reinforcing steel in the masonry concrete will generate capac2 ties for the walls whict. e considerably higher than those considered in Section 3. 2.2 of PGE-1020 in the range of dead load that exists in the Compl ex walls. i ~ ~ _d ui O k-77* {J'

Q. 47. Page 1 of 3 pages Provide the detailed bases for each of the variations assumed in Poble B-2 in the calculations of the peak broadening pe re'en t age. ~ Answer: As indicated in PCE-1020, the variation in frequency is due to variation in the mass and stiffness. During the develop-ment of the response spectra for interim operation, the var-iation in mass was estimated to be 15%. Since the weight of the new structural elements being added is small and known to at least this accurracy, the variation in total weight will still be taken at 15%. The variation in the initial modulus results from variation in the properties of the re-inforced concrete, concrete bicek, steel plate, etc.

Again, the variation in the properties of the new materials are known as well as those of the existing structure.

Therefore the variation used here is the same as for interim operation. The variation in the stiffness reduction factors are due to vsriations in the shear stress, dead load and experimental uncertaintics. The significance of variations in these para-meters can be put into prospective by estimating the frequ-ency if no stiffness reduction was used. By using various intermediate results in the Ftiffness reduction iteration process and approximate calculations, the frequency of the uncracked structure for the fundamental N-S mode is ectimated to be 8.1 cps. When the stiffness reduction factors are used, the frequency of the same mode as determined in the y 4e I D a nye!w%n["[: e ey g gpi;vg ms -"'Wb3Q U' g s s b jelb

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Q. 47. Page 2 of 3 pages STARDYNE analysis is 7.6 cps, indicating a 6 percent reduc-tion. Considering the accuracy of the calculations, the expected reduction should be in the 5 to 7 percent range. This provides a useful upper limit on the variation in frequency due to the various parameters. The variation in the shear stress is also based on the in-termediate results from the stiffness reduction iteration process. From the wall which shows the maximum variation in shear force as the stiffness reduced factors converged, the sheer force was 3150 kips for the uncracked stiffness. It reduced to 2350 kips on the first iteration and then be-came 2960 kips and 3000 kips on successive iterations. This provides confidence that the shear force is known to 1200 kips or 7%. Other walls showed less variation. This indi-p cates the +10% variation used in PGE-1020 is conservative. The variation in the dead load results from two effects. First is the variation of the actual weight of the struc-ture. The 5 percent variation used for the mass is approp-riate for this consideration. The second source of variation is the load path. Since the majority of the dead load in the walls is due to self-weight and not the floor system, the load path is very simple, straight down the wall, and can be calculated to a variation of less than 5 percent. Tne load path for the other weights - equipment and non-structural walls - is known well enough so that 120 percent used in PGE-1020 is conservative. me a[jS')O.3.,id? f;' ;,:, j[l ((eh p 9 'S I' ,i gl,: J. M q. " L ;7 lr. ([j ? c. m x m., m, [, y } 6.4 m:,' c U -w Mv u Lt % e ;,; rs y Q,j j y 'Q; e M li9 N

Q. 47. Page 3 of 3 pages As indicated in Table B-2 of PGE-1020, the variation in the total structural stiffness is 115 percent resulting in frequency shift at +7.2 percent. Thss frequency shift is greater than the estimated frequency reduction, 5 to 7 per-cent, due to the inclusion of the stiffness redbetion factors. This 5 to 7 percent frequency shift corresponds to an average reduction in stiffness of 12 percent, (0.94)2, 0.B8. Since most structural elements have a stiffness reduc-tion factor of 0.85 or higher, the 115 percent variation in stiffness reduction factor will accommodate a 100 percent increase in the amount of stiffaess reduction. A smaller variation is indicated by the shear stress-deflection curves shown in Figures A2-2 and A2-3 of PGE-1020. Some of the variations among these specimens are due to differing eteel j ration. If this variation was eliminated as is done for the stiffness reduction factor relationship shown in Figure B9, B10 and Bil, the variation would be smaller. These two groups of specimens are the only ones with similar enough properties to provide a meaningful comparison. This consis-tency of results and the small amount at overall stiffness reduction indichtes the 115 percent variation in the stiff-ness reduction factor due to experimental uncertainties is adequate. o((@ n.. vLP Q % ,o n ,,. p w$ y n ..e. ij l f I; ' '; 7 F".,. ]' :'; " u <p.'g fr g OA Y I 4-ti i/0 NY}}