ML20117K895

IPEEE Plymouth,Ma

ML20117K895
Person / Time
Site:	Pilgrim
Issue date:	07/09/1992
From:	Marciano E GEI CONSULTANTS, INC. (FORMERLY GEOTECHNICAL ENGINEER
To:
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ML20117K845	List: ML20117K844 ML20117K851 ML20117K860 ML20117K895 ML20117K905
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Attachment C 9

l GEI Consultants, Inc. Report Pilgrim 1 IPEEE l Plymouth, Massachusetts 1 l

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9609120220 960905 PDR ADOCK 05000293 P PDR

4. gets-w s-a mures Anhas l 2

Y gel Consu tants, Inc.

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1 PILGRIM 1 IPEEE I PLYMOUTH, MASSACHUSETFS l

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l Submitted to Stevenson & Associates Ten State Street Woburn, MA 01801 i

1021 Main Street Project 92012 Winchester, MA 01890-1943 July 9,1992 617-721-4000

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PILGRIM IPEEE PLYMOUTH, MASSACHUSETTS l

l , July 9,1992 . :p .;7-- l 1

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Submitted'to Stevenson & Associates i Ten State Street I l

! Woburn, Massachusetts 01801 '

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! GEI Consultants, Inc.

1021 Main Street Winchester, MA 01890-1943 .

l Project 92012 l

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) //) $0bY '

l Eug'ene A. Marciano, Ph.D.

l Project Manager l J

, EXECUTIVE

SUMMARY

The purpose of this report is to provide estimates of the liquefaction potential, and the seismically induced settlements, permanent horizontal displacements, and transient horizontal displacements of the ground versus peak ground surface acceleration due to an earthquake, for the Pilgrim Nuclear Station in Plymouth, Massachusetts.

The stratigraphy of the Pilgrim site consists of 30 to 50 feet of compacted fill materials above approximately 30 to 50 feet of glacial outwash deposits, which are underlain by bedrock at a depth of approximately 80 feet. The fill consists of sand and gravelly sands with less than 6% fines. He outwash deposits are granular, consisting pr*Aami=tely of poor- to well-graded sauds with some zones of gravelly-sandsimThe fill is heavily compacted. The outwash depositiare very'deise'is a rWult ofloadihidue to giacisition.

For the outwash deposits, previous cross-hole testing by.Weston Geophysical gave shear wave velocities ranging from about 1,700 to 2,700 feet per second (fps) (shil profile 2).

l In addition, shear wave velocities wem calculated for the outwash' soils using empirical correlations available in the literature and blowcount and laboratory test data from the

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Pilgrim site. The results range from 800 to 1,400 fps (soil profile 1). For the outwash, the shear wave velocity profile that produces the more severe loading should be used.

! It is reasonable to expect that the two shear wave velocity profiles bound the true shear l wave velocity profile at the Pilgrim site.

i For the compacted fill, cross-hole test results were not available. ne shear wave l velocities for the fill were calculated using empirical correlations and the results of laboratory tests and range from, about 300 fps at the ground surface to over 1,000 fps at a depth of about 40 feet.

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Liauefaction The soils at the Pilgrim site are highly dilative due to their dense state, and thus their l undrained steady-state shear strengths are greater than their drained strengths. Therefore,

a liquefaction stability failure is not possible, regardless of the magnitude and peak l ground acceleration of the earthquake. A conservative stability analysis conducted using l the drained shear strengths gives a factor of safety of 1.9 against failure.

Settlements For soil profiles 1 and 2, the maximum calculated settlements at the ground surface are

, 0.29 inch and 0.72 inch, respectively. These settlements correspond to peak ground l

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accelerations of 0.35 g and 0.7 g, which are the largest peak accelerations that would develop at the ground surface for soil profiles 1 and 2, respectively.

Differential settlements can be expected within the foundation imprint of any one building and within the areas between buildings due to natural variability of the compressibility of the soil deposits. These can be taken equal to 50% of the total settlements and can be taken to occur over a distance of about 25 feet for structures on individual spread footings and for the areas between buildings. For structures founded on a continuous mat foundation, the differential displacement can be taken to occur over a distance of about 50 feet.

Differential settlements can also be expected botvsa any one building and,the ground due tof the and between adjacent

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of tifsoilbuildings,-such-as- those-within-the Power B

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different" thick 6 esses strata bendith'tsiifous structures an'd ground surface. Those between a building and the surrounding ground will occur over a distance of only a few feet. The distance over_which the differential settlements between adjacent buildings will occur is dependent on the interaction of the foundation mat with the foundation soil' sui can occur abruptly at construction or expansion joints between or within the buildings.

For a given value of 5cceleration, the settlements for the NUREG/CR-5250 and EPRI  !

hazard results are identical.

Permanent Horizontal Disolacements For soil profile 2, the maximum calculated permanent horizontal displacement occurs for the Intake Structure for the NUREG/CR-5250 hazard results. It equals 1.7 inch and occurs for a peak ground acceleration of 0.7 g. The permanent horizontal displacements for soil profile, I are slightly less than those for soil profile 2. Estimates of the differential pennanent horizontal displacements between structures can generally be taken equal to the difference between the permanent horizontal movements of each structure, which will be less than 1.7 inches.

The permanent horizontal displacements for the EPRI hazard results are about one third of those calculated using the NUREG/CR-5250 hazard results.

Transient Horizontal Displacements For soil profiles 1 and 2, maximum transient horizontal displacements at the ground surface are 1.05 inches and 1.12 inches, respectively, ii

l Differential transient horizontal displacements can be expected between any one building and the surrounding ground, between buildings separated by some distance, and between ,

adjacent buildings within the Power Block. The differential displacement can be  !

conservatively taken equal to the absolute sum of the peak displacements of the building and the surrounding ground or the absolute sum of the peak displacements of the two buildings. l The differential displacement between a building and the surrounding ground can be conservatively taken to be uniformly distributed over a distance of about 25 feet from the foundation. The differential displacement between two separated buildings can be 1 reasonably taken to be uniformly distributed over the distance between the two buildings. l The differential;displeemant beiween any two buildings within' the Power Block is conservatively bounded by*the absolste sum of the peak displada6ents'of the two structures. The differential dispir-mt between any two buildings within the Power Block may occur abruptly across; construction joints or expansion joints between or within the building foundations or structures.

For a given value of peak ground acceleration, the settlements for the NUREG/CR-5250 j and EPRI hazard results are identical. '

The above described proceduits for determining the differential transient displacements are generally conservative procedures that do not account for the effects of flexibility and 1 rocking of structures. The effect of rocking is likely to be small due to the high stiffness of the soils at the Pilgrim site and is likely to be compensated by the effects of soil-  ;

structure-interaction or, i.e., flexibility of the foundations. The Condenser Tanks appear l to be the most likely structures to be significantly affected by rocking, due to their high center-of-gravity and small depth of embedment.

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TABLE OF CONTENTS I EXECUTIVE

SUMMARY

TABLE OF CONTENTS LIST OF TABLES LIST OF FIGURES Page No. ,

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1. INTRODUCTION 1  !

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I T1,1 PurpoiMi and Scope 1 1.2 Background -

1 4 1.3 Project Personnel 2 1.4 Authorization 2

2. SUBSURFACE CONDITIONS AND SOIL PROPERTIES 3 2.1 Introduction 3 2.2 Soil Profile 3 2.3 Existing Foundations 4 2.4 Ground Water Table Elevations 5 2.5 Total Unit Weights 5 2.6 Shear Strengths 5 2.7 Shear Wave Velocities 7

3. SOIL LIQUEFACTION EVALUATION 10

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3.1 Potential for Seismic Pore Pressure Build-Up 10 3.2 Seismic Stability 11

4. PERMANENT DISPLACEMENTS DUE TO SEISMIC EVENTS 13 4.1 Critical Locations for Evaluation of Permanent Displacements 13 4.2 SHAKE Analyses 13 4.2.1 Results for Profile 1 14 4.2.2 Results for Profile 2 15 4.2.3 Discussion of SHAKE Results 16 l

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. TABLE OF CONTENTS j (Continued) 1 4

Page No.

l 4.3 Pseudostatic Analyses 16 i

4.3.1 Purpose and Method 16 4.3.2 Soil Strengths for the Pseudostatic Analyses 17

! 4.4 Analytical Methods for Calculating the Permanent Displacements 18 l -. -

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4.4.1' ' Displacements Due 'to Slope Movements ' 18 l

4.4.2 Seismically Induced Settlements of I2 vel Ground 20 I 4.4.2.1 Densification of Soils Above the Water Table - 20 1

4.4.2.2 Consolidation of Soils Below the Water Table 21 1

4.4.3 Settlement of Structures 22 l

i 4.5 Results of Displacement and Settlement Calculations 22 i

I 4.5.1 Displacement Due to Slope Movements 22 j 4.5.2 Vertical Settlements 23 i ^

j 5. TRANSIENT DISPLACEMENTS 26 k

6.

SUMMARY

AND CONCLUSIONS 28 REFERENCES i

LIST OF TABLES

! Table 1 - Permanent Horizontal Displacements (Inches) for Soil Profile 2 Table 2A - Settlements (Inches) for Soil Profile 1 Table 2B - Differential Settlements (Inches) for Peak Ground Surface Acceleration of 0.35 g Table 3A - Settlements (Inches) for Soil Profile 2 i

Table 3B - Differential Settlements (Tehen) for Peak Ground Surface ""Acceleration of 0.7 g Table 4 - Transient Horizontal Displacements (Inches) for Soil Profile 1 Table 5 - Transient Horizontal Displacements (Inches) for Soil Profile 2 f

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LIST OF FIGURES l

Fig.1 - Site Location Map 1

Fig. 2 - Site Plan )

l Fig. 3 - Stability Analyses for Section I l

Fig. 4 - Stability Analyses for Section II Fig. 5 - Stability Analyses for Section III

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AL: :&--------

Fig. 6 - In-Situ Shear Wave Velocity Versus Depth Fig. 7 - Liquefaction Potential Based on Standard Penetration Test Dataj _

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t Fig. 8 - Shear Wave Velocity Input Versus Depth (Feet) for SHAKE Analyses Fig. 9 - SHAKE Results, Record NR0098-1, Peak Acc. = 0.23 Fig.10 - SHAKE Results, Record NR0098-2, Peak Acc. = 0.2g i

! Fig.11 - Sununary Plot of Volume Decrease of Sands Under Cyclic Straining  !

l Fig.12 - Volume Changes of Sands and. Silts,Due to_R,econsolidation After Cyclic Leding d: b; ~ '- 5' I , -

er l Fig. A Standard Penetration Test Vs. Effective Overburden Pressure - Unit < -r j l ,

Fig. A Standard Penetration Test Vs. Effective Overburden Pressure - Units 2 and 3 l

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1. DiTRODUCTION l

1.1 Purpose and Scope

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The pmpose of this report is to provide estimates ofliquefaction potential and seismically

~ induced permanent and transient displacements of the ground at the Pilgrim Nuclear Station. These are provided as a function of the peak ground acceleration.

The scope of the work is:

a. Perform an evaluation of the liquefaction potential and seismic stability, a : ,,cg.c=; w , z .. = t - .y .a y- - . . . n. ,,

~_.g. .a - a,p~p.py _

b. Estimate seismically induced displacementa of' sloping ground using published  ;

correlations of displeamant with ground mo+1on intensity.  :

m. . 4. .

c. Estimate seismically induced settlementa using published correlations of volumetric

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strain with ground motion intentity.

d. Estimmte transient displacemant of the ground and building foundations during a seismic event.

e. Present the results of the above items as a function of the peak ground surface acceleration.

1.2 Background _.'.

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'Ibe Pilgrim Nuclear Station is located in Plymouth, Mannehneetts. The location of the site is shown in Fig.1. _ _

A site plan showing the major structures, including the Power Block and the Intake Structure, is shown in Fig. 2. 'Ibe power block consists of the Reactor Turbine, Auxiliary Bay, and Rad-Waste Buildings. Three sections through the main structures are shown in Figs. 3, 4, and 5.

Subsurface investigations have been performed for Pilgrim 1, as well as for a potential expansion referred to as Unit 2 that was planned but was not built. The site for Unit 2 is located immediately east of Unit 1, and thus subsurface data in Unit 2 were also used to develop results presented in this report.

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1.3 Project Personnel The following GEI personnel were responsible for performing the majority of the work for this project:

Eugene Marciano, Ph.D. Project Manager Paul Joseph Project Engineer Edmund William.* Project Engineer Gonzalo Castro, Ph.D., P.E. Project Principal 1.4 Authorization of Stevenson &

• Associates?

Ibis work 'Ihe was authorized work was byGEI performed imder 'aConsultants,'

contract Inc.'s'Q signed by Mr. Thoma l

Program, which complies with the requirements of 10CFR50 Appendix B.

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3-l i 2. SUBSURFACE CONDITIONS AND SOIL PROPERTIES 2.1 Introduction The information presented in this report is based on Bechtel Drawings C1 through C9 and MIS through M29, Bechtel's Soils Report for the Pilgrim 2 site (Bechtel,1976), and a compendium of Pilgrim boring logs conducted for both the Pilgrim 1 and 2 sites (GEI, 1978). The Pilgrim 2 Soils Report contains GEI soils data reports conducted for the Pilgrim 2 site. In addition, GEI's project files for geotechnical investigations conducted-for the Pilgrim 1 and 2 sites were used.

Soil Profue m., M,, lid j " '. ~ - .

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The stratigraphy of.the' Pilgrim I site consists of 30 to 50 feet of comycted fill materials, designated as type A and typ: B. fills on Bechtel Drawing C8, above approxi-mately 30 to 50'feifof glacial outwash deposits. - De soil depositsfam underlain by bedrock at a degh of approximately 80 feet.

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The type A and B fills were specified to be compacted to a minimum of 98% and 96, respectively, of the maximum dry density as determined by ASTM D1557. The type A and B fills consist of sand and gravelly sands with less than 6% fines.

l The outwash deposits are very dense as a result ofloading due to glaciation subsequent to their deposition. The outwash deposits are granular, consisting predominantly of poor-to well-graded sands with some zones of gravelly sands.

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A comparison of the boring logs for the Pilgrim 1 and 2 sites and the~ geological history of the sites indicates that the outwash deposits have similar soi!' descriptions and ranges of blowcounts at the Pilgrim 1 and 2 sites. Bey have the same depositional history and were both subjected to glacialloading.

Plots of blowcounts versus effective ve.rtical stress for several boring logs from the Pilgrim 1 and 2 sites are presented . Figs. A-1 and A-2 in Appendix A. These blowcounts correspond to the outwash soils and are limited to blowcounts obtained using the standard penetration tests for which no gravel was observed in the split-spoon samples. It can be seen that the blowcounts at the two sites are similar and very high, indicating similar properties of the outwash materials in Units 1 and 2. Note that the shallower soils were excavated below the Power Block structures in Unit 1, as shown in l

Figs. 3,4, and 5. The range of effective overburden pressures at the excavation bottoms l

is shown in Figs. Al and A2.

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I Boring logs from the Soils Report for three borings (Borings 505, 609, and 610) representative of the Pilgrim site conditions are included in Appendix A, and the blowcounts are plotted in Fig. A2 where they are identified by the open square symbols.

l Conections for the effect of the gravel content were made to the blowcounts, if required.

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However, the majority of the blowcounts did not require correction. Nearly all of the blowcounts in these borings are in excess of 50 blows per foot, and many exceed 100 blows per foot and are similar to the blowcounts from the other borings.

2.3 Existing Foundations i The Reactor Building's foundation bears directly on the glacial outwash. The Turbine Building is underlain by about 10 feet of compacted fill above the glacial outwash. The

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~ foundations of the Diesel Ocierator Building an' ~thed Canw' tenner Tanks att close to the _

' ground surface and thd:W(oGided on the fill. The' Intake Strs5t6re~is fodnded iis the glacial outwash. The foundation elevations for these structures are given in Figs. 3,4, and 5.. .

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The following vertical bearing pressures were obtained from the Soils Report (Bechtel, 1976) for the Pilgrim I structures:

Structure Gross Bearing Net Bearing Pressure Pressure (ks0 (ks0 Reactor Containment Building 7.7 1.7 Reactor Auxiliary Building 2.3 -0.5 to 3.0 Radwaste Building 4.5 -0.3 Turbine Building 2.2 -0.9 Intake Structure 4.1 -1.8 The net bearing pressure is equal to the gross bearing pressure minus the total vertical stress in the soil at the bearing level prior to excavation.

The dimensions of the Condenser Tanks were estimated based on Bechtel Drawing C8 and a recent site visit by GEI. The tank load was estimated assuming that the tank is i

filled with water, giving a value of about 3.5 kips per square foot (ksf) at the ground surface. The load exerted by the Diesel Generator Building near the ground surface was taken to be 2 ksf.

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l 2.4 Ground Water Table Elevations The elevation of the ground water table in this area can be expected to undergo

, fluctuations due to tidal effects and normal rainfall. Based on observation well readings l conducted by gel (1983) over nearly a 3-year period within and surrounding the Pilgrim 1 area, the highest recorded ground water elevations varied from +4 feet at approximately 100 feet from the shoreline to +8 feet at the southern end of the Turbine Building, about 600 feet from the shoreline. 'Ibe corresponding values for the lowest recorded ground water elevations are +1 to +5 feet. The above values do not include the potential effects of flooding, storm surges, or other extreme events on the ground water table.

l The mean high water and mean low water tidal elevations from the nearest National

Oceanic and AtmospherkiAdministration tide station, located,in Bostbo, areM.98fand l

-4.56 feet, respectively; OEI (1983).1 :

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2.5 , Total Unit Weights

.. 4 Based on the data available in the Soils Report for Pilgrim 2, the average total unit weights are 126 pounds per cubic foot (pcf) for the compacted fill above the water table, 137 pcf for the compacted fill below the water table, and 129 pcf for the outwash deposits. Bechtel indicates in the Soils Report (1975) that the unit weight of the bedrock is 168 pcf.

2.6 Shear Strengths The compacted fills and glacial outwash at the Pilgrim site are very dense, as evi_denced by the high blowcounts in the outwash and the minimmn specified relative compaction values of 98% and 96% (ASTM D1557) for the class A and B fills, respectively.

The results of triaxial compression tests conducted by GEI were reported in the Soils Report (Bechtel,1976). Consolidated drained (S) triaxial compression tests and consolidated undrained (R) triaxial compression tests with pore pressure measurement were conducted for undisturbed tube samples of the glacial outwash, as well as for samples representative of the fill.

Drained Strengths Because of the dense compacted state of the fill and the dense state of the outwash due to glacial loading, these materials are dilative during shearing. Therefore, during drained loading they increase in volume (dilate). The shear stress increases to a maximum value as the soil dilates, reaching the drained peak shear strength of the soil. The shear stress l then drops off until the soil is completely remolded. At that point, the soil reaches the j steady-state condition, which is a state of continuous deformation at constant volume, I

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constant normal effective stress, constant shear stress, and constant rate of shear strain

! (Poulos et al,1985). The shear stress at the steady-state condition is the minimum l drained strength of the soil and is referred to as the drained steady-state shear strength.

The strength is equal to. i l l peak strength: So, = o', tan $,

steady state ]

strength: Sos = 0'r tan $,

l where So, = drained peak shear strength

, So, = draited steady-state shear strength .

^--

c', = effe:tive normal' stress on the failure surface FI' ' -

(, = peak friction angIe*

= the steady-state friction angle

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, Based on the results of the S tests,$ the average peak and steady state friction angles for .

I the outwash are 38.8 degrees and 33.4 degrees, respectively. Tests on samples representative of the compacted fill indicated peak friction angles of 40.5_ta .43, degrees ,

and steady state friction angles of 36 to 39 degrees. l Analysis of the blowcount data for Borings 505,609, and 610 using correlations by Peck, Hanson, and Thornburn (1974) and Gibbs and Holtz (1957) indicate a drained peak friction angle of about 42 to 45 degrees for the glacial outwash. 'lliis is higher than the value of 38.3 degrees from the triaxial compression tests. I Undrained Strength 1 w During undrained shearing of a dilative cohesionless soil, the shear stress increases gradually and approaches the undrained steady-state dwar stmngth as the soil is remolded, i.e., the peak strength is about equal to the steady-state strength. The undrained steady-state shear strength, S , and thus in effect the peak undrained shear strength as well, is given by the same expressions as for drained strengths given above.

However, for dilative material, the pore pressure decreases during shearing, resulting in an increase in effective stress and consequently larger shear strength for undramed conditions than for drained conditions.

The results of the R tests give a friction angle of 33.2 degrees for the peak and the steady-state strength. Note that this is about equal to the drained steady-state friction angle of 33.4 degrees, as expected. The results of R tests indicate that dilation continues i

until the pore pressure becomes low enough to cavitate. 'nius there is no well-defined undrained strength since it is a function of the effective stress at which cavitation occurs l which, in turn, is a function of the initial pore pressure and effective stress. In the i

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analyses presented in this report, the strength available during the earthquake is con-servatively assumed to be equal to the drained strength, thus neglecting the possibility of negative-induced pore pressures.

2.7 Shear Wave Velocities The results of seismic cross-hole testing conducted by Weston Geophysical at the Pilgrim 2 site in 1972 and 1976 are available in the Soils Report (Bechtel,1976). The results are plotted in Fig. 6, and the shear wave velocity ranges from 1,700 to 2,700 fps.

There is no compacted fill in the Pilgrim 2 area, and thus the cross-hole test results are not available for the fill. For the outwash deposits, the following shear wave velocities based on the cross-hole results were ~~

recommended by Bechtel.(1976) for design of pilgrim 2. .-. =.y : ,'c E, ? - yV i Ff- Mi#e k - -

, 3ng.y .

1 Depth Elevation Shear Wave Velocih

_ (feet) (feet) (fps) 0

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j 35 to 51 -13 to -29 1,950 .._...

l 51 to 71 -29 to -49 2,300

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71 m 49 m 7,M0 l bedrock bedrock

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bedrock bedrock 5,900 l We have also calculated the shear wave' velocities ~of theloutwashIsidls and the compacted fills based on blowcount and unit { weight'datR%hich were input'into- ,

empirical correlations, and'on laboratory testing data from the Soils Report (Bechtel, 1976). * @' a- - -

i The shear wave velocities were calculated using the following data:

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a. Blowcount data within the glacial outwash corrected for the influence of gravel content, if necessary, and an empirical correlation between shear were velocity and blowcount by Ohta and Goto,1978, as presented in Sykora (1987).

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b. Impulse shear wave velocity tests on undisturbed samples of glacial outwash.

c. Resonant column test results on specimens prepared by compaction of materials from bulk samples obtained from the glacial outwash. The bulk samples were l l

obtained from borings in the vicinity of the Pilgrim 2 cross-hole survey tests and are believed to be representative of both the outwash and the fill. l l

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d. Hardin and Drnevich's (1972) empirical relationship for granular materials. The l range of unit weight for the outwash deposits was determined from in situ field density test results conducted by GEI (Bechtel,1976). The range of unit weight of the compacted fills was estimated using the results of compaction tests conducted by GEI on samples of the outwash materials (Bechtel,1976).

The results of the various estimates of the shear wave velocities are shown in Fig. 6. All of the plotted points and curves in this figure are based on a ground water table elevation of +5 feet, i.e., a depth of about 17 feet below the ground surface.

The shear wave velocities estimated from the empirical correlations and laboratory test results described above fall within a relatively narrow band (see Pip. 6). _ However, for l

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the loweroutwash than those soils, plottSd,fidepths obtained larger thsit%25)miit froni the cr6ss'-lioWM 6y i@la' factor m w mem m.

The empirical2correlations are an average ofilata for a wide range of soils and thus, l involve considerable scatter.1 ne laboratory tests may underestimate somewhat the in

! situ shear way'e velocity becAuse of unavoidable disturbance of the " undisturbed"_ samples used in the tests. However, it is not likely that the laboratory tests would underestimate

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the shear wave velocities by a factor of two. k 6~

633 .

! The shear wave velocities measured using the cross-hole method appear to be unusually high even considering the high densities and high overconsolidation of the outwash soils l at Pilgrim. For example, recent shear wave velocities determined in glacial till in Boston were about 1,800 fps as compared to 2,500 fps for the Pilgrim outwash at similar depths.

The glacial till is denser.than the outwash, and it is also highly overconsolidated. De procedures used to perform the cross-hole tests for Pilgrim liicluded the use of explosives for the signal source and relatively large spacings between th'e' source and receiver holes.

The use of explosives for the source generates.a larger percentage,of compiessive wave (P wave) energy than shear wave (S wave) energy. De velocity of the S wave is typically about half that of the P wave, and thus the P wave arrives before the S wave.

The result of this is that the high P wave content tends to obscure the arrival time of the S wave recorded at the receiver holes. In addition, the large spacings (approximately 150 feet) between the source and receiver holes may have resulted in refraction of the wave through deeper, denser layers, which would result in an overestimate of the shear wave l velocity.

Based on the above discussion, it is our opinion that the actual shear wave velocities of the outwash soils are bounded by the results obtained from the empirical correlations and laboratory test data and those obtained from the cross hole measurements. Since the range between the lower and upper bound is large, it is recommended that analyses be i performed for both lower and upper bound values as shown in Fig. 8, designated as i profiles S1 and S2, respectively. Alternatively, one could perform additional cross-hole determinations using closely spaced boreholes (10 to 15 feet) and signal generation devices that enhance shear wave energy.

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9-l For the compacted fill, we recommend that the empirical / laboratory results be used, as shown i,n Fig. 8.

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3. SOIL LIQUEFACTION EVALUATION The potential for seismically induced liquefaction involves consideration of several related phenomena and mechanisms of failure. In this section we deal with the potential for pore pressure build up and with the potential for overall instability of the grotmd.

These two subjects are discussed separately in the following sections.

3.1 Potential for Seismic Pore Pressure Build-Up Two approaches were used to evaluate the potential for pore pressure build-up induced l by earthquakes of various levels of peak ground surface acceleration. _ s_.

n Errdd$nEQ9559 = ?YER = - - NE . ~

Thifirfipproach isl5EiYaE55"plSial"ciEtYiN611iesveedr4&6f p6re '

pressure build-up in sand deposits as a function of ~ earthquake seismic' shear stresses induced in the ground by the earthquake. The sites are level ground sites, and 'the soil 1 characteristics are represented)yLSPT blowcounts normalized to a standard confining I pressure (1 ton per square foot'[tsf)).

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ne fieldibservations consiste' generally of sand boils or sand volcanoes, which are evidence of high pore pressures and settlement. They are not an indication ofinstability, as instability was not possible for the level ground deposits. l As part of the liquefaction evduation for the Pilgrim 2 site (Bechtel,1976), the blowcounts were analyzed using the then available correlation by Seed et al (1975),  !

shown in Fig. 7. De curve shown was based on the data plotted by Seed et al (1975) I and shown as open and closed; circles. In addition, we have added to,the plot the more l

recent curves by Seed et al (1983) for ea@*= magnitudes 7.5 and 8.5.;

m -

The square and triangular data poidts in Fig. 7 ire the ah shear stress determined using Seed and Idriss's (1971) simplified expression for shear stress and are plotted versus the corrected blowcount for each standard penetration test in Borings 505,609, and 610. Only the blowcounts for which the spoon samples had no gravel content were shown. De square and triangular data points are for peak ground surface accelerations of 0.15 and 0.25 g, respectively.

For a peak acceleration of about 0.5 g, the shear stresses computed using Seed and Idriss's (1971) expression would be double those indicated by the triangles. Therefore, for a peak acceleration of about 0.5 g and higher, a few of the lowest Pilgrim blowcounts will plot on or close to the boundary lines in Fig. 7, indicating a potential for pore pressure build up in localized zones of the outwash. Note, however, that the Ni values of the Pilgrim site are 34 or higher, while the empirical data are based on sites which,

with one exception, have Ni values lower than 25. Thus the analysis of the Pilgrim case is based on extrapolation of field data.

A second approach is to estimate pore pressure increases based on the shear strains induced by the earthquake and the laboratory data collected by Dobry (National Research Council,1985). Dobry's data indicate that about 10 cycles of a seismic shear strain of l about 0.3% or higher is needed to reach 100% pore pressure. Based on the results of SHAKE analyses discussed in Section 4.2, an effective strain of 0.3% (defined as 65%

l of the peak strain) is reached somewhere in the profile when the earthquake has'a peak l ground surface acceleration of 0.40 g and 0.75 g for the high and low estimates of shear wave velocity defined in Section 2.7, respectively.

l 3.21 SeismicSf$bility~ W y[ .

', _=

_ N 6pe- - -

A liquefaction stability failure will occur if: 1) the undrained steady-state shear strength is less than the" driving stresses (i.e., the stresses required to. maintain equilibrium) and

2) sufficient deformation is induced, such as by an earthquake, to reduce;the shear l strength to theIundrained steady-state shear strength.  ? '~

For the dilatEve soils at the Pilgrim site, the undrained steady-state shear strength is greater than the drained strength. %erefore, a liquefaction stability failure, as defined above, is not possible at the Pilgrim site, regardless of the magnitude and peak ground acceleration of the earthquake.

A stability analysis was conducted to determine the minimum factor of safety based on the drained steady-state shear strength. The computer program, STABL5, was used to perform the stability analyses. The Modified Bishop method of slices for circular failure surfaces was used.

The geometry used for the stability analysis is shown in Fig. 3 and is based on Bechtel Drawings No. C1 through C9 and MIS through M29. He gross loads given in Section 2.3 for the structures shown in Fig. 3 were applied at the level of the foundations l of these structures.

The stability analysis was conducted using steady-state friction angles of 36 and 33 degrees for the compacted fill and the glacial outwash, respectively. These are based on the results of the triaxial compression tests discussed in Section 2. These were used l to compute the drained steady-state shear strength of the soil along the potential fai!we l surface.

l

ne slopes are covered with riprap, consisting of large blocks of stone. The riprap was I taken to be about 10 feet thick, as shown in Fig. 3, based on a recent site visit by GEI.

The unit weight and friction angle for the riprap were estimated to be 130 pcf and 40 degrees based on previous experience with similar materials.

l The maximum measured ground water table elevations described in Section 2 were used to define the phreatic surface on land for the stability analysis. These are shown in Fig. 3. The elevation of the water surface of the channel was taken to be the mean low water tidal elevation discussed in Section 2.

A search was conducted to determine the critical surface for stability failure. The resulting critical circle for the stability analysis is shown in Fig. 3. The critical factor of safety for this circle is 1.9. 'Ihis factor of safety is based on the drained steady-state

' shear strength of the soil. The undrained steady-state strength of the dilative soils is

- - - - higher ~than the drained strength, and therefore, the actual factor of safety against a ~

stability failure is'i6ucti higher. 1 *" 9.

• l l

t I I l l l

l

l 3 .

i l 4. PERMANENT DISPLACEMENTS DUE TO SEISMIC EVENTS I L

Permanent displacements and settlements were calculated at critical locations throughout the site. Newmark's (1%5) sliding block analogy was used to calculate the permanent

)

l Msplacements due to sloping ground. Previously integrated solutions available in the inerature were used for this purpose. Correlations of volumetric strain with the seismic shear strain (Castro,1987) were used to calculate the settlements due to seismic shaking.

The computer program, SHAKE (Schnabel et al,1972), was used to calculate the shear strains and shear stresses versus depth for the soil profile. The results of the SHAKE analyses represent the case of one-dimensional wave propagation through the soil profile in the free field. These results do not munt for the effects of soil-structure-interaction or rocking of the structure, which are discussed later in this report.

.n. .,

pp- .

g v.;.c pp~y .

r 9.

4.1 Critical Locations for Evaluation of Permanent Displacements Displacements due to slope movements were calculated for the Intake Stmeture, the Condenser Tanks, the Reactor Cantninment Building, the Turbine Building, and the Diesel Generator Building.

Vertical settlements due to densification during seismic shaking of the soil above the ground water table and due to dissipation of pore pressures developed below the water table during seismic shaking were calculated for these structures. Vertical settlements were calculated for the structures arid for the ground surface to enable calculation of differential settlement of piping and ducts.

4.2 SHAKE Analyses The computer program, SHAKE (Schnabel et al,1972) was used to analyze the response of the soil deposit to an earthquake. The purpose of the SHAKE analyses is to determine the maximum shear strains and the maximum shear stresses versus depth as a function of the peak ground surface acceleration.

The bedrock was taken to be at about El. -58 feet based on the available boring logs for the Pilgrim I site. The strata thicknesses and total unit weights used for input into the SHAKE analyses consist of the following:

Stratum Thickness Total Unit Weight (feet) (pcf)

Compacted fill above 18 126 water table Compacted fill below 17 137 water table Glacial Outwash 45 129 Bedrock Halfspace 168

'Ihe SHAKE analyses were conducted for the two alternative in-situ shear wave velocity profiles discussed in Section 2 and shown in Fig. 6. The specific values of shear wave velocity used for each of the two profiles are shown in Fig. 8.

~

For each of the two soil profiles, analyses west enaaeted for a range of peak ground accelerations for each of two seismic records. The two records were generated by Stevenson & Associates and are referred to as NR0098-1 and NR0098-2 in this report. l NR0098-1 is a synthetic record generated using the NUREG-0098 response spectrum scaled to a peak acceleration of 0.5 g. NR0098-2 was generated by changing the peak value of record NR0098-1 from 0.5 g to 1.0 g; the acceleration values for all other points in this time history are identical to those of NR0098-1.

Analyses were first conducted using NR0098-1, which provided strain-compatible moduli l for the soil profile. 'Iben, analyses using NR0098-2 were conducted using the strain-compatible moduli determined for NR0098-1. An analysis for NR0098-2 scaled to a peak acceleration of 0.2 g was conducted using the strain-compatible moduli determined by the analysis for NR0098-1 scaled to 0.1 g and so forth for higher levels of peak ,

acceleration. It is our understanding that this procedure was used by Stevenson &

Associates to conduct their soil-structure-interaction (SSI) analyses for the Pilgrim site. l 4.2.1 Results for Profile 1 Profiles 1 and 2 refer to the low and high values of shear wave velocity defined in Section 2.7, respectively. SHAKE analyses were conducted using NR0098-1 scaled to peak accelerations of 0.1,0.2,0.3,0.35,0.4, and 0.45 g. The record was applied  !

to the surface of the deposit. Results were not obtained for 0.45 g since the  !

program did not converge to strain-compatible values of shear moduli due to the severity of the ground motion. This indicates that for the NUREG-0098 design spectrum, a peak ground acceleration higher than 0.4 g is not possible for soil profile 1. In addition, for peak ground surface accelerations of 0.3,0.35, and 0.4 g, the peak acceleration at the bedrock is 0.67,1.9, and 11 g, respectively. ' Itis

= .

c NORTH 200 -

150 -

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@ CLACIAL OUTWASH 112 129 33 40 $E '

g ff

@ RIPRAP SLOPE 130 130 40 40 as g { 9,

[ g, 0 34 0 46 NOTES: 1. REFER TO FIG. 2 FOR PLAN LOCATION OF SECTIONS.

PEAK HORIZONTAL ACCELERATION ( g) 9

2. SECTION I IS BASED ON *YARDWORK PLANT StTE EXCAVATION SECTION ET (DRAWING NO. C-7) SY Stevenson & Associates

$ BE N Pilge 1 PEEE Wh. Mossoche Ptymouth. Mossochusette Stem MMS FOR SECTION I OEI Consultants, Inc. Project 92012 July 1992 Fig. 3

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APPRomasATE SCALf., FEET SOIL PROPERTIES i*-

LAYER 7M 7eet Ye YD =T= E sTRucrvRE NO. (pcf) (pcf) (dog) (deg) g- 9 O COMPACTED CRANULAR FILL 126 137 36 40 SURFACE A 0 k'

@ CLACIAL OUTWASH 112 129 33 40 h @ 1.0 ---- _

vARo.cRrracAL g ,,, g , wee U i i e NOTES:

1. REFER TO FIC. 2 FOR PLAN LOCATION OF SECTIONS.

W"

' I I I 1 i i

2. SECTION H IS BASED ON "YARDWORK PLANT $!TE "" f f f EXCAVATION SECTION C" (DRAWING NO. C-7) SY "' . "3, , i i BECHTEL s e a "8 87 o.24 o34o40

3. PSEUDOSTATIC FACTORS OF SAFEW EQUAL TO ANO PEAK HORIZONTAL ACCELERATION 4gp LESS THAN 1 CAN OCCUR MOMENTARILY OURING SHAkiWO 1

FOR PEAK ACCELERATION VALUES CREATER THAN (5.24 5 CAUSING A PERMANENT BUT FINITE HORf20NTAL FASPLACEMENT.

5 THE PSEUDOSTATIC FACTOR OF SAFETY OOES NCF REFLECT THE Stevenson & Associates STAaftRY OF THE SOfLS AND STRUCTURES AT TW SITE. Woteurn, Moosechusette Pil rim 1 IPEEE STA01UTY ANALYSES

? THE FACTOR OF SAFETY AGAINST A STABIU1Y FAs6URE AT -

% g OR RCTION N I

THE PILGRIM SITE IS GREATER THAN 1.9.

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200 - c NORTH iso -

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@ RtPRAP SLOPE 130 130 40 40

• o.s ea o,3

, o.s o7 NOTEL o ao

, 1. REFER TO FIG. 2 FOR PLAN LOCATION OF SECTIONS.

1 PEAK HORIZONTAL ACCELERATION tgs

2. SECTION lil IS BASED ON *YARDWORK PLANT SITE EXCAVATION" DRAWING NO. C-7 BY BECHTEL

$ Stevenson & Associates Pagnm 1 iPEEE WM hh Plymouth, Massachusetta STAOlUTY ANM.YSES FOR SECTION lil I

ont comsmie==== Ime. Project 92012 July 1992 Fig. 5

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l SHEAR WAVE VELOCITf (FEET /SEC.) l t 0 1000 2000 3000 0 l 7

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e Resonant column test results a Crosshole test results,1972 m Crosshole test results,1976

__ __ Range for compacted fills (based on Hardin and Drnevich expression)

Range for outwash deposits (based on Hardin and l Drnevich exoression l Best fit all dato, exc)luding crosshole tests Stevenson & Associates Pilgrim 1 IPEEE IN-SITU SHEAR WAVE Woburn, Massachusetts Plymouth, Massachusetts VELOCIT( VERSUS DEPTH

!k gel Consultants, Inc. Project 92012 May 1992 Fig. 6 l l

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' I DATA POINTS AND CURVE  !

AFTER SEED, ARANGO & j CURVES FOR MAGNITUDE

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0 0 10 20 30 40 50 N i- BLOWS PER FOOT i

LEGEND 9,, UQUEFACTION OBSERVED (1975)

O,0 NO LIQUEFACTION OBSERVED (1975)

A PILGRAM DATA FOR Q = 0.25 g l 1

i

$ O PILGRAM DATA FOR Om = 0.15 g l k

3

& 63 VALUES OF Ni > 50 l E Stevenson & Associates UQUEFACTION POTENTIAL Pilgrim 1 IPEEE g Woburn, Massachusetts Plymouth, Massachusetts BASED ON STANDARD

- PENETRATION TEST DATA 5 OEI Consultants Inc. Project 92012 May 1992 Fig. 7

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SHEAR WAVE VELOCITY (FPS)  ;

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>++e+ Soll Profile-2 i Stevenson & Associates Wobum, Mossochusetts Deformation Analysts Shear Wave Velocity input l Pilgrim 1 IPEEE versus Oepth (Feet) for SHAXE Analyses GEI Corntitants, Inc, Project 92012 May 1992 Fig. 8 i

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STRAIN COMPATIBLE SHEAR lilAVE vnocnv (ep.)

Stevenson & Associates Deformation Analysis SHAKE RESULTS

- m en0ntr-t Woburn, Mossochusetts - Pilgrim 1 IPEEE Record NR0098-1

-son. rnontr-2 Peak Acc. - 0.29

" ' "?"N"i--__ "' !**' ' _ _ _ _ -

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0 500 1000 1300 2000 2300 3000 0.0 0.1 0.2 OJ 0 200 400 600 800 1000 O o.01 0.02 O_03 STRAM COW'ATRE SIEMt WAVE PEAK ACGLDIATION (g) MAX. SIEAR STRESS (pof) MAX. SHEAR STRAIN $s)

VElDCITY (fpe)

- a m_., Stevenson & Associates Deformation Analysis SHAKE RESULTS

- son.rn0Far-2 Wobum, Massachusetts Pilgrim 1 IPEEE Record NROO98-2 Peak Acc. - 0.2g h GEI Consultants, Inc. Project S2012 Fig.10

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l Refereness: Silver and Youd,1972 ishibashi et al, Seed,1971 1985 Type of Teet: Simple Shear Simple Shear Torsional Shear l

Type of Sample: Ory Sand Saturated Sand Saturated Sand Relative Density: 45% to 80% 75 % 42% to 76% l Confining Pressure: 500 to 4000 psf 100 to 4000 pef 2880 psi Estimated relationship for fill above water table at Pilgrim 1 Stevenson & Assoc.iotes

SUMMARY

PLOT OF Wobum, Mossochusetts Pilgrim 1 IPEEE VOLUME DECREASE Plymouth, Mossochusetts OF SANDS UNDER CYCUC STRAINING OEI Consultants, Inc. Project 92012 May 1902 Fig. 11

5 . . .

2 -

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0 01 0 01 002 0.05 0.1 0.2 03 1 2 $ 10 Mesimum Cyclec Sneer Streen.%

after Castro (1987)

@ Tatsuoka et al,1984 Torsional Tests Clean Sand, loose to dense 0,, - 1000 to 6000 pel

@ Castro,1968 Triaxial Tests Silt, Undisturbed Samples B - 6000 psf

@ Castro et al,1987 Triaxial Tests Loose Silty Sand, Undrained Samples B , - 5000 to 6400 psf

-- - Estimated relationship for fill and glacial outwash below water table at Pilgr'n 1 Stevenson & Associates VOLUME CHANGES OF Wobum, M0ssachusetts Pilgrim 1 iPEEE SANDS AND SILTS DUE TO Plymouth, Massachusetts RECONSOUDATION AFTER CYCUC LOADING i OEI Consultants, Inc. Project 92012 May 1992 Fig. 12 1

PS PSAR A.*ENDMENT 26 August 31, 1976 S/sndard Atefrafm Tes/ (bhusperM) n a D 20 30 m so so 70 80 M Mo l I

, \ NOTE 5:

!. Blowcount.s owre deterinired ' rom 1066 l .

A borin s cVilled on the 1%7 to 1%8

.i i e inves ^ atton: BQ B-3L, B-37, 8-30

, B-55, sv.

i \ p 2. Effective overburden pressure nos AM ,

determined based on water /evel

-\ I\

• i readm .taken at the me of drillsng in each

\ .

e

\_ t , 3. SPT re ters to the sfonobro'pt/h*frofto/>

3000  ; test inoa'e hr accordance with l

I l u Ab7M Designation D-ISoto.

-> +: -

• i

' +. Blowcounts were plotted on Gibbs i SEE Q \ "' -

l t Notti Relative Density chart for

% 4f0%

' normaffy congotjggfgg 30 l (s).

NOTE 2 g , l I

5 Effective overburden ssures were

-y SOM

(

{t

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'* determined from tA followmg unit ove ' ts :

fS orated unit We ' t=B7p ds buoyant Unit Neig t = Mpc'f cf

' j 6. crefrom Alowcoun/s mqoV

,,. ihutos: Sand (S9) ana!Si// band gogg onh'tr> cot on soils cArssi n'

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• + Indicales blowcount was Srec+er than 100 blows per fact pgo ._ . . . .. __ .. ..

(t) See Reference ! in /sst E taferences.

PILf, RIM STATION

,. _,,._B._,,,,,,

Bra BIP: &# N3 + EM M N 2, Rarige of eescove stresses from conom STANDARD PENETRATI(*

1. % .A7% <@7, ,

of Su tc case et Reactor Commsmem TEST VS. EFITCTIVE jg Sm (addeo ty GEL 1962L OVERBURDEN FRESSURE-UNIT 1 FIGURE  :.o A-1 .

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s t e %1

.,%a gg*< ,ao Apdr 4 : 1' 11 251 8

ut :d : a

.I n::. J(.;:)(. 3 44 4,P s .t ca 4 ee e; n 4s *J .r : .%

h t  ?

ej{i ,l 2 ,,3 sed, * *1

,%R.wtI44y=Iw$..n.

a m $%.g% w tw4.c w24 t y  ? , . f *g u s 3, g.

5 E

4 -i +  %

• 6 .s 44*O 30 jg lja-s a

E bh#

f5f!

-e

!I" h.at.I .

. .5....

Ip *,if 7 hj'id'~

v  !

E.

g

.; o t j, I b .

{ ,e

, a d, 01 .. -

, e t... s C

4 o l 4R i

(h 4 . O o.

. g g%

g ' g) .ty l

l A .

4k R

ki S

(I E

.,- si k k I. k k k k k I k AN 4k

~ \

. w

.n, 1

1 1

i

. mo__ . , _ -m __..~u ..w_2 . , _ _ _ _ ,m.~~.__.m_ m _ _ -. _- _ . _ . . .-~ ., ._ _ _ . . __ ._.

4 303 310IACATEEl Ngees, ett E3449, et gBC L51ATEMI Vertled SEASSIG N A. DATE FTART/FWISE Ouseher 30,19tg / (eleber 10.1p*1

! CaesBO 9 8 an- CCBB SEEE W. A. at TOTAL DSPTg e1. 4 a Def1MD DT Amerleen Drtialum 4 Bertas Ca.. J. Tammara M A/ "A

_ .-- 3 L ftS u +96. 9 It DEPTE TO WATSE/DeTS 1A00819 ST W. E. Paal F. D. ' " _

S t. 8488PLS BA W p 1EMDTM RtMARIIB ON Sf2 E/T Tyg ODIL AND BDCE DWCarTEIS WL Buga Type y CD REC PElrTTRA- ADVANCE OF Bff LBED 70 (ASTM Dessf et and00486-491 4 A bl.

Oreve41r seat Wiele gradedi shoe 941 eMeter to seruanted grovel mas * '

else la met ebes $1 esup&astle thans grer SWL a St. , ela m 9 IE. 18 19 $*f/8 A . A 5 1

~,,e .ndDfwm

- m.tn.ee.mesem e .e. -Wt.hoe.gMF. 41 Rue gWe b% wth shed 41

. . ,,,- ,or

,1. ,, e.h

. , .... . J l

14. b .

f 4

> - 4 L 4 i , <

' l

16. , . Sent predenteestk the and mesem greseeds shed 61 Amo gravel as 8 mm amm  ; I sises otal layere and poetste to l4 em thestuksee of fhureemdv elks rests enmanner St 44 SA SS 14

I O' " E M '

l ,

= 6 -= .

14. , Ruter het hadder '~'

- . , J fsees T.6 to T.6 i di ,

!  :. 8 = J l

It-. ~

, , _ Sehr sand.Wttene gradens about M1 engtaly p&estle ihnese shee $1 hee grovel to =

l g g g 34 saa mas stees gror-treue SM-MLL

= 19 " *"

l 16

a. ,, = J

. eessus to 14.0 t

',,13 L J e-. . .

= 14 h k e . Asser het haalder <

I4* '

• IO* * ~

WASH: Stity seed. Widely graded; haes slightee ptestne sentales ene rert fregens d '3' I '*= 8 I R l . 35 m* te sise; brown.

I4 =

4<,, . J

' 14

- h .l 8<. L J emeg 19.8 8 ,

" 44

= = J SIty sad.Wldeb graded; shout SS1 sitghtly plasste Bases about IFI fine grevet; me '

se ta 3g a 33

. . rock fragmens SS em elset brese etth several rwt broen pasass e5W-SMt. <

4 , - -

i  : '

1 98 = J

-B ., ( $3. 9* .

Appretaste arme ewe determened frem oesh.

1 34 testug to 18.9 h = .l L . .

=

hto k

-4.; . .

k80 = d Send. tutformatine grained with several perhete af memum send; abeut 51 messestie <

4

-8 *L - fines; brown S Ph 5 j S4 44 8h 10 it j l 90 W u . mmmenre . _ susenemme, teses/4t af a leo-h 3'Of 88 l

4 . hasamee taatsg Se A to dette a sput epsen eempler 1) Test pit essenertas prosess la aporessee REC

. . . *, 16,,,,et m eemple resseeredg ore,s e e e.arta.g4.hertug. as essor leM reenage rnoRrW rrATrno. .e.

. . ,,m,, ,ee,m.,,,e

,,m >> mswo of .r. . r. we. eurr ,,o. ,

i 8

• esty tube F. Fause pesent P = Pesher N . Deessa a Bisse of 1964". 3Mrow reg 0W COMPANT 0+ Quesytsyg 0

• OSI team. Fahrwere it.1974 (Revisodi

.A W, er, . o -e e. .em -

PAGE I d 3 i

g 505

.er..t m. .,,,

m _

i l

1 i

l l 3033eG LOCAfinN h M. 4 % F le e t. en BIC LSI A TEDBf % ereleal g ga g ge, M.4 pagg gygggf ppggg l eteree 30, Ith / tetsbee 88. Stet j I l CAggeO g) 3 g CCdl8 8818 E 4- tm. TOTAL DEPTM *'* 4 DetLLSD ST 4merteu fictittu ud no,qu con J. Tenteg l

OpoggtD BL inELL) " M. l in DIPTE TO WATBR/DATR II ? h/ %4 LDGC ED ST _H.t. 199 4 0 l eomers l g L, GAJs P12 R A Af 8 f N TP RF ttA RIGS nN STf F MY PF $0tL AFD ROCR DWC3!Pf gogg WL Depth Type N AFC PD N FT RA. AfiV ANC R OF SFT (BED Til eB8 5ASTat Dket.49 and DH86 49) I T10M FIDRfNO ADVANCF A R No 08. in, ana nn-

- I l N $ +T/9 b. "

l

.gg ." Trteams roller l" . l 401 "

l

31

J l

.i .. .

l LH '

=  ; l l

.ig s ,

se.d. twform.n gr. .edl sie a en .ee n i. 3. e iAw i... af .iii, .

,, ,. u is fue e ei a i=ne aae.. ii- api. -

.x '-  ;

.u .. .

3a L

.le ., . -

'de

- =

sand. Medlem te nee grata d with arous Si e orse grainedi eleens one is == leur of modem to eeerse seedi ene 4 mm laver et mensdaede stko fine sand; browe 6PL 2

$$ 163 3 km. 18 la L41 l. ,

31 . .  ;

4

. . l

44

=

]

.s4 , ,,

M4 .

J E. " Sead. t utfornoftas gratned; less than 4*' asap 4astle flees 4 about l'. subengular flee " .

gravela several 3 3 mm tatet eneeues le flee sand levers; brous SPL

$9 184 S h. 18 la

, 48

.g a a. . J

_u  :. .l

.M a ,, -

.l Ls3 l- .l

1. . Saad. Medium te fine granmeds ebens S; aceplastle flees; abena il subreunded fine .

3B .. =

gre,ets several 3-4 em thlet levers sikt has seed; brewe tSPL "

810 le S ta. 18.8 le

>' ele eseed 63 to

$f R 4dinared

• M -'.

NW eestas to St .

A

~

'M

J '

i l

l lead, tarered. 38 mm soit flee send stra abe;s it' noms 4astle Game; to em Rae

. Se ggg gg, g g,, g, g gg -

to mesum sand. M tr. eseplaatte name, ebens 19 esarse sand and flee 1

gravels 114 se eteen meSuse to See sand abas 19" eseree saad and itse grovels

.H .",, . #" 8 E .

Advessed NW

, essess en se R _ _ , ,

i n . samadarg , __ _ resissamme, blese/R of a 444 3, NOTES nemmer faDieg De in, to drive a sent**pese sesasier 1) Test get e,wstertag eveten to spersaan arc . Lang1h et sample rose seed dustag bortas, se esser level reedtage  !

e . spam esesm esmans $ Grune ease ented eed. Pf!4 AIM STATION NO. 600 U . Uedtsended samples LtTT NO. 3 l

g.gw y p,,, 80BTON E9 BON COtt PANT _ _,

F. Pimme pesem P . Pasher

0. Osterters c . C BI Dnie. F ebrw es t r. i s u #'d ***

- P , ~ . m.c .i ~ ,,,, , ,, ,

505

. . . - .... e.,,

N ee

_ - _ _ . . . - _. ._ . . . . _ . . . . . _ ~ . _ . - . _ - - --

l 1

I S EAR BdC h.A- DAT8 ST AR T/ FDsSN e etshe r 10.13't / v =eam e 1a. tan 303 psG 1,DC ATION N9994. et i *4449. 4 9 OEC LSW AT1001 % Mlr el CORR 8118 tL na. TUT AL DEPT 1s SL6 fl DalL12D BY w arumsfM11h. A n came. 1 T -

CA408G S t =-

1

24. 9 fl DEPTE TO W ATES/DATE '1 b ft I 48 LDOGED ST w.D. f9" D. 3 ***er*

l Caos[D EL Insin

' AMP tf7 F M yf4 80lb AND ROCK DESCafPTIDee El. SA88 PLE  ? s Nr-?tt e r u t a ns one

• Bff tarp Tn (A.tT M D34st-49 amal D94 AA-49 88k Dese Type T nic PD NIT R 4 40VANCF or and Tyg gyggm anggucy A A me. ta. In. ann m-

'88 31/t taen Trteene rolled '.* *

-88"- bit.

. =

.e <

Advaased Nw easing to 43 R .

, j 43.. i

• ant. Levered; ts em medhsm sand etth ahoia 19. enesdaatte ftmee and ebens it -

Lu

=

eenroe eend and nas gearets SS em ftne send etG sema St. ensadastle fw; to saa mesten to Bae sand eith ehed S^. asagdeette nasal let snm fine seed with "

.44 - dit to 3in. 11 le -

ebois 19 W=atte fines, eemains wwent 8 inen lavers essemen eends bm m

=

Lse

.44 . ,,

>:.4A 4e - L 4dvsmeed sw .

easlag to es a Eapd. Reedom te noe greened; ebens IF. eef=essular to a ' gro,ea to se em

[, te  ; ""

maatmens stasi shout 4, eeerse san.is she- 9'. annpeastle ftnee etP),

$13 81 8 te. 't le

=

=

=40 ~.

l. =

= ft

=

. 1

.sg -.

.94 I

Cesve4 fek en  ; .

.g4 ..

refler bei et fl A

- Sand, tendeum te nas gremdeds strettfledpoos Den f. eamp4esete finesi ene 3 mm =

- fg 3 in, 13.4 layer eart brees selv nme send; brewe 8 f*,

414 las le L Y

• M ..

P '

Lfe Gesved fels en reiter be ft.S A ,

6a .,

""' Gesvol feh en rene, >m. .e n ,

.ag .,

~

W 1and. Flas to medeum a rstned, lees thes r; nonpleetle fines; trees (5P).

h93 b

.e. , . $15 1H 1 ts 10,5 18 -

Cavius of hele et ""',I_ '

-" *

• 74 is es R Apprettmets strela enange determined f reva eneht l

.u

.H

' =

44 .

l. 33

- Sim. Raptd eletance reactions lee dew strengtAs less than 31 ocettered e.edium saad

.gg 9 g rutes: erw 40 mm thtet laver eth, medewe to fine sand with ancia v. fine se

"=

i 914 (in Iha. 18 17 heBer het to 91.0 angear grs,W (M LL l (39,9) (B fTSROCK)

. ;go a en d en '

N . Miamened penseretsu ressaamase, basure/h of a 146-h IsoTg bammes ammam M h. to drove a sudis,eseen asust*' 11 Test yet doettartug 8veten to secre85em Rf C . Langth el samste recovered dettag tortag, no water leest readinge Pt14 RIM STATION NO, see 8 3 pia spean sanota $ Ceemadesame in Blows of 31'G. M/t*, and 91/8". L1(FT NO. 3 y , ggy ,,,,g,, BGTON EDBON COM PANY

$ = Sehr mens II.Dumase F fand sesem P . Piaster Date.Fersary If, W?S iReviseds O. Oneerters sAMP On.naside sameer af saanpung epse O = gel Pact s of _ 3 l 505 aer*e. wnse s 's i essww1s.EAb apas.gnens .ag

O BO9 00G JDC AfluN #M3 ode F4361. 94 SeC Laua TEN venel ggaa880 NA DAT$ 8f48T/rgeon 88ereh 18. late / teore'. I4. leve t

>- w COAS Of33 WA nm. TOTAL DSPTM 87.I R DelLLED BY Amertess Detutag 4 Swtag co. . E. Ause f CA40sG10 fil. SI DEPTM TO WATER /DAfg fa) M# LDOGFD 97 w. E. FIN CROLisD ELittst *11. 99 ft g l. SA M P t2 SA MI LTNGTH Rf hLAltKS ON 5!2 E ?T Yl't SOIL AND BOCK DSBCRf9Pttuse RISL Dept Trum ADVANCE OF Bff t,BFD TO sAFT u Deest.ee end D6484-401 n Uf3 REC Pt W E7 RA.

ene TI W tosgua ADVAMCF A A me. Gn . la hoRf W

11. in turned weuseau e in. Treceae tend. wWo4v areded. ese emeut set seenguter to serended g,emi to se me  :

bescentie dritung rester mm snea, euoi nmen. sw)  :

. e.1 es ate. 14. s is Se < , ,

= en

-3 '

Oeenslanal echhted

, and heuMere near ,

1

e. ..

te , , ,

l Le i

~ 2 mener neu we n--ery.

l g,, (s.t Mt I ls. Mene 8 ) gg g ,es,,,g L6

• A

. 6.04 14 ' ..s . j f

It < . <

" 38 ~ "

sand, thsferanly graded measun to ame greenedi ehest $1 . amass emmenes 6-3 30 $ te. 11 16 eseestmaal M W W s h h M I

! le < . ,

1 - t:

I l

, . =

l i

e. . <

I L le *

• I

. - Sand, lavoredt Se en ettente estav Rae sends 35 se sleen mates to eenroe send 5 6-4 et I le. 11.5 la suh enee e6r gremia M me em See a maeum saadi12 mm eh 4< . flee to seerse saad; 15 mm oneen See send weib a trace maeum senda 36 na hae ,

= le send week a trase af has gravels brous. WP) "

4 g -

k 3 . <

, Lo t Seed, tavered; SS me sleen fine to eserne send with a trace ftme gresets SI en *'

! etena Ane to medtuum send; It me famely laminated stil, One sends 175 mm elms .

I . s.s te a en. 18 to  ; unidee m Ane te medlem md wHh a treme mrw eend and Ame gmmig bme. SN e '. .

- at

~ .

-t ,

, 14 - W '

l l

Send. Uusformly stededs nas greamed; aheel $'d moeplastle flees: centelns severel I

• me thlet levere af stity nas send esd eseestenal small broen apase ever sep of 4 ', S4 1 12 2 te. II. S 18 -

semple bem SP) <

F to L ,

j .e ,

' Se

. - J l

l

= .

I .

• 4 ,

Centeusd) y, p - meadsed sonstrettee restatanse, blose/ft af a len lb MteOTW Dmm e ta. m $w mig W 4 R. Beme-hammer lelttug Se te. to drive e epitt.epsen sampler REC - Lesette of semele renovered de M tu5 M M to dw we.

CBI AAer bortag esapleted. 3 it. ID WW

PILEltIM WT ATION NO. Geo I I""*

0

• eestag tast,ajled to 97.1 R. but see 0 - Umstsermed essasise L'NTT NO. I greed erberkAlled.

g

,0 e .. m. . e .m.r to.el re .e t ---Av_

. - . . or .rm. o.ed.

4) Bleue ed 50,T**

O. Ostestors C Ctl Dans- Mer 19.1*Te

, A .. oD . O.et e.eter e, .am- -

PACE 3 et 3 609 m_ -

e psC LDI A TE)N %

• mar el S E A p sto NA DAf 5 87 ART /F9tSM Meb l8.18?8 / asareh le.1978 90AptG LEJCATIG8 W9e58.ese Fl351.34 M4 TUTAL DErfM ett a pe:LLgp of Ameneen Drtittag 6 hortas Co. . K. Alles Ca80sG ID fil.19 b m com g el3E et l

DEPTE TO D ATER/DAf g

• A/ LaGCED ST **E M" GnouwD SL eMW 11.89 n

81. SAMP12 4A M p LDdOTN Rt eda RAS ON $12f *TTPE gott AND BOCK DilBCSIPTIDess agc ADVANCE OF BFT LEED TO (AM M DBeef-et and D5444-es)

( M8L (mesh Trea p OD Pthfi AA.

and TN)N DoggMO ADVANCE h h me. 1 ta. BAR N (Jamesed bene notag 4 ta. Treeene lead. Layord; SM Inn eteen umsform flee to modem eendt 64 mm ehghtly "

beatenho drtinag roller bu strettned has to seeree sends to mm altgtaly othy flea senda crasse brown.

- 5.1 las tis, la le used run bm and tee. SP)

-le . ,' ,

. 33 3

1 i ,

all ' , "

~

"I*

- sand. tassersty gredada manne te noe greamed; treee af e<=ree eendi eemein.  :.

9A 11 10 seestal I am b&seiretatead levers; broes SP)

S-8 17e 44 < 4

=

L34

- r. - _ . . 5, J

=

4

.le <, ,

= .

. se l '

18 .

l  ;,4e L Saad. Layered 4164 mm eiena See he eenroe senda 3 me eltty has saad 46 me flee f l s-9 las 8 km. 11.8 la to meeems send web s trees eenroe samen M me themty lamanssed othy Ane sends j ad flee aandy adhst $$ ene e$ly eMty See to eenroe send sad shoes SS sub- "

• rounded Rae to searse grosol up te la uma broen. SP and WW)

-80 <

L41

- a l

33

. A Lee 1

~ Send. Imperah t?6 mm aldsty graded Rae to esarse send and tense er subengular to

• le embruended gesvol to le om emb eeveral tusseheds of etery has senda 96 mm

.g e <, s.le 131 1 In. 13

,l L sitettly enty fine to sessee send and gtsvol to la maa breen. igw)

..e

= $

-34 . .

= ~

. ee n.r l

b 5 l -2 * ', "

", l 1

- 50 6ead. Lavereda 35 mm ftse la medhma sends SS una Rae to memesm amad and fame

[

le gengla 31 mm See to seeree sends se am mademn to seerse send; 100 mm l I g.3i in 3 te. 12 '

n==a ased attiv nee send and I am layere et aika to em Auty Rae to eenroe  ;

' , l

' seed; broen. SW and gP)

-38 a ,

~ ~

~u 7  :

• 33 ', . , 1

-u

~

m }

=

sand. Lawred;ISS mm eteen entform flee to medium saad; ISS me ettshtly othy

• noe se medlem sand and subr==^ d to embasgular gravel to se mas brossa. SP

'I'  ; ed 8,)

l ' , , , , 3-13 107 3A 11.6

- 1g q l

. E I

, 3. . .

~

~

! ' 58 l

iCounteued)

-88 aIse j se . stuutard penetros>m resistence. blooedt of a leo-b %fgg a ta. ID SW eastas to 4 R. Beste- l hammer felltng 30 tn. to detvo a aput-spean sampler alte drtilise fhad used to edsnace hone. l Rrc . unsa of e nete race owd m ano, he,i,g ese,iet,g '= M "^rio" "o aa e -**i=e - es*e & ceme* e-- emieg muisd to e .i 3 n.se, b D aNw UNTT NO. I B U = Undisserted eamoine greaed er bachfllied.

131 He esser leve4 roesage ehtained ese to M NCOMfAg __, , ,,,__,.

s . Snetty este p . Dumme bestemate dettinag field. j

r. ri e ,tsee p . Pnete, G.C Dens- Mov le. lets O . Ostsetme

..., oo . ~m se.mm.r ., EI e.m.it. .-

PACE I of 3 l

so9 eerveuro..= eamcasens ce

[

! l I

800 D8G LDCATEJM N'88 8. 488 E835 t . 34 esc L9d ATIDN Vretkel BEAR 0tG MA DATE START /P3ssa him,mk ia_ inte / *=" ta _ ines NA 8tI DetLLED BY AmeMeas Drituer a tortur Co. . E. Alana

( CAsetG E) th. * = Com 8 Sitt in. TOT AL OtPTM t CROUWD ELindSL) *tt. 84 A DEPTE TO D ATER/DAf t II) A/_ IDOGEDBY W. f. Ptit E l. SAar P t2 EAMP 1.EWTM agas. ARKS OM Sttt/77PE totL AND tOCE DggCBMgue Inst Dessa Type y 0D mEC PENtT RA. ADV AMCE OF Brf ISED TO LAff M D8487-40and (Seas 40) ,

and TioM soseeG ADVAMet I R R m*- to , ts . sostoso , 1 1

80 Unsesed hele ustq 6-la. Tetsame Sand, tktierely gradeds moeum to Ses gresand ett about 8% emarse send and .

bemnamme delittag rener bit eben 38% sehensidar greret to 10 ma man sises breve. SP) -

6-13 tee t in. 13 38 med "

5 de .  :

.m

. A

. A as

= M '= A l seed. 14perada ans se moeum sed sah ' i men layer et euer ans sed " ,

,g , se man has to nadamm asad ett sham sl -- ' to subungear gresat as l

',. es s.te m 8 ta. 18 tt '. la en man, senes bem, en --- ----+---

q l

. - . j l

4e <, , ,

l

- es 46 - ' ,

= fo -

sand, taperesi las ma moeum to eenres send use a trase ans emas and ens "

l j

s.:s les te. Il to gre' ens tu me emety stretsnad aan to amass saada to em amessa to emesse

, . mand est e tross reams se susweed ene greests besme, pP)  ;

M. ' '

l =N "

5 l l l i

l '?s. e') J j

1

.a <, , .

1

= T4 " "

r <

L -

ors ==Up samL tidser gredada sensus to Ans ges abse as4 2 l

to aC i a ta, angder ene to eaaree grami to m em man. essa samasame aneus 1st ass-

-Se . s-ts tas to ts l M % broen, m  ; '

l

. = A l

-u -, .

l

= 99 " "

l

. - J l

-u . l

.= 04 ==

Semi. thiderely graded 4 moeuun to nas gredned ett shout 59 emerse sand sad s-t ? m 8 in. In 18 - shout 10% Aas to eaarne subreeded as subesgular grosol to 30 mm. sad abma ,

- og am asesi brees. ar)

j 40

' ,. 81

~ -

• g

-es ., - ,

= 64 "'

• ' ,. Sand. Wldety gradedi meena to fles greened with shout 10% emerce send and sheet 8.8 to A' S-te M) 3 in. plased la N in ans a swee em to sw gW k M ans. and abad M

' , slighth 3dastle Reesi broen. SW)

, -45.5'- *'

N T. t .

1 PItat AS LE BEDROCM . BorTOM OF SORDdQ

)

i i

i n E8

. standard ,

-, tem.gpenetra.rle

. resistanse, bi.ses/n

~e - sam.ie af a 846-b e,e _ eld .d mse ,.t

, REC . taugth d e@ resemed NH MAN M. 800

. . . d. e.mese e  ; cro-.eemt gg}i.g ape.r bortaganed . im n. .,.

h '. 3.es. E) Nu .ene.

gre.ied varr m. :

B U . Meses1had samoans beststues. DorTON EDBOM COMPANY

! C (3) We emer tard e eheshted dies to -

g 8 - M tube W.Dunese bemasmses drukas

- +-

F. Fmed peseen P - Petsher M) eseo et 40/5' eum 1464 hammer and O. Osassents C . GEt et/e" and st/6" estag 3e64 hexamer. Dame- Iday it.1994 SAMP OD . Oisette semeser af canesens opse M Shoes of 130/8" estig 144'4 bemmer and j 85/S" and les/ea se64 kamm e swoo na w e =Yfso4 tamme,er. p,gg g ,g y see

$11 estaus SM-Ra hommer.

A . ._ . e .eu.e

., .. . . . . . - . . _~ _ c~_---..n.n _ _-.~. ,. . ..

e 1

l 1

BoepsG WCATpast 49d ?t. e5 f1490. 68 98C L8dATWIf Wertteel SEAS WG MA DA78 START /P9fSN _ adore 9,1996 / taeren 13, late t

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91C2672-LRS2-007 gel Consu tants, Inc.

i 1021 Mam Street Winchester, MA 018901943 617 721 4000 July 10,1992 Project 92012

_ _ Mr..Domas J. Tracy Vice President Stevenson & Associates Ten State Street Wobum, MA 01801

Dear Mr. 'lYacy:

Re: Uncertalaties in Soil Failure Analyses Pilgrim 1 IPEEE, Pilgde Station, Plymouth, Mananchusetts his letter is in response to Dr. Fred Mogolesko's request at the June 11,1992, meeting with BECo, Stevenson & Associates, and GEI Pnnsnitants, Inc. for a letter describing any uncertainties ===aciated with the soil failure analyses.

We understand that the Probabilistic Risk Analysis (PRA) involves awan point estimatian to evaluate the seismic vulnerability of the plant, and therefore, best estimates of the soil displacements are needed for input into the fragility analyses for the PRA. Due to the fairly high density of the soils at the Pilgrim site, it was datarmineA prior to pt ing the soil failure analyses that relatively simple methods of analysis would be adequate.

Some reasonably conservative assumptions were made to apply these methods of analysis.

The uncertainty associated with the analyses conducted and the conservative nature of the assumptions made are described below for each set of results from the soil failure analyses. These results are: 1) stability against liquefaction failure, 2) permanent horizontal displacements,3) settlements, and 4) transient horizontal displacements. Next, the earthquake input is discussed. De impact of the soil parameters, the earthquake input, and the methods of analysis on the results are then summarized.

De uncertainties for the values of shear wave velocity for the soil strata at the Pilgrim site are discussed in the revised report prepared by GEI. De analyses of settlements, Cimcord, New Hampshire Ralcich. North Carohna D nver. Colonh

. l l

l Mr. 'Ihomas J. Tracy July 10,1992 l

permanent horizontal displacements, and transient displacements were performed for two shear wave velocity profiles that are upper and lower bounds for the probable actual in situ values.

i l

Liquefaction Stability l The blowcounts and laboratory test data for the Pilgrim site indicate that the outwash l I

deposits are very dense and highly dilatant. 'Ibe compaction specifications and laboratory test data indicate that the fill at the Pilgrim Unit 1 site is heavily c:-i+M and also highly dilatant. The strength of these highly dilatant soils would be higher for the undrained shearing that occurs during cardw==ka shaking than they are for drained shearing.

4s A cannervative determinatian of the factor of safety using the drained steady-state signs of the soils gives a value of 1.9, which is quite high. Typically, factors of safety of 1.5 are considered to be adequate. Since the soil response will be undrained l during shaking, the actual factor of safety against liquefaction instability is much higher !

than 1.9. I Permanent Horizontal Displacements A finite amount of per==aaat horizontal displacement can be expected to occur during shaking as a result of movements of the ground downslope toward the waterfront. These displacemante have been calculated for the structures as a function of the peak ground N08-l

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'Ibe expression used to estimate the permanant horimatal displacement boundinearly all of the values of displacement calculated for a large number of western United States

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care hg=6a acceleration records by Newmark (1965) and later by Franklin and Qiang I (1977). The characteristics of eastern seismicity were taken into account by using the l hazard results in NUREGKR-5250 and EPRI Report No. NP-6395-D for the Pilgrim site l to determine the values of ground velocity to input into the expression for displacement.

'Ihc values for the yield acceleration used to calculate the displacements are based on  !

conservatively selected values of yield strength. The yield strengths were conservatively I based on the peak drained strength of the soil, since there would be some uncertainty as l to whether the higher undrained strengths could be fully mobilized. 1 I

'Ibe values for the maximum acceleration of the potential soil mass subject to permanent movement are based on the maximum shear stresses calculated using the computer l program, SHAKE. 'Ibe maximum shear stresses are conservative for the reasons described below in the section on the earthquake input.

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t. Mr. '1homas J. Tracy July 10,1992 Settlements Conrelations available in the literature were used to determine the relationships between the volumetric compression and peak seismic shear strain for the soils at the Pilgrim site.

'Ibe correlations are based on test data involving several different granular materials and, therefore, involve considerable acatter. Mid-range values were used to determine the relationships. Better estimates can only be mada by using site-specific test data.

The volumetric compression increases with the peak seismic shear strain. The peak seismic shear strains calculated using the computer program SHAKE are conservative for the reasons described below in the section on the earthquake input.

i Transient Displacements , .

7

q~c h results of the SHAKE analyses w- one dimensional wave propagation through the soil in the free field and do not account for the effects of the weight and flexibility of the structure or the effects of rocking of the structure,

, Since the weights of the buildings are approrima'aly equal to the excavated soil that they

replace, the presence of the heing will have a small effect on the accuracy of the calculated shear strains in the soil.

'Ibe buildings were assumed to be rigid by taking the building displacernent to be equal to the dispLaramaat calculated by SHAKE at the elevation of the base of the building foundation; this is conservative since any fienWiity of the building foundation will result in smaller differential transient horizontal dispteu-- :- between the building and the surrounding soil

" +

'Ihe effect of rocking on the structures is likely to be amail due to the fairly high stiffness of the soils, with the possible exception of the Condenser Tanks due to their high center-of-gravity and small depth of embedment.

'Ihe differential transient displacements were taken to be the absolute sum of the peak transient displacements of the points being considered, e.g., the building and the adjacent soil, which is conservative.

h transient displacements were calculated by integrating the peak seismic shear strains calculated using the computer program SHAKE. The peak seismic shear strains are conservative for the reasons described below in the section on the euthquake input.

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l

Mr. Romas J. Tracy July 10,1992 Earthquake Input The acceleration time history used for the SHAKE analyses is conservative for determining the peak seismic shear stresses and strains in the soil deposit, which are thus conservative for calculating the settlements and the permanent and transient horizontal i displacements. h time history was synthesized by Stevenson & Associates using the NUREO/CR-0098 spectrum. For the same value of peak acceleration, the NUREO/CR-0098 spectrum envelopes the response spectra provided in NUREO/Ot-5250 and the EPRI Report No. NP-6395-D for the Pilgrim site. %e NUREO/Ot-0098 gives a maxi-mum spectral velocity of 80 in/sec/g compared to values of 30 and 18 in/sec/g for the NUREO/CR-5250 and EPRI spectra, respectively, he effect of the spectral shape on the peak seismic shear stadsses 'andistrains is

~

illustrated in Pigs. 9 and 10 in the report by GEI. De results for the acceleration record based on the NUREGER-0098 spectrum scaled to 0.2 g are presentM in Pig. 9. He results for an acceleration record for which all the acceleration values except the peak value are half of those of the acceleration record for Pig. 9 are presentM in Pig.10.

nus the ordinates of the insponse spednuu for the acceleration record for Pig. 9 are essentially twice those for Fig.10, with the exception of very high frequencies where the peak accelerations are ideatical. It can be seen that the peak seismic shear stresses and strains in Fig. 9 are about double those in Fig.10.

Summary The results of the soil failure analyses are conservative estimates of the soil dispIncemente rather than mean value estimates. He degree ofconservativeness for each set of results is affected by uncertainties ameaciatM with the values of the soil giarameters, the earthquake input, and the methods of analysis.

De soil parameters relevant to the soil failure analyses are 1) shear wave velocity, V,;

2) modulus reduction and damping curves, G/G_ and D versus shear strain; 3) density;

4) drained and undrained shear strengths; and 5) volumetric compression versus shear strain.

%e probable actual shear wave velocities at the site are bounded by the two sets of values determined and used for the soil failure analyses. The average modulus reduction and damping curves presented in Seed and Idriss (1970) for sands were used for the SHAKE analyses. The shapes of these curves have little effect on the resulting accelerations, sh ar stresses, and shear strains, especially for very stiff soils such as those at Pilgrim.

%e densities and drained strengths of the soils are well known from the results ofin situ tests and laboratory testing on both compacted, remolded samples and undisturbed

. Mr. 'Ihomas J. Tracy July 10,1992 samples. h test data also indicate that the undrained strength is greater than the drained strength.

The relationships between compression of the soil and seismic shear strain are reasonable.

The carthquake input record is reasonable and is conservative for the purpose of calculating the soil displacements. h effects of eastern United States seismicity have been taken into account by selecting the peak ground velocity, V, based on the harnrd results for the Pilgrim site.

W results of SHAKE analyses represent one-dimensional propagation of shear waves vertically through the soil strata. h effects of soil-structure-interaction or rocking of the structures on the settlemants and the perm-* horienntal dieal-*= are small.

Conservative approximations have been made to determine values for the transient displacements from the results of the SHAKE analyses.

Please call me if you have any questions.

Sincerely yours, GEI CONSULTANTS, INC.

Eugene A. Marciano, Ph.D.

Project Manager EAM:ms l

i I l i

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l . indicates that the largest realistic peak ground surface acceleration for soil profile 1 is less than 0.35 g.

SHAKE analyses were conducted using NR0098-2 scaled to peak accelerations of 0.2,0.4,0.6, and 0.8 g. The record was applied to the surface of the deposit.

l l Typical results for each of the records scaled to 0.2 g are shown in Figs. 9 and 10.

The maximum shear strains and shear stresses and peak accelerations are plotted versus depth.

l l Comparison of Figs. 9 and 10 indicates that, for the same peak acceleration,

! NR0098-1 produces larger values for the maximum shear stresses and maximum shear strains than NR0098-2. This result is not unexpected since theisingle, high -

frequency, narrow peak in the NR0098-2 record has'little 'effect on the soil profile.

This is confumed by the fact that the stresses and strains in the soil profile are about the same for the NR0098-1 record and for the NR0098-2 record with twice the peak acceleration of the NR0098-1 record. j Based on the above, the results of the analyses obtained with record NR0098-1 were used to determine shear stresses and strains in the soil and accumulated deformations as a function of the peak ground acceleration.

4.2.2 Results for Profile 2 l

SHAKE analyses were conducted using NR0098-1 scaled to peak accelerations of 0.1,0.2,0.3,0.4,0.5,0.6,0.7,0.75,0.80,0.85, and 0.9. The record _was applied to the surface of the deposit. Results were not obtainad for 0.9 g since ~the pmgram did not converge to strain-compatible values of shear moduli due to the severity of

! the ground motion. This indicates that for the NUREG-0098 design spectrum, a peak ground acceleration higher than about 0.85 g is not possible for soil profile 2.

In addition, for peak ground accelerations of 0.6,0.7,0.75, and 0.80 g, the peak acceleration at the bedrock is 0.67,1.2,2.3, and 5 g, respectively. This indicates that the largest realistic peak ground surface acceleration for soil profile 2 is less than about 0.7 g.

SHAKE analyses were conducted using NR0098-2 scaled to peak accelerations of 0.2, 0.6, 0.8,1.0, and 1.4 g.

Typical results for each of the records scaled to 0.20 g are shown in Figs. 9 and 10.

l The maximum shear strains and shear stresses and peak accelerations are plotted

! versus depth.

i

l Comparison of Figs. 9 and 10 indicates that NR0098-1 produces larger shear stresses and shear strains than NR0098-2. Based on the discussion in Subsection 4.2.1, the results of NR0098-1 are the appropriate values to use.

4.2.3 Discussion of SHAKE Results The results of the SHAKE analyses represent the case of one-dimensional wave propagation through the soil profile in the free field. These results do not account for the effects of soil-structure-interaction or rocking of the structure. Rese effects are not likely to have e substantial impact on the values of the settlements and the pennanent horizontal displacements. Rey can have a significant effect on the l transient displacements of the structures. In general, the effects of soil-structure-interaction would be to reduce the differential transient displacements.:ne Teffect

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of rocking could be to increase t$e trans6it dispikements.7HowesGNfairly high stiffness of the soil strata at the Pilgrim site make it likely that for most of the structures the effect of rocking will be small and will be compensated by the effects of soil-structure-interaction. The Condenser Tanks appear to be the.most likely I structures to be significantly affected by rocking, due to their high center-of-gravity and relatively small depth of embedment. If the issues of transient displacements

and thus soil-structure-interaction and rocking are found to be critical, a two- or i three-dimensional model of the soil and the structure should be used to more j accurately estimate differential transient displacements.

i 4.3 Pseudostatic Analyses i 4.3.1 Purpose and Method i . .

Pseudostatic stability analyses went conducted for the critical structures in the three

,i profiles shown in Figs. 3, 4, and 5. De purpose of these analyses was to determine the yield accelerations for each critical structure. %e yield acceleration is the value of horizontal acceleration which gives a pseudostatic factor of safety of 1.

The computer program, STABL5, was used to perform the stability analyses. The Modified Bishop method of slices for circular failure surfaces was used. Circular failure surfaces are critical for all of the cases except for the Intake Structure. A wedge analysis using Janbu's method of stability analysis in STABL5 was conducted for the Intake Structure. The wedge was taken to coincide with the bottom of the structure. This is equivalent to analyzing the Intake Structure as a gravity-retaining structure.

The geometries used for the stability analyses are shown in Figs. 3,4, and 5 and are based on Bechtel Drawings No. C1 through C9 and MIS through M29. The

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bedrock was taken to be at El. -75 feet, which is the deepest elevation reported for the boring logs at the Pilgrim 1 and 2 sites. The gross bearing pressures given in Section 2.3 for the structures shown in Figs. 3, 4, and 5 were applied at the

foundation levels of the structures. In addition, a horizontal shear load equal to the l

horizontal acceleration times the gross bearing pressure was applied at the foundation level.

The selection of the yield strengths for the fill and outwash materials is discussed in Subsection 4.3.2. The strength parameters for the riprap and the ground water ,

table elevations used are the same as those for the stability analysis described in i Section 3. l l

For each structure, a search.was conducted to determine the critical surface and I factor of safety for seker'al valdes"of horizontal' afcelsration. "1he'sesrch was

~

conducted for surfaces restricted to intersecting at least part of the stni6ture. The

~ I value of horizontal acceleration giving a pseudostatic factor of safety of 1 is the  !

yield acceleration. This value was detennined by integolation between the ,

acceleration values for factors of safety bracketing 1.0, as shown in Figs. 3,4, and l 5.

Note that the pseudostatic factor of safety will be equal to or less than 1 only if and ,

when the peak value of the acceleration time history of the soil mass exceeds the yield acceleration and then only for fractions of a second during the course of shaking. The result will be a finite permanent displacement of the soil mass and not instability of the soil mass. The method for calculating the permanent displacement of the soil mass is described in Subsection 4.4.1. In general, the peak value of acceleration for the soil mass must be substantially larger than the yield acceleration for the calculated permanent displacement to be significant.

4.3.2 Soil Strengths for the Pse'udostatic Analyses During undrained shearing of dilative soils, the shear stress increases until it reaches the value of the undrained steady-state shear strength. As the soil is  !

sheared, the pore pressure decreases, and therefore, the effective stress increases, l' For a highly dilative soil, such as is the case for the soils at the Pilgrim site, the pore water pressure can decrease to a net pressure of -1 atmosphere in theory.

However, phenomena such as release of dissolved gases due to the reduced pore pressure can limit the net pore pressure to much smaller values, perhaps on the order of-0.5 atmosphere. A reasonably conservative estimate of the resulting shear  ;

strength for the stability analyses conducted at the Pilgrim site is the drained peak shear strength.

i l

. The peak friction angle in the triaxial compression tests discussed in Section 2 was 38.8 degrees and was estimated as 42 to 45 degrees based on the blowcount data.

The peak friction angles in the triaxial compression tests for the compacted samples ranged from 40.5 to 43 degrees. Based on these results, a value of 40 degrees was selected to estimate the yield strengths of the compacted fill and the glacial outwash. The strength along the critical surface was calculated using the following expression:

Sy = o', tan $

where S, = the yield strength o', = effective normal stress on the failure surface

- $ -- = - friction angle of 40*-

m. -

n.w .

with o', computed assuming drained conditions. This provides a reasonably conservative estimate of the yield strengths.

1 4.4 Analytical Methods for Calculating the Permanent Displacements 4.4.1 Displacements Due to Slope Movements The displacements due to slope movements were estimated using Newmark's-(1965) method of deformation analysis. In this method, one considers the movements of the soil mass above the critical surface. Pennanent displacement of the soil mass occurs when its acceleration time history exceeds the yield acceleration. The yield acceleration is the value of horizontal acceleration that gives a pseudostatic factor of safety of 1 for the given slope geometry when applied to the soil mass in question. 'Ihe displacement is computed by double integration

~

of the difference between the acceleration time history and the yield acceleration whenever the acceleration' exceeds the yield value. Integration is continued until the relative velocity of the soil mass is zero. Integration resumes each time the acceleration exceeds the yield value. The displacement is assumed to occur in the downslope direction only.

The following approximate expression was used to calculate the permanent displacements.

' h V2 A D= 1 _N __

2gN A,N

. where:

D= Permanent displacement of the soil mass above the critical surface V= peak particle velocity of the soil mass N = yield acceleration in units of g A= maximum value of the acceleration of the soil mass g = acceleration of gravity his expression was developed by Newmark (1%5) and provides a reasonable upper bound for the displacements for values of N/A greater than about 0.15. It was determined based both on theoretical considerations and comparison to the results ofintegrating actual recorded acceleration time histories for different values

~ ~ ~ of yield accelerationc

_ ,g n f , __ gp .

x The above expression was used to c.tenlaw the permanent horiental displacement for a range of values of the peak ground acceleration. For a given value of peak ground acceleration, the maximum value of the acceleration of the soil mass and the peak value of the velocity need to be determined to enter into the above expression for displacement.

The maximum value of the acceleration of the soil mass was computed by dividing the peak shear stress calcu1=ad by SHAKE at the depth of interest by the total vertical stress at that depth. De soil profile used for the SHAKE analyses was used to compute the total stresses. For each critical surface, the maximum average accelerations were computed just above the depth of the water table, at or near the bottom of each critical surface, and approximataly midway between the water table and the bottom of the critical surface. %ese three values were generally ^close in value and decreased slightly with depth. The average of these three values was used for the ground motion parameter, A.

Based on previous work using acceleration time histories recorded in the western United States, the ratio of the peak ground velocity to the peak ground acceleration was estimated to be 48 in/sec/g for competent soil conditions. However, this ratio is dependent on the prevalent frequency content of the ground motion and thus can be expected to be different for earthquakes occurring in the eastern United States.

The effect of the frequency content of eastem earthquakes is taken into account by calculating the ratio using the hazard results provided in NUREG/CR-5250 (1989) for the Pilgrim site. Using these hazard results, the ratio was calculated to be 20 in/sec/g. The method of calculation is described below.

Based on the hazard curves and the uniform hazard spectra provided in NUREG/CR-5250 for a retum period of 10,000 years and a damping ratio of 5%

of critical damping, the median values for the maximum spectral velocity and the

. 20-peak ground acceleration are about 23 cm/sec and 0.3 g, respectively. Based on NUREG/CR-0098 (1978), the median value of the amplification factor (i.e., the ratio of the maximum spectral velocity to the peak ground velocity) for spectral I

velocity is 1.65. Dividing the maximum spectral velocity of 23 cm/sec by 1.65 gives a value of 14 cm/sec for the peak ground velocity. Dividing 14 cm/sec by the peak ground acceleration of 0.3 g gives 46 cm/sec/g, or 18 in/sec/g, which for the analysis presented in this report was approximated to 20 in/sec/g.

I A similar computation was performed using the median EPRI Response spectrum and peak ground acceleration for a 10,000 year return period. The result is a peak ground velocity of about 11 in/sec/g.

! The displacements were calculated for the NUREO/CR-5250 hazard results'l There-fore, the value of V was taken equal to 20 in/sec/g multiplied by the applicable l value of peak ground surface acceleration.

l l 4.4.2 Seismically Induced Settlements of Level Ground i

Settlement of the ground surface and structures due to seismic shaking is the result of: 1) densification during shaking of cohesionless materials located above the water table and 2) reconsolidation of materials located below the water table, shortly after the earthquake, due to dissipation of pore pressures developed during shaking.

The results of several experimental investigations of the densification of cohesionless materials above the water table and the reconsolidation of materials below the water table is presented in Castro,1987 (see Pigs.11 and 12).

4.4.2.1 Densification of Soils Above the Water Table .

The strains due to densification of sands and silts for drained conditions

, increase with the number of cycles of uniform shear-strain amplitude and the value of the shear-strain amplitude. The strains decrease with increasing initial density of the materials. Based on the available correlations, the log of the vertical strain was taken to vary linearly with the log of the shear-strain amplitude between the following values for the compacted fill at the Pilgrim I site, as plotted in Fig.11.

Shear Strain Amplitude Vertical Strain

(%) (%)

0.01 0.01 O.5 0.6

. -. _ . _ . . - - . - . - - ~ . . . -. _ - - -. ..

. l The above relationship is for 10 cycles of uniform shear-strain amplitude.  !

For an earthquake, the effective shear-strain amplitude should be used, which is taken to be 65% of the maximum shear-strain amplitude induced l by the earthquake. i The vertical strain increases with the number of cycles of shaking. The strain for 40 cycles is close to double that for 10 cycles. Cycles in excess  ;

l of about 40 produce little additional strain. l l

l Seed et al (1983) indicated the following number of equivalent, uniform i amplitude cycles versus earthquake magnitude, which is applicable to l earthquakes in the westem United States: I l .

Magnitude * ~ Number of & - CW "'

~

l ' Cycles l 8.5 26 l

7.5 15 6.75 10 J

t Earthquakes in the eastern United States generally have had a higher

, frequency content than westem earthquakes and consequently have a larger number of equivalent uniform cycles for a given magnitude. Therefore, the

! number of equivalent uniform cycles may be closer to 40 than 26 for a magnitude 8.5 and higher event. Therefore, we calculated the strains in the l compacted fills above the water table based on 10 cycles and then increased l the resulting values by 100%. l l a-l 4.4.2.2 Consolidation of Soils Below the Water Table l

l The results of experiments conducted using constant amplitude cyclic stress tests (Castro,1987) and recorded earthquake time histories (Nagase et al, i 1988) indicate that the reconsolidation strain for materials below the water j table is dependent on the maximum value of shear strain that occurs during l the earthquake. Based on the available correlations, the log of the l consolidation strain of the glacial outwash and the compacted fills below the I water table was estimated to vary linearly with the log of the maximum shear strain between the following values, as plotted in Fig.12:

)

i

Maximum Shear Consolidation Strain Strain

(%) (%)

1 0.09 0.01 0.20 0.05 0.50 030 2.0 1.0 4.43 Settlement of Structures . _ _ _ . . _ _

my ~ - .,

. m n e3,7 The effect of the structures on the vertical strains in the underlying foimdatinnLsoils depends on the influence of the structure on the seismically induced shear strains.

These strains are affected by the seismic base shear exerted by the structme at the bottom ofits foundation. The seismic base shear depends on the mass and dynamic response of the building.

The net bearing pressures under the structures range between -2 and +2 ksf (see Section 2.2), with the extreme low and high values ecs. dug at approximataly El. -25 feet. A stress change of 2 ksf represents the equivalent of removing or adding approximately 15 feet of soil. His represents a change in total mass, which is fairly small compared to the total thickness of the soil deposits of about 80 feet. Therefore, the seismic base shear exerted by the building is not expected to alter the seismic shear strains in the underlying foundation soils appreciably.

Consequently, the seinmbHy induced settlements of the buildings should be about the same as those computed in the soil column _at the elevation of the foundation base. '%; 1 4.5 Results of Displacement and Settlement Calculations l l

4.5.1 Displacements Due to Slope Movements )

The computed horizontal displacements for soil profile 2 for the critical surfaces shown in Figs. 3,4, and 5 are presented in Table 1. For a peak ground surface j acceleration of 0.7 g, all of the displacements are less than 1 inch, with the  ;

exception of the Intake Structure, which has a calculated displacement of under l 2 inches. ]

The results for soil profile 1 are slightly lower than those for soil profile 2, since the maximum values of average acceleration are slightly lower than those for soil l

J

-23 profile 2. This can be seen from Figs. 9 and 10, where the maximum shear stresses for soil profile 2 are only a few percent higher than those for soil profile 1.

The results for the EPRI hazard levels would be about one third of those calculated using the NUREG/CR-5250 hazard results, due to the lower value of 11 in/sec/g l for the peak ground velocity.

The computed displacements listed in Table 1 are those of the center of gravity of the soil mass above the potential failure surfaces shown in Figs. 3,4, and 5. It is not possible to compute differential horizontal movements from the results of the analyses. The computed movements are toward the shoreline and are identified by the critical surfaces in Figs. 3,4, or 5 and the name of the structures located within the soil mass above the critical surface. ' Thus estimates of permanent horizontal

~

i differential movements can be obtai5ed liy comparing thidmputed m&5diie for l each structure. For example, for a 0.7 g earthquake the Intake Structure would l

move 1.7 inches while the Reactor Containment Building would move only

~

0.2 inch, both movements toward the shotcline, i.e., roughlp north. .. Thus the l distance between the two structures would increase by about 1.5 inches. -. 1 i

i 4.5.2 Vertical Settlements

! l l The vertical settlements within the soil profile were calculated by integrating the i vertical strains calculated as described in Subsection 4.4.2 with respect to depth.

l The settlements are principally a function of the peak ground acceleration and l the depth and density of the soil strata below the level at which the settlements are being evaluated. 'Ihe settlements were calculated at El. +22, +12, +2, -3, l and -28 feet to correspond to the approximate depths of the foundations of the l critical structures. The results are presented versus the peak ground acceleration in Tables 2 and 3 for soit profiles 1 and 2, respectively.

The structures are founded at approximately the following elevations:

l Structure Elevation (feet)

Containment Tanks +22 Diesel Generator Building +22 Turbine Building -5 Reactor Containment Building -26 l Intake Structure -28 f

. 24-As discussed in Subsection 4.4.3, it is reasonable to assume that the building settlements are about equal to those of the soil profile at the depth of the foundation base.

The estimated building settlements are generally small, less than 1/3 inch up to a peak ground surface acceleration of about 0.35 g for soil profile 1 and less than 3/4 inch up to a peak ground surface acceleration of about 0.7 g for soil profile 2.

This is consistent with the high blowcounts in the outwash materials, which plot well below the Seed et al (1975,1983) curves in Fig. 7, which were discussed in Section 3. The compression of the fill above the water table is generally greater than for the material below the water table, due to its drainage during seismic shaking. _. =. ^

v~ + . i :37 %55 Differential settlements 'can be~bnuexpected within the# bon imp w

building and within the areas between buildings due'to natural variability of the compressibility of the soil deposits. ? Differential settlements can also be expected between any one building and the ground and between adjacent buildings, such as those within the Power Block, due to the different thicknesses of the soil strata beneath the various structures and beneath the ground surface.

In general, for a structure founded on individual spread footings, differential settlements equal to about 50% of the total settlement over distances of 25 feet can occur due to natural variability of the compressibility of the soil deposits below the structure. This is based on experience with settlements of foundations on cohesionless soil deposits. his can also be expected to be true for the settlements of the ground surface and at the various depths below the ground surface. For structures founded on a structurally continuous mat foundadon, the. differential settlement can be taken to be 50% of the total settlement distributed over a distance of about 50 feet. .

De differential settlements between a building and the surrounding ground can be taken equal to the difference between the total settlement at the ground surface and the total settlement of the building. The differential settlements between the ground surface and the buildings will occur over a distance of a few feet from the building.

The differential settlements between adjacent structures due to the different elevations at which these structures are founded, such as for the Reactor Containmen'. and Turbine Buildings in the Power Block, can be taken equal to the difference in their calculated total settlements given in Tables 2 and 3. The computed differential settlements are listed in Tables 2 and 3 for the maximum possible accelerations of 0.35 g and 0.7 g for profiles 1 and 2, respectively. Note that these differential settlements are in addition to the differential settlements resulting from natural variability of the soil deposits.

1

I 4

i l

. If the foundations of the adjacent structures within the Power Block consist of a structurally continuous mat, then this will smooth the settlements over a fm' ite l distance. The distance over which the differential settlement between adjacent i

structures is distributed will depend on the interaction of the foundation mat with the foundation soil with the building settlements imposed under the foundation l imprint of each building. If there are construction or expansion joints between any of the adjacent building foundations, then the settlement profile is likely to be discontinuous across the jo'mt.

"& ~

l 1

l l

1 l

l l

l J

4 i

. 5. TRANSIENT DISPLACEMENTS The peak transient horizontal displacements of the ground due to an earthquake were calculated by integrating the maximum shear strains obtained from the SHAKE analyses with respect to depth. This is slightly conservative, since the peak shear strains do not all occur exactly simultaneously. The resulting transient horizontal ground displacements versus the peak ground acceleration are presented in Tables 4 and 5 for soil profiles 1 and 2. As discussed in Section 4, the effect of rocking of the structures is not taken into account by use of the results of the SHAKE analyses. It is likely that its effect is small for the structures at the Pilgrim site, with the possible exception of the Condenser Tanks.

The transient ^ displacements'of the embedded building foundations..and,the. transient displacements"of the giround at'some distance away from the building would be different.

For the purposes of computing differential displacements of piping and ducts entering the building, it is reasonable to approximate the displacement of the' building by taking it equal to the displacement of the ground at the elevation of the base of the foundation.

Differential displacements can be expected between any one building and the surrounding ground, between buildings separated by some distance, and between adjacent buildings within the Power Block. The differential displacement can be conservatively taken equal to the absolute sum of the peak displacements of the building and the surrounding ground or the absolute sum of the peak displacements of the two buildings. This is based on the conservative assumption that the peak displacements will occur simultaneously with their directions 180 degrees out-of-phase.

The differentini displacement between a building and the surrounding ground can be conservatively taken to be uniformly distributed over a distance of about 25 feet from the foundation. For example, this gives a differential displacement of about 1-1/4 inches over 25 feet for ducts entering the Reactor Containment Building near the ground surface for a peak ground acceleration of 0.7 g for soil profile 2. For the same conditions, this gives a differential displacement of about 1 inch over 25 feet for pipes entering the Reactor Containment Building at a depth of about 18 feet.

I The differential displacement between two separated buildings can be reasonably taken  !

to be uniformly distributed over the distance between the two buildings. For example, l this gives differential displacements of about 1/4 inch between the Reactor Containment i Building and the Intake Structure and about 1-1/4 inches between the Condenser Tank I and the Reactor Containment Building over the distances between these structures. i i

The differential displacement between any two buildings within the Power Block depends f not only on the soil profile, but also on the stiffness characteristics of the structures and  !

on the nature of their structural connections, if any. The differential displacement cannot i

l

/

l

. be estimated based on the results of the SHAKE analyses, which represent the case of one-dimensional wave propagation through the soil profile in the free field. A realistic estimate of the differential displacement can be obtained by performing appropriate two-or three-dimensional soil-structure-interaction analyses that model the soil profile and the stiffness characteristics of the structures and the connections between the structures. The absolute sum of the peak displacements estimated using the results of the SHAKE analyses provides a conservative upper bound value f'r the differential displacement between adjacent structures within the Power Block. For example, this gives a differential displacement of about 0.78 inches between the Reactor Containment and Turbine Buildings for a peak ground acceleration of 0.7 g for oil profile 2.

The differential displacamant between any two buildings within the Power Block may occur abruptly.across_ construction joints or_ expansiort joints between or within the building foundations or structures.

)

Ifit is deurminad based on the results of the fragility analysis that more refined estimatas of_the differential displacements are: required, then two- or three dimannional soil-structure-interaction analyses can be performed using existing computer codes to develop more realistic estimates.

The displacements of the structures can occur in any direction, and thus the differential displacements can be transverse or parallel to the distance between the structures. The effects of simultaneous occurrence of two horizontal components of differential displacement or of the resulting forces or stresses imposed on pipe and ducts can be taken into account using any of the methods available for combining loads due to seismic excitation in multiple directions. .

De methods described above for estimating the differential displacements using the transient displacements given in Tables 4 and 5 account for motions of the structures relative to each other or of the structures relative to the adjacent ground. Buried conduits away from the effects of the structures are subject to strains due to the propagation of seismic motions across the site. Formulas to estimate these strains are presented in  ;

ASCE (1983,1984).

i

/

\

b l f 6.

SUMMARY

AND CONCLUSIONS The permanent and transient displacements and settlements due to an earthquake were  !

calculated versus the intensity of the ground motion. . The results were obtained using conservative methodology and previously existing information concerning the soils and stmetures at the site. The results were determined using the hazard curves and response j spectra from NUREG/CR-5250 and EPRI (1989) and seismic time histories provided by

Stevenson & Associates.

I j Based on the results of previous cross-hole testing by Weston Geophysical and the results

of calculations using available empirical correlations and the Pilgrim soils test data, two

~  !

' shear wave velocity profiles were determined for the outwash deposits. ' Displacements and se'ttlements"are provided for both of the' shear wave velocityiirdfilEIt is reasonable to expect that the two shear wave velocity profiles bound the true shear wave velocity l

\

profile at the Pilgrim site.

.' :q

! The results of the SHAKE analyses conducted for the purpose of calculating the displace-

] ments indicate that the largest realistic peak ground accelerations for soil profiles 1 and j 2 are less than 0.35 and 0.7 g, respectively. This conclusion is based on analyses using ,

1 the acceleration time history provided by Stevenson & Associates having a peak ground i

acceleration of 0.5 g (NR0098-1) scaled to peak ground surface accelerations of 0.1 to 0.9 g in the SHAKE analyses conducted for this report.

1

For soil profile 1 and a peak ground acceleration of 0.35 g, the settlements, permanent

horizontal displacements, and the_ transient horizontal _disalmts_are less _than O.29 inch,0.2 inch, and 1.05 inches, respectively. For soil' pro 61e 2 and a' peak ground acceleration of 0.7 g, the settlements, permanent horizontal displacements, and transient 4

horizontal displacements are less than 0.72 inch,1.7 inches, and 1.12 inches, respectively.

i This is consistent with the characterization of the soils at the site as very dense.

i A liquefaction stability failure is not possible at the Pilgrim site due to the dense state j of the in situ soils and compacted fill.

i 1

i 3

a i

4 i

REFERENCFE

1. ASCE, Committee on Seismic Analysis of the ASCE Structural Division Committee on Nuclear Structures and Materials (1983). " Seismic Response of Buried Pipes and Structural Components."

2. ASCE, Committee on Gas and Liquid Fuel Lifelines (1984). " Guidelines for the Seismic Design of Oil and Gas Pipeline Systems."

3. Bechtel (1976). Soils Repon prepared by Bechtel as part of Pilgrim 2 PSAR, dated August 31,1976, Amendment 26 (cantaina GEI soils data reports).

^ n+ , ms-s a

4. Castro, G. kl987). "On the Behavich of S' oils During Earthqskes - Liquefaction,"

Soit Dynamics and Liquefaction, A. S. r'abnat Editor, Elsevier.

h. .,

5. EPRI Report No. NP-6395-D (1989). "Probabilistic' Seismic Hazard Evaluations at Nuclear Power Plant Sites in the Central and Eastern United States: Resolution of the Charleston Earthquake Issue," April.

6. Geotechnical Engineers Inc. (1983). " Analysis of Groundwater Levels, Pilgrim Station Unit 1, Plymouth, Massachusetts," February 28.

7. Geotechnical Engineers Inc. (1978). Soil Borings - Location Plans, Logs, and Test Pits for Pilgrim Station, Plymouth, Massachusetts.

8. Gibbs, HJ. and Holtz, W.H. (195i)["Research on Deinia; the _ Density of Sand by Spoon Penetration Testing," 4th ISCMPE, LW Vol.1, p. 35.

l

9. Hardin, B.O. and Drnevich, V.P. (1972). " Shear Modulus in Soils: Design Equations and Curves," Journal of the Soil Mechanics and Foundations Division, l ASCE, Vol. 98, No. SM7, pp. 667-692.

10. National Research Council (1985). " Liquefaction of Soils during Earthquakes,"  !

National Academy Press, Washington, D.C. l

11. Nagase, H. and Ishihara, K. (1988). " Liquefaction Induced Compaction and Settlement of S,and During Earthquakes," Soils and Foundations, Vol. 28, No.1, pp. 65-76, March.

12. Newmark, N.M. (1%5). " Effects of Earthquakes on Dams and Embankments,"

Fifth Rankine Lecture, Geotechnique, Vol.15, No. 2.

4

l 1

. l I

13. Newmark, N.M.; Blume, J.A.; and Kapur K.K. (1973). " Seismic Design Spectra for Nuclear Power Plants," Journal of the Power Division, No. PO2, November, j pp. 287-303.

14. NUREG/CR-5250 (1989). " Seismic Hazard Characterization of 69 Nuclear Power Plant Sites East of the Rocky Mountains, Vol.1-8," January.

i

, 15. NUREG/CR-0098 (1978). " Development of Criteria for Seismic Review of l Selected Nuclear Power Plants," May.

16. Peck, R.B.; Hanson, W.E.; and 7hornburn, T.H. (1974). Foundation Engineering, John Wiley & Sons, Inc. ,

, s. . o au -- -

~

17. Poulos, S.J.; Castro, O.; ah France,'J.W. NkNMs Ekiluation '

Procedure," Journal of Geotechnical Engineering, ASCE, Vol. 3, No. 6, pp. 772-792. -

Ni ,

f $.

18. Schnabel, P.B.; Lysmer J.; and Seed, H.B. (1972). " SHAKE, A Computer Program for Earthquake Response Analysis of Horir= tally Layered Sites," Report No. EERC 72-12, University of California, Berkeley. n

19. Seed, H.B. and Idriss, I.M. (1971). "A Simplified Procedure for Evaluating Soil Liquefaction Potential," Journal of Soil Mechanics and Foundation Engineering, l ASCE, Vol. 97, No. 9, pp.1249-1273.  !

20. Seed, H.B. and Idriss, LM...(1970). " Soil Moduli and Damping Factors for Dynamic Response Analyses," Report No. 70-10,: University of California, Berkeley. ,

. 1.

21. Seed, H.B.; Arango, L; and Chan, C.K. (1975). " Evaluation of Soil Liquefaction Potential During Earthquakes," Report No. EERC 75-28, Earthquake Engineering Research Center, University of California, Berkeley.

22. Seed, H.B.; Idriss, I.M.; and Arango, I. (1983). " Evaluation of Liquefaction l

l Potential Using Field Performance Data," Journal of Geotechnical Engineering, ;

ASCE, Vol.109, No. GT3, pp. 458-482.

23. Sykora, D.W. (1987). " Examination of Existing Shear Wave Velocity and Shear .

Modulus Correlations in Soils," Waterways Experiment Station, U.S. Army Corps l of Engineers.

l

)

TABLE 1 - PERMANENT HORIZONTAL DISPLACEMENTS (INCHES) FOR SOIL PROFILE 2 Pilgrim 1 IPEEE, Plymouth, Massachusetts Location Yield Acceleration Peak Ground Surface Acceleration

'(9) 0.2 g 0.3 g ' O.4 g 0.5 g 0.7 g l 4 Condenser Tank, Fig. 3 0.34 0 0 0 0.1 0.6 Reactor Containment Building, Fig. 3 0.48 0 0 0 0 0.2 Power Block, Fig. 3 0.50 0 0 0 0 0.1 Intake Structure, Fig. 4 0.24 0 0 0.2 0.4 1.7 Yard, Fig. 4 0.34 0 0 0 0.1 0.6 Yard, Fig. 4 0.40 0 0'M 0 0 0.4 Diesel Generator Building, Fig. 5

{0.40 0 0i 0 0 0.4 l ,. v i

'l !l-U gel Consultants, Inc. u 9

1 TABLE 2A - SETTLEMENTS (INCHES) FOR SOIL PROFILE 1 Pilgrim 1 IPEEE, Plymouth, Massachusetts i

Depth Elevation Peak Ground Surface Acceleration (feet) (feet) 0.1 g 0.2 g 0.3 g 0.35 g 0 +22 0.02 0.07 0.17 0.29 10 +12 0.02 0.06 0.15 0.26  ;

20 - +2 g u 0.1.8

- M. ~07 ;T ]4l0.02 M0.09 25 -3 0 0.02 0.08 0.17 50 -28., .0 0.01 0.06 0.12 80 -58 I-0 9 0 0 0 TABLE 28 - DIFFERENTIAL SETTLEMENTS (INCHES) FOR PEAK GROUND SURFACE ACCELERATION OF 0.35 g Pilgrim 1 IPEEE, Plymouth, Massachusetts l

Ground Condenser Diesel Turbine Reactor intake Surface Tank Generator Buildng Buildng Structure Buildng Ground Surface, 0 0 0 0.12 0.17 0.17 El.22 Condenser Tank, 0 0 0 0.12 0.17 0.17 El.22 Diesel Generator 0 0 0 0.12 0.17 0.17 Bldg., El. 22 Turbine Bldg., 0.12 0.12 0.12 0 0.05 0.05 El.-5 Reactor Bldg., 0.17 0.17 0.17 0.05 0 0 El. 26 1

Intake Structure, 0.17 0.17 0.17 0.05 0 0 l El. -28 1

(

s Project 92012 gel Consultants, Inc. July 9,1992

e

)

TABLE 3A - SETTLEMENTS (INCHES) FOR SOIL PROFILE 2 l Pilgrim 1 IPEEE, Plymouth, Massachusetts i Depth Elevation Peak Ground Surface Acceleration I (feet) (feet) 0.1 g 0.2 g 0.3 g 0.4 g 0.5 g 0.6g 0.7 g 0 +22 0.02 0.05 0.10 0.17 0.27 0.45 0.72 10 +12 0.01 0.04 0.07 0.13 0.22 0.38 0.63 20 ~ ,~ 01 ~ 1, 0 0.03 0.07 f.+2.~ l " 0.01. 0.14 0.27 25 -3 0 0~ 0.01 .0.02 0.05 0.11 0.22 50 _..-28 0 .0. 0 0 0 0 0.01  !

80 - -58 0 0 0 0 -0 0 0 TABLE 38 - DIFFERENTIAL SETTLEMENTS (INCHES) FOR PEAK GROUND SURFACE ACCELERATION OF 0.7 g Pilgrim 1 IPEEE, Plymouth, Massachusetts Ground Condenser Diesel Turbine Reactor intake Surface Tank Generator Bulldng Bulking Structure BulkSng Ground Surface, 0 0 0 0.50 0.71 0.71 El.22 Condenser Tank, 0 0 0 0.50 0.71 0.71 El.22 Diesel Generator 0 0 0 0.50 0.71 0.71 Bldg., El. 22 Turbine Bldg., 0.50 0.50 0.50 0 0.21 0.21 El.-5 Reactor Bldg., 0.71 0.71 0.71 0.21 0 0 El. -26 Intake Structure, 0.71 0.71 0.71 0.21 0 0 El. -28 Project 92012 gel Consultants, Inc. July 9,1992

/

l TABLE 4 - TRANSIENT HORIZONTAL DISPLACEMENTS (INCHES)

FOR SOIL PROFILE 1 ,

Pilgrim 1 IPEEE, Plymouth, Massachusetts Depth Elevation Peak Ground Surface Acceleration (feet) (feet) 0.1 g 0.2 g 0.3 g 0.35 g 0 +22 0.15 0.37 0.74 1.05

--10 ..

- - +12 - , , .

0.14 0.36 ..20.7 1.03 18 +4 0.13 0.34 0.68 0.98 25 -3 0.12 0.32 0.64 0.93 50 -28 0.07 0.19 OS0 0.58 80 -58 0 0 0 0 Note: The building displacement can be assumed to be equal to the ground movements at the elevation of the foundation base, i.e., at the elevations given below.

Building Foundation Base Elevation

. - - . - . . - - - . . . .L. . -- . - . . . _ _ . .

Ground Surface +22 Condenser Tank +22 Diesel Generator Bldg. +22 l.

Turbine Bldg. -5 l l

Reactor Bldg.

-26 i Intake Structure -28 l

l f

l 2

Project 92012 gel Consultants, Inc. July 9,1992 l

TABLE 5 - TRANSIENT HORIZONTAL DISPLACEMENTS (INCHES)

FOR SOIL PROFILE 2 Pilgrim 1 IPEEE, Plymouth, Massachusetts l

Depth Elevation Peak Ground Surface Acceleration (feet) (feet) 0.1 g 02 g 0.3 g 0.4 g 0.5 g 0.6 g 0.7 g 0 +22 0.06 0.14 024 0.38 0.56 0.80 1.12 10 +12 0.05 0.12 022 0.35 0.52 ~ 0.75 1.06 y

18 +4 0.04 0.10 0.18 0.30 0.43 0.62 0.87 25 -3 0.04 0.08 0.14 023 0.33 0.46 0.65 50 -28 0.01 0.03 0.04 0.06 0.07 ' '~ 0.09 0.i3 80 -58 0 0 0 0 0 0 0 i

Note: The building displacement can be assumed to be equal to the ground movements at the j elevation of the foundation base, i.e., at the elevations given below.

Building Foundation Base Elevation ' J~

Ground Surface +22 Condenser Tank +22 Diesel Generator Bldg. +22 Turbine Bldg. -5 '

Reactor Bldg. -26 Intake Structure -28 Project 92012 gel Consultants, Inc. July 9,1992

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NOTES l ~

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Map la taken from U.S.G.S. Topographic 7.5 Minute Series

• Map of Monomet, Mass. Quadrangle,1977. - Sc=

Datum is National Geodetic Vertical Datum (NGVD).

Contour Interval is 10 Feet. l Stevenson & Associates Pilgn.m 1 IPEEE Wobum, Massochusetts SITE LOCATION MAP Plymouth, Massachusetts j OEI Consultants, Inc. Project 92012 May 1992 Fig. 1