ML20209A397
| ML20209A397 | |
| Person / Time | |
|---|---|
| Site: | Millstone  |
| Issue date: | 06/22/2020 |
| From: | Dominion Energy Nuclear Connecticut |
| To: | Office of Nuclear Reactor Regulation |
| Shared Package | |
| ML20209A356 | List:
|
| References | |
| 20-223 | |
| Download: ML20209A397 (201) | |
Text
Millstone Power Station Unit 3 Safety Analysis Report Chapter 2: Site Characteristics
Table of Contents tion Title Page GEOGRAPHY AND DEMOGRAPHY ..................................................... 2.1-1 1 Site Location and Description..................................................................... 2.1-1 1.1 Specification of Location............................................................................ 2.1-1 1.2 Site Area ..................................................................................................... 2.1-1 1.3 Boundaries for Establishing Effluent Release Limits................................. 2.1-1 2 Exclusion Area Authority and Control ....................................................... 2.1-2 2.1 Authority ..................................................................................................... 2.1-2 2.2 Control of Activities Unrelated to Plant Operation .................................... 2.1-3 2.3 Arrangements for Traffic Control............................................................... 2.1-3 2.4 Abandonment or Relocation of Roads........................................................ 2.1-3 2.5 Independent Spent Fuel Storage Installation (ISFSI) ................................. 2.1-4 3 Population Distribution............................................................................... 2.1-4 3.1 Population Distribution within 10 miles ..................................................... 2.1-4 3.2 Population Distribution within 50 Miles .................................................... 2.1-5 3.3 Transient Population ................................................................................... 2.1-6 3.4 Low Population Zone.................................................................................. 2.1-6 3.5 Population Center ....................................................................................... 2.1-6 3.6 Population Density...................................................................................... 2.1-7 4 References For Section 2.1 ......................................................................... 2.1-7 NEARBY INDUSTRIAL, TRANSPORTATION, AND MILITARY FACILITIES........................................................................... 2.2-1 1 Locations and Routes.................................................................................. 2.2-1 2 Descriptions ................................................................................................ 2.2-1 2.1 Description of Facilities.............................................................................. 2.2-1 2.2 Description of Products and Materials........................................................ 2.2-3 2.3 Pipelines...................................................................................................... 2.2-5 2.4 Waterways .................................................................................................. 2.2-5 2.5 Airports ....................................................................................................... 2.2-5 2.6 Highways .................................................................................................... 2.2-6
tion Title Page 2.7 Railroads ..................................................................................................... 2.2-7 2.8 Projections of Industrial Growth................................................................. 2.2-7 3 Evaluation of Potential Accidents............................................................... 2.2-8 3.1 Determination of Design Basis Events .................................................... 2.2-12 3.1.1 Missiles Generated by Events near the Millstone Site ............................. 2.2-12 3.1.2 Unconfined Vapor Cloud Explosion Hazard ............................................ 2.2-19 3.1.3 .................................................................................................................. 2.2-19 3.1.4 Hydrogen Storage at the Site .................................................................... 2.2-22 3.1.5 Toxic Chemicals ....................................................................................... 2.2-22 3.2 Effects of Design Basis Events ................................................................. 2.2-24 4 References for Section 2.2 ........................................................................ 2.2-25 METEOROLOGY ...................................................................................... 2.3-1 1 Regional Climatology ................................................................................. 2.3-1 1.1 General Climate .......................................................................................... 2.3-1 1.1.1 Air Masses and Synoptic Features.............................................................. 2.3-1 1.1.2 Temperature, Humidity, and Precipitation ................................................. 2.3-2 1.1.3 Prevailing Winds......................................................................................... 2.3-2 1.1.4 Relationships of Synoptic to Local Conditions .......................................... 2.3-3 1.2 Regional Meteorological Conditions for Design and Operating Bases ...... 2.3-3 1.2.1 Strong Winds .............................................................................................. 2.3-3 1.2.2 Thunderstorms and Lightning..................................................................... 2.3-4 1.2.3 Hurricanes ................................................................................................... 2.3-4 1.2.4 Tornadoes and Waterspouts........................................................................ 2.3-4 1.2.5 Extremes of Precipitation............................................................................ 2.3-5 1.2.6 Extremes of Snowfall.................................................................................. 2.3-5 1.2.7 Hailstorms ................................................................................................... 2.3-5 1.2.8 Freezing Rain, Glaze, and Rime ................................................................. 2.3-6 1.2.9 Fog And Ice Fog ......................................................................................... 2.3-6 1.2.10 High Air Pollution Potential ....................................................................... 2.3-6 1.2.11 Meteorological Effects on Ultimate Heat Sink........................................... 2.3-6
tion Title Page 2 Local Meteorology...................................................................................... 2.3-7 2.1 Normal and Extreme Values of Meteorological Parameters ...................... 2.3-7 2.1.1 Wind Conditions ......................................................................................... 2.3-7 2.1.2 Air Temperature and Water Vapor ............................................................. 2.3-7 2.1.3 Precipitation ................................................................................................ 2.3-8 2.1.4 Fog and Smog ............................................................................................. 2.3-8 2.1.5 Atmospheric Stability ................................................................................. 2.3-8 2.1.6 Seasonal and Annual Mixing Heights ........................................................ 2.3-9 2.2 Potential Influence of the Plant and Its Facilities on Local Meteorology .. 2.3-9 2.3 Local Meteorological Conditions for Design and Operating Bases ........... 2.3-9 2.3.1 Design Basis Tornado ................................................................................. 2.3-9 2.3.2 Design Basis Hurricane ............................................................................ 2.3-10 2.3.3 Snow Load ................................................................................................ 2.3-10 2.3.4 Rainfall...................................................................................................... 2.3-10 2.3.5 Adverse Diffusion Conditions .................................................................. 2.3-10 2.4 Topography ............................................................................................... 2.3-10 3 On-Site Meteorological Measurements Program ..................................... 2.3-11 3.1 Measurement Locations and Elevations ................................................... 2.3-11 3.2 Meteorological Instrumentation................................................................ 2.3-11 3.3 Data Recording Systems and Data Processing ......................................... 2.3-12 3.4 Quality Assurance for Meteorological System and Data.......................... 2.3-12 3.5 Data Analysis ............................................................................................ 2.3-13 4 Short-Term (Accident) Diffusion Estimates............................................. 2.3-13 4.1 Objective ................................................................................................... 2.3-13 4.2 Calculation ................................................................................................ 2.3-13 4.3 Results....................................................................................................... 2.3-13 5 Long Term (Routine) Diffusion Estimates ............................................... 2.3-14 5.1 Calculation Objective ............................................................................... 2.3-14 5.2 Calculations .............................................................................................. 2.3-14 5.2.1 Release Points and Receptor Locations .................................................... 2.3-14 5.2.2 Database.................................................................................................... 2.3-14
tion Title Page 5.2.3 Models ...................................................................................................... 2.3-14 6 References for Section 2.3 ........................................................................ 2.3-14 HYDROLOGIC ENGINEERING ............................................................. 2.4-1 1 Hydrologic Description............................................................................... 2.4-1 1.1 Site and Facilities........................................................................................ 2.4-1 1.2 Hydrosphere................................................................................................ 2.4-1 2 Floods.......................................................................................................... 2.4-2 2.1 Flood History .............................................................................................. 2.4-2 2.2 Flood Design Considerations...................................................................... 2.4-3 2.3 Effect of Local Intense Precipitation .......................................................... 2.4-4 3 Probable Maximum Flood on Streams and Rivers ..................................... 2.4-7 4 Potential Dam Failures, Seismically Induced ............................................. 2.4-7 5 Probable Maximum Surge and Seiche Flooding ........................................ 2.4-7 5.1 Probable Maximum Winds and Associated Meteorological Parameters.... 2.4-7 5.2 Surge and Seiche Water Levels .................................................................. 2.4-8 5.3 Wave Action ............................................................................................. 2.4-10 5.3.1 Deep Water Waves ................................................................................... 2.4-10 5.3.2 Shallow Water Waves............................................................................... 2.4-12 5.3.3 Wave Shoaling .......................................................................................... 2.4-13 5.3.4 Wave Refraction ....................................................................................... 2.4-13 5.3.5 Wave Runup ............................................................................................. 2.4-14 5.3.6 Clapotis on Intake Structure ..................................................................... 2.4-14 5.4 Resonance ................................................................................................. 2.4-14 5.5 Protective Structures ................................................................................. 2.4-15 6 Probable Maximum Tsunami Flooding .................................................... 2.4-15 7 Ice Effects ................................................................................................. 2.4-16 8 Cooling Water Canals and Reservoirs ...................................................... 2.4-16 9 Channel Diversions................................................................................... 2.4-16 10 Flooding Protection Requirements ........................................................... 2.4-16 11 Low Water Considerations ....................................................................... 2.4-16
tion Title Page 11.1 Low Flow in Rivers and Streams.............................................................. 2.4-16 11.2 Low Water Resulting from Surges, Seiches, or Tsunamis ....................... 2.4-17 11.3 Historical Low Water................................................................................ 2.4-18 11.4 Future Control........................................................................................... 2.4-18 11.5 Plant Requirements ................................................................................... 2.4-18 11.6 Heat Sink Dependability Requirements.................................................... 2.4-18 11.7 Dispersion, Dilution, and Travel Times of Accidental Releases of Liquid Effluents in Surface Waters ...................................................................... 2.4-19 12 Groundwater ............................................................................................. 2.4-23 12.1 Description and Onsite Use ...................................................................... 2.4-23 12.2 Sources...................................................................................................... 2.4-23 12.3 Accident Effects........................................................................................ 2.4-24 12.4 Monitoring or Safeguard Requirements ................................................... 2.4-28 12.5 Design Bases for Subsurface Hydrostatic Loading .................................. 2.4-28 13 Technical Specification and Emergency Operation Requirements .......... 2.4-28 14 References for Section 2.4 ........................................................................ 2.4-29 GEOLOGY, SEISMOLOGY, AND GEOTECHNICAL ENGINEERING ......................................................................................... 2.5-1 1 BASIC GEOLOGICAL AND SEISMIC INFORMATION ................... 2.5.1-1 1.1 Regional Geology .................................................................................... 2.5.1-1 1.1.1 Regional Physiography and Geomorphology .......................................... 2.5.1-2 1.1.2 Regional Structure ................................................................................... 2.5.1-3 1.1.3 Regional Stratigraphy .............................................................................. 2.5.1-6 1.1.4 Regional Tectonics .................................................................................. 2.5.1-6 1.1.4.1 Domes and Basins.................................................................................... 2.5.1-6 1.1.4.2 Faulting .................................................................................................... 2.5.1-7 1.1.4.3 Tectonic Summary ................................................................................. 2.5.1-10 1.1.4.4 Remote Sensing ..................................................................................... 2.5.1-10 1.1.4.5 Structural Significance of Geophysical Studies..................................... 2.5.1-11 1.1.5 Regional Geologic History .................................................................... 2.5.1-12
tion Title Page 1.2 Site Geology .......................................................................................... 2.5.1-16 1.2.1 Site Physiography .................................................................................. 2.5.1-16 1.2.2 Local Stratigraphy.................................................................................. 2.5.1-17 1.2.3 Site Stratigraphy .................................................................................... 2.5.1-17 1.2.4 Local Structural Geology....................................................................... 2.5.1-18 1.2.4.1 Site Structural Geology.......................................................................... 2.5.1-20 1.2.5 Site Geological History.......................................................................... 2.5.1-21 1.2.6 Site Engineering Geology ...................................................................... 2.5.1-24 1.3 References for Section 2.5.1 .................................................................. 2.5.1-25 2 VIBRATORY GROUND MOTION ....................................................... 2.5.2-1 2.1 Seismicity................................................................................................. 2.5.2-1 2.1.1 Completeness and Reliability of Earthquake Cataloging ........................ 2.5.2-1 2.1.2 Earthquake History .................................................................................. 2.5.2-2 2.1.3 Seismicity within 50 Miles of the Site..................................................... 2.5.2-4 2.1.4 Earthquakes Felt at the Site ..................................................................... 2.5.2-5 2.2 Geologic Structures and Tectonic Activity.............................................. 2.5.2-9 2.3 Correlation of Earthquake Activity with Geologic Structures or Tectonic Prov-inces ....................................................................................................... 2.5.2-14 2.3.1 Correlation with Geologic Structures .................................................... 2.5.2-14 2.3.2 Correlation with Tectonic Provinces ..................................................... 2.5.2-15 2.4 Maximum Earthquake Potential ............................................................ 2.5.2-16 2.4.1 Maximum Historical Site Intensity........................................................ 2.5.2-16 2.4.2 Maximum Earthquake Potential from Tectonic Province Approach..... 2.5.2-17 2.5 Seismic Wave Transmission Characteristics of the Site........................ 2.5.2-17 2.6 Safe Shutdown Earthquake .................................................................... 2.5.2-18 2.7 Operating Basis Earthquake................................................................... 2.5.2-18 2.8 References for Section 2.5.2 .................................................................. 2.5.2-18 3 SURFACE FAULTING .......................................................................... 2.5.3-1 3.1 Geologic Conditions of the Site............................................................... 2.5.3-1
tion Title Page 3.2 Evidence of Fault Offset .......................................................................... 2.5.3-1 3.2.1 Petrographic Analysis .............................................................................. 2.5.3-3 3.2.2 Clay Mineralogy, Fluid Inclusion Analysis, and Radiometric Dating .... 2.5.3-4 3.2.3 Conclusions.............................................................................................. 2.5.3-7 3.3 Earthquakes Associated with Capable Faults .......................................... 2.5.3-8 3.4 Investigation of Capable Faults ............................................................... 2.5.3-8 3.5 Correlation of Epicenters with Capable Faults ........................................ 2.5.3-8 3.6 Description of Capable Faults.................................................................. 2.5.3-8 3.7 Zone Requiring Detailed Faulting Investigation ..................................... 2.5.3-8 3.8 Results of Faulting Investigation ............................................................. 2.5.3-8 3.9 References for Section 2.5.3 .................................................................... 2.5.3-8 4 STABILITY OF SUBSURFACE MATERIALS AND FOUNDATIONS ..................................................................................... 2.5.4-1 4.1 Geologic Features .................................................................................... 2.5.4-1 4.2 Properties of Subsurface Materials .......................................................... 2.5.4-2 4.2.1 Artificial Fill ............................................................................................ 2.5.4-3 4.2.2 Beach Deposits ........................................................................................ 2.5.4-3 4.2.3 Unclassified Stream Deposits .................................................................. 2.5.4-3 4.2.4 Ablation Till............................................................................................. 2.5.4-4 4.2.5 Basal Till.................................................................................................. 2.5.4-4 4.2.6 Monson Gneiss ........................................................................................ 2.5.4-5 4.3 Exploration............................................................................................... 2.5.4-6 4.4 Geophysical Surveys................................................................................ 2.5.4-6 4.4.1 Onshore Seismic Refraction Survey ........................................................ 2.5.4-7 4.4.2 Offshore Seismic and Bathymetric Survey.............................................. 2.5.4-7 4.4.3 Seismic Velocity Measurements.............................................................. 2.5.4-7 4.5 Excavations and Backfill ......................................................................... 2.5.4-9 4.5.1 Excavation ............................................................................................... 2.5.4-9 4.5.2 Backfill................................................................................................... 2.5.4-11 4.5.3 Extent of Dredging................................................................................. 2.5.4-14
tion Title Page 4.6 Groundwater Conditions........................................................................ 2.5.4-14 4.6.1 Design Basis for Groundwater............................................................... 2.5.4-14 4.6.2 Groundwater Conditions During Construction...................................... 2.5.4-16 4.7 Response of Soil and Rock to Dynamic Loading .................................. 2.5.4-17 4.7.1 Subsurface Material Properties Used in SSI Analysis........................... 2.5.4-18 4.8 Liquefaction Potential............................................................................ 2.5.4-19 4.8.1 Structural Backfill.................................................................................. 2.5.4-19 4.8.2 Basal Tills .............................................................................................. 2.5.4-19 4.8.3 Beach and Glacial Outwash Sands ........................................................ 2.5.4-20 4.8.3.1 Dynamic Response Analysis of Beach and Glacial Outwash Sands ...................................................................................................... 2.5.4-20 4.8.3.2 Liquefaction Analysis of Beach and Glacial Outwash Sands................ 2.5.4-22 4.8.3.3 Liquefaction Analyses of Beach Area Sands using 2-Dimensional Dynamic Response Analysis......................................... 2.5.4-24 4.8.4 Ablation Till........................................................................................... 2.5.4-26 4.8.4.1 Dynamic Response Analysis of Ablation Till ....................................... 2.5.4-26 4.8.4.2 Liquefaction Analysis of Ablation Till .................................................. 2.5.4-27 4.9 Earthquake Design Basis ....................................................................... 2.5.4-28 4.10 Static Stability........................................................................................ 2.5.4-28 4.10.1 Bearing Capacity.................................................................................... 2.5.4-28 4.10.2 Settlement of Structures......................................................................... 2.5.4-29 4.10.3 Lateral Earth Pressures .......................................................................... 2.5.4-30 4.11 Design Criteria ....................................................................................... 2.5.4-30 4.12 Techniques to Improve Subsurface Conditions ..................................... 2.5.4-31 4.13 Structure Settlement............................................................................... 2.5.4-32 4.14 Construction Notes ................................................................................ 2.5.4-32 4.15 References for Section 2.5.4 .................................................................. 2.5.4-33 5 STABILITY OF SLOPES ....................................................................... 2.5.5-1 5.1 Slope Characteristics................................................................................ 2.5.5-1 5.1.1 Shoreline Slope ........................................................................................ 2.5.5-1
tion Title Page 5.1.2 Containment Rock Cut............................................................................. 2.5.5-3 5.2 Design Criteria and Analysis ................................................................... 2.5.5-3 5.2.1 Shoreline Slope ........................................................................................ 2.5.5-3 5.2.2 Containment Rock Cut............................................................................. 2.5.5-6 5.3 Logs of Borings ....................................................................................... 2.5.5-7 5.4 Compacted Fill......................................................................................... 2.5.5-7 5.5 References for Section 2.5.5 .................................................................... 2.5.5-7 6 EMBANKMENTS AND DAMS ............................................................ 2.5.6-1 PENDIX 2.5A- AGE OF TILL AT MILLSTONE POINT .................................... 2.5A-1 PENDIX 2.5B- PETROGRAPHIC REPORTS, FINAL GRADE ..........................2.5B-1 PENDIX 2.5C- MINERALOGICAL ANALYSIS OF MILLSTONE FAULT GOUGE SAMPLES......................................................................................2.5C-1 PENDIX 2.5D- POTASSIUM - ARGON AGE DETERMINATION ................... 2.5D-1 PENDIX 2.5E- SEPARATION OF < 2 FRACTION AND CLAY ANALYSIS OF SAMPLES B, C, D (ESF BUILDING) AND P-1 AND P-2 (PUMPHOUSE).............................................................................2.5E-1 PENDIX 2.5F- DYNAMIC SOIL TESTING ON BEACH SANDS...................... 2.5F-1 PENDIX 2.5G- CONSOLIDATED UNDRAINED TESTS ON BEACH SANDS2.5G-1 PENDIX 2.5H- SEISMIC VELOCITY MEASUREMENTS ................................ 2.5H-1 PENDIX 2.5I- DIRECT SHEAR TESTS ON NATURAL ROCK JOINTS .......... 2.5I-1 PENDIX 2.5J- BORING LOGS..............................................................................2.5J-1 PENDIX 2.5K- SEISMIC SURVEY...................................................................... 2.5K-1 PENDIX 2.5L- SEISMIC AND BATHYMETRIC SURVEY ...............................2.5L-1 PENDIX 2.5M- LABORATORY TEST PROGRAM FOR PROPOSED ADDITIONAL STRUCTURAL BACKFILL SOURCES.....................................2.5M-1
List of Tables mber Title 1 1990 Population and Population Densities Cities and Towns Within 10 Miles of Millstone 2 Population Growth 1960-1990 3 Population Distribution 1985 (0-20 km) 4 Population Distribution Within 10 Miles of Millstone - 1990 Census 5 Population Distribution Within 10 Miles of Millstone - 2000 Projected 6 Population Distribution Within 10 Miles of Millstone - 2010 Projected 7 Population Distribution Within 10 Miles of Millstone - 2020 Projected 8 Population Distribution Within 10 Miles of Millstone - 2030 Projected 9 Population Distribution 1985 (0-80 km) 10 Population Distribution Within 50 Miles of Millstone - 1990 Census 11 Population Distribution Within 50 Miles of Millstone - 2000 Projected 12 Population Distribution Within 50 Miles of Millstone - 2010 Projected 13 Population Distribution Within 50 Miles of Millstone - 2020 Projected 14 Population Distribution Within 50 Miles of Millstone - 2030 Projected 15 Transient Population Within 10 Miles of Millstone - 1991-1992 School Enrollment 16 Transient Population Within 10 Miles of Millstone (Employment) 17 Transient Population Within 10 Miles of Millstone State Parks and Forests (With Documented Attendance) 18 Low Population Zone Permanent Population Distributions 19 Low Population Zone School Enrollment and Employment 20 Metropolitan Areas Within 50 Miles of Millstone 1990 Census Population 21 Population Centers Within 50 Miles of Millstone 22 Population Density* 1985 (0-20 km) 23 Population Density 1985 (0-80 km) 24 Population Density Within 10 Miles of Millstone 1990 (People per Square Mile) 25 Population Density Within 10 Miles of Millstone 2030 (People per Square Mile)
mber Title 26 Population Density Within 50 Miles of Millstone 1990 (People per Square Mile) 27 Population Density Within 50 Miles of Millstone 2030 (People per Square Mile) 28 Cumulative Population Density 1985 29 Cumulative Population Density Within 50 Miles of Millstone 1990 (People per Square Mile) 30 Cumulative population Density Within 50 Miles of Millstone 2030 (People per Square Mile) 1 Description of Facilities 2 List of Hazardous Materials Potentially Capable of Producing Significant Missiles 3 Summary of Exposure Distance Calculation 4 Aggregate Probability of Explosion or Violent Rupture Capable of Missile Generation 5 Types of Tank Car Missiles 6 Tank Car Fragment Range (Feet) at 10-Degree Launch Angle 7 Estimated Ignition Probabilities 8 Probability of an Unconfined Vapor Cloud Explosion 1 Monthly, Seasonal, and Annual averages and Extremes of Temperature at Bridgeport, Conn. (1901-1981) 2 Mean Number of Days with Selected Temperature Conditions at Bridgeport, Conn.
(1966-1981) 3 Monthly, Seasonal, and Annual Averages and Extremes of Relative Humidity at Bridgeport, Conn. (1949-1981) 4 Monthly, Seasonal, and Annual Frequency Distributions of Wind Direction at Bridgeport, Conn. (1949-1980)
-5 Occurrence of Bridgeport Wind Persistence Episodes within the same 22.5-Degree Sector (1949-1965) 6 Monthly, Seasonal, and Annual Frequency Distributions of Wind Direction at Bridgeport, Conn. (1949-1980) 7 Monthly, Seasonal, and Annual Wind Speed Extremes at Bridgeport, Conn. (1961-1990)
mber Title 8 Mean Number of Days of Thunderstorm Occurrence at Bridgeport, Conn. (1951-1981) 9 Monthly, Seasonal, and Annual Averages and Extremes of Precipitation at Bridgeport, Conn. (1901-June 1982) 10 Estimated Precipitation Extremes for Periods up to 24 Hours and Recurrence Intervals Up to 100 Years 11 Monthly, Seasonal, and Annual Averages and Extremes of Snowfall at Bridgeport, Conn. (1893-June 1990) 12 Monthly, Seasonal, and Annual Averages of Freezing Rain and Drizzle at Bridgeport, Conn. (1949-1980) 13 Average Monthly, Seasonal, and Annual Hours and Frequencies (percent) of Various Fog Conditions (1949-1980) at Bridgeport, Connecticut 14 Monthly and Annual Wind Direction and Speed Distributions for Surface Winds, at Bridgeport, Conn. (1949-1980) 15 Monthly and Annual Wind Direction and Speed Distributions for 33-Foot Winds at Millstone (1974-1981) 16 Comparison of Wind Direction Frequency Distribution by Quadrant at Bridgeport, Conn. and Millstone 17 Comparison of Average Wind Speed by Quadrant at Bridgeport, Conn. and Millstone 18 Occurrence of Wind Persistence Episodes Within the Same 22.5-Degree Sector at Millstone (1974-1981)
-19 Millstone Climatological Summary (1974-2000) 20 Comparison of Monthly and Annual Average Dry-Bulb and Dewpoint Temperature Averages at Bridgeport, Conn. and Millstone 21 Comparison of Monthly and Annual Average Relative Humidity Averages at Bridgeport and Millstone
-22 Mean Number of Days with Heavy Fog at Bridgeport, Conn. and Block Island, Rhode Island (1951-1981) 23 Wind Direction/Stability Class/Visibility Joint Frequency Distribution at Millstone 24 Persistence of Poor Visibility ( 1 Mile) Conditions at Millstone (Hours) (1974-1981)
mber Title 25 Bridgeport Pasquill Stability Class Distribution (1949-1980) 26 Millstone Stability Class Distribution Using Delta-T for Stability Determination 27 Millstone Stability Class Distribution Using Sigma Theta for Stability Determination 28 Comparison of Pasquill Stability Class Distribution at Bridgeport, Conn. and Millstone 29 Persistence of Stable Conditions (E, F, and G Stabilities) at Millstone (1974-1981) 30 Seasonal and Annual Atmospheric Mixing Depths at Millstone 31 On-site Meteorological Tower Measurements 32 Millstone Meteorological Tower Instrumentation 33 Monthly Summary of Data Recovery Rates/Meteorological System 34 Distances from Release Points to Receptors 1 Connecticut Public Water Supplies within 20 Miles of Millstone 3 2 Maximum Wave Heights Generated by Slow, Medium, and High Speed Storms (Deep-Water Fetch) 3 Maximum Shallow Water Waves (after Refraction) Slow Speed Probable Maximum Hurricane 4 Maximum Shallow Water Waves (after Refraction) Medium Speed Probable Maximum Hurricane 5 Maximum Shallow Water Waves (after Refraction) High Speed Probable Maximum Hurricane 6 Lowest Tides at New London, Connecticut 1938-1974 7 Circulating Water System and Service Water System Heat Loads 8 Dilution Factors and Travel Time
- 9 Category I Structures - Roof Survey 10 Input Data to Program HEC-2 Water Surface Computations 11 Computed Water Surface Elevations at Safety-Related Structures 12 Roof Area and Ponding Level Due to PMP (1)Category I Structures 13 Overflow Length of the Parapet Wall on the Roof Used in PMP Analysis - Category I Structures
mber Title 1-1 Rock Formations of the Coastal Plain off Southern New England 1-2 Rock Formations of Western Connecticut 1-3 Rock Formations of Eastern Connecticut and Western Rhode Island 1-4 Rock Formations of Central Rhode Island (and not Included in Previous Descriptions) 1-5 Rock Formations in Northern and Eastern Rhode Island and Southern Massachusetts 1-6 Rock Formations of Central Massachusetts 1-7 East of Clinton-Newbury Fault System, Eastern Massachusetts, and New Hampshire 1-8 Descriptions of Lineaments from LANDSAT Photographs (Shown on Figure 2.5.1-10) 2-1 Modified Mercalli (MM) Intensity Scale of 1931 2-2 List of Operating Seismic Stations 2-3 Chronological Catalog of Earthquake Activity within 200 Miles of the Site 2-4 List of Earthquakes within the 50-Mile Radius 3-1 List of Faults 3-2 List of Samples 3-3 List of K/Ar Age Determinations of Fault Gouge 4-1 List of Joints - Final Grade Floors of Structures 4-2 List of Foliations - Final Grade Floors of Structures 4-3 List of Slickensides - Final Grade Floors of Structures 4-4 List of Joints - Final Grade Containment and Engineered Safety Features Building Walls 4-5 List of Foliations - Final Grade Containment and Engineered Safety Features Building Walls 4-6 List of Slickensides - Final Grade Containment and Engineered Safety Features Building Walls 4-7 List of Joints - Final Grade Walls of Structures 4-8 List of Foliations - Final Grade Walls of Structures
mber Title 4-9 List of Slickensides - Final Grade Walls of Structures 4-10 Rock Compression Test Results 4-11 Direct Shear Test Results From Joint and Foliating Surfaces 4-12 Summary of Static Soil Properties for Beach Sands
- 4-13 Natural Water Contents of Split Spoon Samples 4-14 Foundation Data for Major Structures 4-15 List of Approximate Boring Locations, Ground Elevations, and Groundwater Elevations
- 4-16 Summary of Water Pressure Test Data 4-17 Groundwater Observations 4-18 Factors of Safety Against Liquefaction of Beach Sands 4-19 In-Place Density Test Results on Category I Structural Backfill Beneath the Service Water Intake Pipe Encasement 4-20 In-Place Density Test Results at Control and Emergency Generator Enclosure Buildings 4-23 Emergency Generator Enclosure - Soil Properties with Structure Effects from SHAKE Analysis 4-24 Bearing Capacity of Major Structures 4-25 Results of Two-Dimensional Liquefaction Analysis of Beach Area Sands
List of Figures mber Title 1 General Site Location
-2 General Vicinity 3 Site Layout 4 Site Plan 5 Towns Within 10 Miles 6 1985 Population Distribution 0-20 km 7 Population Sectors for 0-10 Miles 8 Counties within 50 Miles 9 1985 Population Distribution 0-80 km 10 Population Sectors for 0-50 miles 11 Roads and Facilities in the LPZ 12 LPZ Population Sectors Distribution 1 Major Industrial, Transportation and Military Facilities 2 Instrument Landing Patterns at Trumbull Airport 3 Air Lanes Adjacent to Millstone Point 4 New London County-State Highways and Town Roads
-5 Propane Concentration Outside and Inside the Control Room 1 Topography in the Vicinity of Millstone Point 2 Topographical Profiles within 5 Miles of Site 3 Topographical Profiles within 5 Miles of Site 4 Topographical Profiles within 50 Miles of Site (Sheet 1) 5 Topographical Profiles within 50 Miles of Site (Sheet 1) 6 General Topography - 50 Miles (Sheet 1) 7 Meteorological Instrument and Data Quality Assurance Flow Diagram 1 Facilities Located on the Site 2 Public Water Supplies within 20 Miles of Site
List of Figures (Continued) mber Title 3 Locations of Hydrographic Field Survey Stations, June to October 1965 4 Tidal Currents Measured by Essex Marine Laboratory 5 Bottom Profiles Established by Essex Marine Laboratory 6 Frequency of Tidal Flooding at New London, Connecticut 7 Site Grade and Drainage Basins for PMP Runoff Analysis 8 Bottom Profile Along Path of Maximum Surface Winds 9 Coincident Wave and Surge Slow-Speed Probable Maximum Hurricane 10 Coincident Wave and Surge Medium-Speed Probable Maximum Hurricane 11 Coincident Wave and Surge High-Speed Probable Maximum Hurricane 12 Locus of Hurricane Eye, Hurricane Type: Large Radius, Slow Speed of Translation 13 Locus of Hurricane Eye, Hurricane Type: Large Radius, Medium Speed of Translation 14 Locus of Hurricane Eye, Hurricane Type: Large Radius, High Speed of Translation 15 Wave Transects on Long Island Sound 16 Areas Under Effect of Wave Shoaling and Wave Refraction 17 Wave Refraction Diagram, Block Island Sound Grid 18 Wave Refraction Diagram, Millstone Grid, Angle of Approach South 30 Degrees East 19 Wave Refraction Diagram, Millstone Grid, Angle of Approach West 85 Degrees South 20 Wave Refraction Diagram, Millstone Grid, Angle of Approach West 45 Degrees South 21 Wave Refraction Diagram, Millstone Grid, Angle of Approach West 17 Degrees South 22 Topography and Runup Transects, Millstone Location 23 Intake Transect A
-24 Runup Transect B (West) 25 Runup Transect C (East) 26 Wave Clapotis at Intake
List of Figures (Continued) mber Title 27 Inputs to One Dimensional Setdown Model 28 Large Radius, Probable Maximum Hurricane Isovel Field 29 Large Radius, Slow Speed of Translation Time Variant Wind Field - Millstone 30 Setdown Versus Wind Speed 31 Boundary of the Modeled Area 32 Onsite Well Locations 33 Probable Seepage Path From Boron Recovery Tank and Waste Disposal Building to Long Island Sound 34 Scupper Details - Control, Hydrogen Recombiner, and Containment Enclosure Buildings 35 Roof Plug Sealing Detail - Hydrogen Recombiner Building 36 Hatch Cover Details - Circulating Water Pumphouse Service Water Pump Cubicle 37 Hatch Cover Details - Control Building Mechanical Room 1-1 Regional Physiographic Map 1-2 Regional Pre-Pleistocene Sediments of the Continental Margin 1-3 Site Surficial Geology 1-4 Regional Geologic Map 1-5 Regional Geologic Section 1-6 Regional Tectonic Map 1-7 Stratigraphic Correlation Chart for the Site and Surrounding Region) 1-8 Regional Stratigraphic Correlation Chart (Sheet 1) 1-9 LANDSAT Photographs of Connecticut, Rhode Island, Southern Massachusetts, and Eastern New York 1-10 Lineament Map from LANDSAT Photographs 1-11 Regional Aeromagnetic Map 1-12 Regional Bouguer Gravity Map 1-13 Site Bedrock Geology 1-14 Tectonic Map of Eastern Connecticut 1-15 Contour Diagram of Poles to Foliation Planes - Final Grade
List of Figures (Continued) mber Title 1-16 Contour Diagram of Poles to Joint Planes - Final Grade 1-17 Contour Plot of Bearing and Plunge of Slickensides - Final Grade 1-18 Generalized Location of Faults 2-1 Location of Seismic Stations 2-2 Epicenters of Earthquakes within 200-Mile Radius 2-3 Location of Earthquakes within the 50-Mile Radius 2-4 Isoseismal Map, Earthquake of November 9, 1727 2-5 Isoseismal Map, Earthquake of November 18, 1755 2-6 Isoseismal Map, Earthquake of May 16, 1791 2-7 Isoseismal Map, Earthquake of August 10, 1884 2-8 Isoseismal Map, Earthquake of March 1, 1925 (February 28, 1925 EST) 2-9 Isoseismal Map, Earthquakes of December 20 and 24, 1940 2-10 Tectonic Provinces 3-1 T-2 Fault Zone, Final Excavation Grade - Northern Section 3-2 T-2 Fault Zone, Final Excavation Grade - Southern Section 3-3 T-3 Fault Zone, Final Excavation Grade 4-1 Geologic Map of Final Grade, Service Water Line Walls - East 4-2 Geologic Map of Final Grade, Service Water Line Walls - West 4-3 Geologic Map of Final Grade, South Wall of Discharge Tunnel 4-4 Geologic Map of Final Grade, North Wall of Discharge Tunnel 4-5 Geologic Map of Final Grade, East Wall of Discharge Tunnel 4-6 Geologic Map of Final Grade, Floors of Structures 4-7 Geologic Map of Final Grade, Service Water Line Floor - West 4-8 Geologic Map of Final Grade, Pumphouse Floor 4-9 Geologic Map of Final Grade, Service Water Line Floor - East 4-10 Geologic Map of Final Grade, Southeast Quadrant of Containment Walls 4-11 Geologic Map of Final Grade, Southwest Quadrant of Containment Walls 4-12 Geologic Map of Final Grade, Northwest Quadrant of Containment Walls
List of Figures (Continued) mber Title 4-13 Geologic Map of Final Grade, Northeast Quadrant of Containment Walls 4-14 Geologic Map of Final Grade, Engineered Safety Features, Building Sump Walls 4-15 Geologic Map of Final Grade, Auxiliary Building Pipe Tunnel Pit Walls 4-16 Geologic Map of Final Grade, North Wall of Excavation 4-17 Geologic Map of Final Grade, Northeast and Southeast Pumphouse Walls 4-18 Geologic Map of Final Grade Engineered Safety Features Building Wall 4-19 Geologic Map of Final Grade Discharge Tunnel Floor 4-20 Geological Map of Final Grade Discharge Tunnel Floor 4-21 Geological Map of Final Grade North Wall of Discharge Tunnel 4-22 Geological Map of Final Grade South Wall of Discharge Tunnel 4-23 Geologic Map of Final Grade Discharge Tunnel Floor 4-24 Geologic Map of Final Grade Discharge Tunnel Floor 4-25 Geologic Map of Final Grade West Wall of Discharge Tunnel 4-26 Geologic Map of Final Grade East Wall of Discharge Tunnel 4-27 Geologic Map of Final Grade Discharge Weir Rock Face 4-28 Corrected Blow Count Plot, Pumphouse Area Sands, Onshore Boring Composite 4-29 Corrected Blow Count Plot, Pumphouse Area Sands, Borings P1 to P8 Composite 4-30 Grain Size Distribution Curves (Sheet 1) 4-31 Boring Location Plan 4-32 Plot Plan Showing Locations of the Borings and the Geologic Sections 4-33 Geologic Profile, Sections A-A', B-B' 4-34 Geologic Profile, Sections C-C', D-D', E-E' 4-35 Geologic Profile, Sections F-F" and G-G' 4-36 Top of Basal Till Contour Map 4-37 Groundwater Contour Map 4-38 Groundwater Observations in Boreholes 4-39 Bedrock Surface Contour Map 4-40 General Excavation Plan
List of Figures (Continued) mber Title 4-41 Shorefront and Dredging Plan 4-42 Modulus vs Effective Confining Pressure, Structural Fill 4-43 Lateral Pressure Distribution 4-44 Gradation Curves for Category I Structural Fill 4-45 K2 vs Shear Strain for Beach Sands 4-46 Earthquake Induced Shear Stresses in Beach Sands 4-47 Cyclic Stress Ratio vs Confining Pressure for Beach Sands 4-48 Cyclic Stress Ratio vs Penetration Resistance of Sand 4-49 Factor of Safety Against Liquefaction of Beach Sands 4-50 Idealized Soil Profile Liquefaction Analysis of Ablation Till Under Discharge Tunnel 4-51 Geologic Profile, Section H-H
4-52 Geologic Profile, Section I-I
4-53 Location of Field Density Tests - Service Water Intake Line 4-54 Location of Field Density Test - Emergency Generator Enclosure and Control Building 4-55 Geologic Profile, Section J-J' 4-56 Geologic Profile, Section K-K' 4-57 Grain Size Distribution Curves - Pumphouse Area Outwash Sands (Sheet 1) 4-58 Equivalent Numbers of Uniform Stress Cycles Based on Strongest Components of Ground Motion 4-59 Plan of Settlement Monitoring Benchmark Locations 4-60 Control Building Settlement (Sheet 1) 4-61 Emergency Generator Enclosure Settlement 4-62 .Solid Waste Building Settlement 4-63 Liquid Waste Building Settlement 4-64 Fuel Building Settlement 4-65 Geologic Profile Section L-L' 4-66 Geologic Profile Section M-M'
List of Figures (Continued) mber Title 4-67 Geologic Profile Section N-N' 4-68 Geologic Profile Section O-O' 4-69 Geologic Profile Section P-P' 4-70 Geologic Profile Section Q-Q' 4-71 Geologic Profile Section R-R' 4-72 Soil-Structure Interaction Emergency Generator Enclosure 4-73 Shear Modulus Curve Type 2 Soil (Structural Backfill and Basal Till) 4-74 Damping Curve Type 2 Soil (Structural Backfill and Basal Till) 4-75 Shorefront Profile Used in Liquefaction Analyses 5-1 Section through Shorefront 5-2 Typical Wedge Geometry 5-3 Design Loads for Ring Beam 5-4 Shorefront Slope Stability Section - Sloping Rock Profile 5-5 Summary of CIU Test Results - Beach Area Outwash Sands 5-6 Potential Failure Wedges West Side of Containment Excavation 5-7 Rock Surface Near North Edge of Main Steam Valve Building
stability of the soil and rock underlying the Millstone Nuclear Power Station - Unit 3 ndations was evaluated using the results of detailed field and laboratory investigations, both r to and during construction. The field investigations consisted of borings, standard etration tests, piezometer installations, water pressure tests, geologic mapping, and seismic eys to determine compressional and shear wave velocity. Laboratory testing was conducted to rmine the physical properties of the soil and rock. A detailed listing of the site investigation gram is included at the beginning of Section 2.5. Evaluations of the subsurface conditions, soil rock properties, and results of stability analyses are presented herein. Analyses incorporate vibratory ground motion associated with the safe shutdown earthquake (SSE) where ropriate.
4.1 GEOLOGIC FEATURES geologic setting and site structural geology of the Millstone 3 site is discussed in tions 2.5.1.2 and 2.5.3, and the local geology is shown on the site bedrock geology map ure 2.5.1-13).
rock surface, mapped prior to excavation, is fresh with few zones of weathering. The thering is not excessive and occurs generally in highly jointed areas or along a fault zone. The of rock has been glacially smoothed and eroded by outwash waters. Many of the joints have n filled with glacial till. In the southern portion of the main excavation in the discharge tunnel
, six low angle joints (394, 398, 424, 425, 577, 645) exhibiting slight displacement due to the ging action of the glacial ice have been mapped. The location of these joints are shown on ures 2.5.4-1 through 2.5.4-5.
evidence of large stress concentrations developed during the rock excavation for Millstone 3.
re was no observable stress relief in the form of popping rock, rock bursts, or notable rock vement. No significant problems were noted from rock stresses in the Millstone Point quarry le and Gregory 1911; Dale 1923). However, Niles (1975-76) indicated that the thin webs of k between closely spaced holes had popped while line drilling, and that the drills had become nd.
close spacing of the drill holes and the binding of the drills were probably caused by the ase of the residual stress in the rock mass.
Geologic Mapping During Construction al excavation grades and most of the top of rock were geologically mapped during excavation the safety related structures. A summary report of the mapping of the bedrock surface and e subsequent reports concerning faults subsequently uncovered at the site have been mitted to the Nuclear Regulatory Commission (NRC) (NNECo. 1975, 1976, 1977, 1982).
ults of site geologic mapping are discussed in Sections 2.5.1.2 and 2.5.3. Field sketches were pared for the floors of structures at the scale of 1:120, and the walls and the major fault zones e prepared at a scale of 1:60. These scales have been reduced in this document for publication
bols, and slickensides correspond to descriptions listed in Tables 2.5.4-1, 2.5.4-2, and 2.5.4-espectively.
geologic maps of the containment and engineered safety features (ESF) building excavation ls are shown on Figures 2.5.4-10 through 2.5.4-14 and 2.5.4-18. Tables 2.5.4-4, 2.5.4-5, and 4-6 list the joint, foliation, and slickenside information for the containment and ESF dings, respectively.
excavation walls of the auxiliary building pipe tunnel pit and the north wall of the excavation shown on Figures 2.5.4-15 through 2.5.4-17. The service water pipeline walls and discharge nel excavation floor and walls are shown on Figures 2.5.4-1 through 2.5.4-5 and 2.5.4-19 ugh 2.5.4-27. Lists of joint, foliation, and slickenside information are given in Tables 2.5.4-7, 4-8, and 2.5.4-9, respectively.
igneous intrusions and biotite concentrations that cross the site are numbered for continuity for distinguishing the different intrusions that cross discontinuities caused by faulting and ation differences in the excavation.
faults uncovered at the site are shown on Figures 2.5.4-6 and 2.5.4-19 and are listed in le 2.5.3-1. The nature and age of the faults are discussed in detail in Section 2.5.3.2.
4.2 PROPERTIES OF SUBSURFACE MATERIALS eries of investigations was conducted in the field and in the laboratory to determine the perties of the subsurface materials existing at the site and the compacted backfill materials cessed from offsite sources. Materials underlying the site include beach sand, unclassified am deposits, ablation till, basal till, and hard, crystalline bedrock of the Monson Gneiss mation. The field investigations included soil and rock borings, geologic mapping, piezometer allation and monitoring, water pressure testing of the bedrock, seismic refraction and ection surveys, and cross-hole and up-hole seismic surveys. The field testing is described in il in Section 2.5.4.3. The laboratory investigations included index property and gradation rminations of onsite soils, moisture-density relations, and direct shear testing of compacted kfill, shear modulus, and damping determination and cyclic and static triaxial testing of beach ds, unconfined compression testing of bedrock core samples, and joint and foliation friction rmination for bedrock surfaces.
oratory testing of site soils and backfill source materials was conducted in the Stone &
bster Engineering Corporation (SWEC) Soils Laboratory. Field testing for backfill control ng placement was conducted in the SWEC Field Quality Control Laboratory, located onsite.
mpacted backfill test results are discussed in Section 2.5.4.5.2. Intact rock core specimens e tested for unconfined compressive strength and unit weight by Prof. K. Tsutsumi of Tufts versity. Results of these tests are presented in Table 2.5.4-10. Direct shear tests along jointed foliated rock surfaces on specimens selected from NX core samples were performed in the EC Soils Laboratory. A description of these tests is presented in Section 2.5.5.2 and data are
he pumphouse. The results of this study are presented in the GEI report, Appendix 2.5F.
solidated undrained (CIU) tests were also performed on samples of the beach sands in the EC Soils Laboratory. The results of these tests are tabulated in Table 2.5.4-12 and endix 2.5G.
rlying the bedrock at the Millstone site are five groups of soils. They are from youngest to est: artificial fill, beach deposits, unclassified stream deposits, ablation till, and basal till. Each hese is discussed in the following sections.
4.2.1 Artificial Fill ficial fill material is comprised of a mixture of till, waste rock materials excavated from the lstone 1 and 2 sites, and some quarry waste. Consequently, it is a heterogeneous mixture.
se fill materials were not placed in controlled thin-lift construction and are not a satisfactory ndation material for structures of any kind. All artificial fill has been excavated when ountered and no structures, pipelines, or electrical ducts are founded on this material.
4.2.2 Beach Deposits beach deposits are the youngest naturally occurring material in the site area. These are ent for the most part only in the cove east of Bay Point, in the area of the circulating and ice water pumphouse. For the most part they consist of uniform silty sand. The beach deposits generally denser than the alluvium deposits due to wave action from Long Island Sound.
ic and cyclic triaxial and resonant column tests were performed on the beach deposits to estigate liquefaction potential and obtain shear strength parameters for slope stability analyses he shoreline area. The results of these tests are tabulated in Table 3 of Appendix 2.5F. These lyses are discussed in Section 2.5.4.7, 2.5.4.8, and 2.5.5.2.
mposite plots of relative density and corrected blow count (N) vs effective overburden stress ed on Gibbs-Holtz relations are presented on Figures 2.5.4-28 and 2.5.4-29, respectively.
se plots show that the beach sand is a medium dense deposit with an average relative density pproximately 70 percent, with most points denser than 60 percent. Some points do plot lower, these low density values are generally indicative of the looser, unsaturated sand near the und surface.
major plant structure, pipeline, or duct is founded on the beach deposits. This material was avated and replaced with compacted select backfill under portions of the service water line, remains in place along the shoreline, adjacent to the circulating and service water pumphouse.
4.2.3 Unclassified Stream Deposits lassified glacial stream deposits west and southwest of Millstone 3 consist of sands with some and gravels. Thicknesses of the deposits vary, and exposed cuts reveal the sediment to be
nt of the unclassified stream deposits is shown on the site surficial geology map ure 2.5.1-3).
major plant structure, pipeline, or duct is founded on unclassified stream deposits. Prior to allation of any foundations, all underlying loose deposits were removed to sound basal till or rock and replaced with compacted backfill, as discussed in Section 2.5.4.5.2.
4.2.4 Ablation Till ation till overlies the dense basal till in the area where the major plant structures are located.
s material consists of glacially transported debris which was deposited as the supporting and/
nclosing ice melted away from it. The ablation till has not been compacted by ice and is, efore, less dense than the basal tills, but is still a strong, stable soil. Both the basal till and the rlying ablation till are relatively impervious. The ablation till is more pervious than the basal because it is irregularly stratified with lenses of sand and gravel and mixtures of cobbles, vels, sands, and silts.
dation analyses and moisture content determinations were conducted on split spoon samples he ablation till. The gradation curves indicate that the ablation till is a silty sand, with typically o 40 percent finer than the No. 200 sieve. The gradation curves are presented on Figure 2.5.4-plotted with the Lee & Fitton (1969) and Kishida (1969) gradation envelopes of soils most ly to liquefy during the earthquake. The ablation tills at the Millstone site are significantly e widely graded and coarser than the soils typified by these envelopes. The natural moisture tent of the ablation till varies from 5 to 15 percent. Moisture content determinations from split on samples of various overburden materials at the site is presented in Table 2.5.4-13.
roximately 500 feet of the circulating water discharge tunnel in the vicinity of Millstone stack ounded on crushed stone and concrete fill overlying ablation till. At all other structures, the tion till was removed to sound basal till or bedrock and replaced with compacted backfill, if uired, as discussed in Section 2.5.4.5.2.
4.2.5 Basal Till al till overlies bedrock at the site area, varying in thickness from less than 5 feet in the phouse area on Niantic Bay to over 40 feet under the turbine building. The basal till is a very se material of low permeability consisting of a widely graded mixture of cobble and boulder-rock fragments, gravel-size material, sand, and some silt binder. The basal till was overridden compacted by ice during the glacial period, accounting for its characteristic very dense state high strength.
dation analyses and moisture content determinations were conducted on split spoon samples he basal till. Although only the minus 1 inch portion of the basal till was tested, the gradation ves presented on Figure 2.5.4-30 show that the basal till consists of a widely graded silty sand
) with 10 to 25 percent finer than the No. 200 sieve and a coefficient of uniformity (D60/D10)
le 2.5.4-14. A detailed discussion of the liquefaction potential of the basal till is presented in tion 2.5.4.8.2 and a discussion of the static stability of structures founded on basal till is uded in Section 2.5.4.10.1.
stic constants have been determined by seismic cross-hole and up-hole surveys. Details of this y are presented in Section 2.5.4.4.3 and Appendix 2.5HThe average Young's modulus (E) rmined for the basal till was 4 x 105 psi and the average shear modulus (G) determined was x 105 psi. A Poisson's ratio of 0.44 has been calculated based on these values of E and G.
4.2.6 Monson Gneiss country rock at the site is the Monson Gneiss. At the site area, the Monson Gneiss is thinly red with light feldspathic and dark biotitic and hornblendic layers. The foliation is well ned and exhibits a consistent northwest trend. Based on data accumulated during geologic ping at the site during excavation, the average foliation attitude of the Monson Gneiss is W, 48NE, (N54W, 48NE grid north). A stereonet projection of the foliation is presented on ure 2.5.1-15.
ting at the site is characterized by an average attitude of N03W, 63NE (N10E, 63SE). (All kes are referenced to true north, which is 13.5 degrees east of grid north. Bearings in ntheses represent grid north.) Minor joint sets observed at the site are N02W, 78NW, (N11E, W) and N69E, 74SE (N82E, 74SE). A low angle joint set oriented at N48W (N35W) dips 7 rees northeast. A lower hemisphere stereonet plot of poles to the joint planes is shown on ure 2.5.1-16 and a complete list of all measured joints and foliations is presented in les 2.5.4-1, 2.5.4-4, 2.5.4-7 and 2.5.4-2, 2.5.4-5, and 2.5.4-8, respectively. In general, the ts are linear and tight and exhibit smooth surfaces. A large number of the joints are coated h chlorite and many exhibit iron oxide staining.
ect shear tests were performed on several joint and foliation surfaces. These tests indicate that average residual shear stress for joint surfaces is 34.5 degrees, and the average residual shear ss for the foliation is equal to 32 degrees. Details of the testing program are presented in endix 2.5I.
onfined compression and density tests were performed on nine core samples of Monson iss and two samples of Westerly Granite. The unconfined compressive strength of the nson Gneiss varied from approximately 4,000 to 14,000 psi, with an average value of 10,000 The unit weight of Monson Gneiss ranged from 161 to 168 pcf, with an average value of 165 The Westerly Granite was slightly stronger and less dense. The average unconfined pressive strength of the two samples was approximately 13,000 psi and the unit weight raged 157 pcf. The results of the rock compression tests are tabulated in Table 2.5.4-10.
eophysical survey was performed, consisting of measuring compressional P wave and sverse S wave velocities using both down-hole and cross-hole techniques. Average values
e 4 x 106 psi, 1.5 x 106 psi, and 0.33, respectively. The geophysical investigations are ussed in detail in Section 2.5.4.4.3 and Appendix 2.5H.
4.3 EXPLORATION otal of 95 test borings, both vertical and inclined, were drilled in the rock and soil at the site.
boring locations are presented on Figures 2.5.4-31 and 2.5.4-32. Table 2.5.4-15 is a listing ll boring coordinates, ground elevations, top of rock elevations, and groundwater elevations at time of drilling. Complete boring logs are presented in Appendix 2.5J. The logs describe the and rock types, the location, elevation, and type of samples recovered, the standard etration test value (N), and the core recovery and rock quality designation (RQD) of the rock. Geologic profiles are presented on Figures 2.5.4-33 through 2.5.4-35 and the basal till ace contour map is presented as Figure 2.5.4-36.
locations of boreholes in which water levels were taken are shown on Figure 2.5.4-37.
undwater elevations were monitored in borings 301 to 310 prior to construction, and the undwater fluctuations over a 2-year period for borings 303, 310, 311, 312, and 317 are shown phically on Figure 2.5.4-38. These wells were disturbed during construction; therefore, there o record reported in these wells subsequent to December 1973. Site groundwater conditions, ed on regional data, site piezometers, and observations during construction, are discussed in il in Sections 2.4.13 and 2.5.4.6.
ter pressure tests were performed in three borings to assess the degree of weathering and meability of the bedrock. The results of the tests are presented in Table 2.5.4-16.
eismic refraction survey to determine compression wave velocities and depths to various strata performed by Weston Geophysical Engineers, Incorporated (WGEI) and is discussed in tion 2.5.4.4.1. The location of the seismic refraction lines and the seismic profiles are ented in Appendix 2.5K.
mic cross-hole and up-hole techniques were employed at the site in order to determine the es of dynamic moduli and Poisson's ratio for the ablation till, basal till, and bedrock. The lts are tabulated and discussed in Section 2.5.4.4.3. The WGEI report on these tests is ented as Appendix 2.5H.
4.4 GEOPHYSICAL SURVEYS physical surveys were conducted to determine the nature and extent of subsurface materials at site. The studies included a seismic refraction survey of the site in the vicinity of the major ctures, an offshore seismic and bathymetric survey employing refraction and reflection niques, and seismic cross-hole and down-hole surveys to determine compressional and shear e velocities of subsurface materials.
eismic refraction survey was performed by WGEI to investigate subsurface conditions at the
. The purpose of the study was to determine compression wave velocities and depths of surface materials, and to prepare a preliminary bedrock contour map of the site area.
d procedures employed during the refraction survey are detailed in Appendix 2.5K.
refraction survey identified three major strata at the site according to seismic velocity. The r surface overburden material, identified as ablation till and discussed in detail in tion 2.5.4.2.4, typically has a seismic velocity ranging from 1,500 to 2,000 fps, indicative of a ium dense to dense, unconsolidated material. The transition between saturated ablation till moderately dense basal till corresponds to a zone with a seismic velocity between 5,000 and 0 fps. The very dense basal till, discussed in detail in Section 2.5.4.2.5, has a seismic velocity pproximately 6,700 fps. The thickness and extent of each of the overburden strata are shown he subsurface profiles in Appendix 2.5K.
harp increase in the seismic velocity was observed at the bedrock surface, indicating the ence of any extensive zones of weathered rock. This was verified during excavation for ctures. Typical seismic velocity values for the bedrock were approximately 12,000 fps, cative of a hard, massive, unweathered rock type. The soundness of the rock has been verified m the logging of rock cores from boreholes and from geologic mapping. The rock contour map ined from the seismic survey and shown on Sheet 3 of 8 in Appendix 2.5K agrees with the tour map of the bedrock surface shown on Figure 2.5.4-39, which is based on actual survey of the rock surface measured during construction.
4.4.2 Offshore Seismic and Bathymetric Survey eismic and bathymetric survey was conducted by WGEI to contour the Long Island Sound om and the bedrock surface offshore from Millstone Point, in the vicinity of the intake and harge structures, as shown on Figure 1 of Appendix 2.5L. Detailed profiling of the bedrock ace was obtained in some areas by means of continuous reflection techniques. Velocity values the different materials were determined from a seismic refraction survey. These values were d in computing depths to the reflecting horizons and for identifying the type of overburden erial and the quality of the bedrock.
bedrock and bottom contour maps for the four areas surveyed are presented in endix 2.5L.
4.4.3 Seismic Velocity Measurements mic velocity measurements using an explosive source were conducted at the site to rmine compressional P wave velocities and transverse, or shear, S wave velocities of the erlying materials. Both down-hole and cross-hole techniques were utilized. Elastic parameters basal till and bedrock obtained from these tests were used as the design basis for foundations hese materials. The field procedure is described in detail in Appendix 2.5H.
l -50 feet. Down-hole velocity measurements were made from el +5 feet to el -99 feet. There good agreement in the values between the two techniques.
ocity measurements of the overburden materials further distinguish between the two tills at the
. Shear wave velocity for the ablation till is approximately one-third lower than for the denser al till.
following seismic velocity profile is representative of materials in the vicinity of the turbine ding:
Seismic P Wave Elevation (ft) Material Technique (fps) S Wave (fps)
+15 to +4 Ablation Till Cross-hole 5,600 1,400
+ 4 to -24 Basal Till Cross-hole 6,800 2,200
-24 to -44 Bedrock Cross-hole 12,800 6,500 following seismic velocity profile is representative of materials in the vicinity of the reactor tainment structure:
Seismic Elevation (ft) Material Technique P Wave (fps) S Wave (fps)
+10 to -50 Bedrock Cross-hole 12,800 6,500
+ 5 to -99 Bedrock Down-hole 13,500 6,500 mic velocity measurements were made using an impact source of shear wave energy to rmine P and S wave velocities of materials underlying the discharge tunnel in the area of Millstone stack. Bedrock is overlain by basal till, ablation till, alluvium, and fill. The rburden in this area is up to 60 feet in thickness. In these tests, geophones were lowered into ch receiving holes to pick up arrival times generated from impact blows on a split-spoon pler positioned at the same elevation. The following seismic velocity profile is representative aterials in the vicinity of the discharge tunnel near the ventilation stack:
Elevation (ft) Material P Wave (fps) S Wave (fps)
+14 to + 2 Fill 1,363-3,060 814-1,238
+ 2 to -13 Alluvium 4,820-5,818 383-684
-13 to -18 Ablation Till 6,053-6,597 398-654
-18 to -30 Basal Till 7,539-7,603 1,246-2,387
rces.
values below are for low strain values from field generated tests and may not necessarily be d as design input. Final values used in design are calculated for individual structures.
Young's Modulus, E Shear Modulus, G Material (psi) (psi) Poisson's Ratio Rock 4 x 106 1.5 x 106 0.33 Basal Till 4 x 105 1.4 x 105 0.44 Ablation Till 2.7 x 104 9.0 x 103 0.49 4.5 EXCAVATIONS AND BACKFILL extent of excavations and backfill for major Seismic Category I structures is shown on ure 2.5.4-40. Final grading, which includes dredging and backfilling in the vicinity of the ulating and service water pumphouse, is shown on Figure 2.5.4-41. Profiles delineating the nt of the excavation and backfill are shown on Figures 2.5.4-33 through 2.5.4-35. Geologic ping of the excavated surfaces is described in Section 2.5.4.1.
4.5.1 Excavation founding materials for major plant structures are listed in Table 2.5.4-14. Most of the major ty related structures are founded on bedrock, with the exception of the control building, rgency diesel generator building, and the hydrogen recombiner building. The control building ounded on basal till. Isolated zones of softened till were excavated and replaced with fill crete or compacted structural backfill. The emergency generator enclosure building wall ings are founded on basal till. The diesel generator pads are supported on approximately 8 feet tructural backfill basal till as shown on Figure 2.5.4-55 (Geologic Profile J-J'). The hydrogen mbiner is founded on concrete fill overlying bedrock.
st of the circulating water discharge tunnel is founded on bedrock. Near the ventilation stack, a distance of approximately 500 feet, the discharge tunnel is founded on crushed stone and crete fill overlying basal till. Section 2.5.4.8.4 and Figure 2.5.4-51 (Geologic Profile H-H")
cribe the founding conditions of the discharge tunnel in this area.
service water intake lines are founded on bedrock in the main plant area; however, between main plant area and the pumphouse they are founded on soil. When soil was encountered as a nding material, all unsuitable overburden was removed to sound basal till. Where the invert ation was higher than the excavated grade, compacted structural backfill was placed in thin to the subgrade elevation of the pipe encasement. All compacted structural backfill was ed in accordance with procedures described in Section 2.5.4.5.2. Figure 2.5.4-52 (Geologic
locations of field density tests of structural backfill placed beneath the service water intake s near the pumphouse, where the deposit of beach and outwash sand was removed above the al till, are presented in Figure 2.5.4-53. Table 2.5.4-19 summarizes the results of the density s in this area.
k in the containment area was blasted and excavated in segmented areas, each approximately eet deep. Rock bolts, discussed in Section 2.5.4.12, were installed in the southwest sector of excavation to prevent potential sliding failures along the foliation. In addition, intercept drains e installed into the southwest excavation face to reduce the hydrostatic pressure on the ation and joint planes. No rock slides were noted during the time the excavation was in ice. However, some areas were overbroken due to blasting and to subsequent scaling rations to remove loosened rock wedges. The overbreak areas were localized and generally ted in size to approximately 2 cubic yards and less. The surfaces of the wedges generally formed to the predominant joint sets mapped at the site and discussed in Section 2.5.4.1. The re and extent of overbreak experienced during site excavation is considered normal for rock of this type and does not indicate instability in the rock mass.
ious techniques were utilized when blasting near the perimeter of structures to limit overbreak minimize damage to adjacent rock. The methods used include line drilling, cushion blasting, plitting, and smooth wall blasting. The purpose of each of these techniques was to develop a ar plane along the perimeter of the excavation so that the excavated rock breaks cleanly from face. In line drilling, the perimeter holes were closely spaced and left unloaded during the
- t. Cushion blasting was used to blast a narrow berm left from a previous blast. A single row of ely spaced holes was drilled along the berm, lightly loaded, and fired simultaneously.
splitting consisted of the firing of a single row of lightly loaded, closely spaced holes, prior to primary blast. The purpose was to produce a crack along the line of presplit holes which the sequent primary blast could break. Smooth wall blasting is similar to cushion blasting except the lightly loaded perimeter holes were the last delay in the blast.
trolled blasting techniques were used to limit the vibrations felt at Millstone 1 and 2 and to lude any structural damage to concrete or bedrock near the blast. Peak particle velocity was sured for each blast, using Sprengnether 3 - component seismographs. No damage to any cture or component in the two operating units or the Millstone 3 construction site was erved as a result of the blasting.
inflow of water into the excavation was controlled by means of pumping from local sumps.
s was possible due to the low permeability of the soils and the tightness of the joints in the rock. Concrete working mats were poured on all foundation surfaces upon excavating each in order to minimize the impact of construction activities on the undisturbed founding aces.
e softening of the basal till in sections of the excavation was observed. The softening is butable to the exposure of the till to the affects of weathering and construction traffic. When
avation, softened till approximately 1 foot in thickness was hand-excavated to firm till and aced with structural backfill. The extent of the softening was verified by excavating two test ches into the till to a depth of 4 feet. No additional softened till was encountered below the ened surface layer. The groundwater level was maintained below the subgrade by pumping m sumps outside the structure, and no seepage infiltrated the excavation after removal of the ened till and placement of the structural backfill.
4.5.2 Backfill egory I structures founded totally or partially on structural backfill include the control building emergency generator enclosure building. In addition, sections of the service water line and e of the buried electrical ducts are founded on Category I structural backfill.
erial used for Category I structural backfill is predominantly obtained from glacial outwash osits located at the Romanella Pit in North Stonington, Connecticut. Test data on borrow erial from the Romanella Pit have been previously reported in July and November 1974 and included in Appendix 2.5M. A small percentage is obtained from other borrow sources having ilar geologic characteristics. A description of the borrow material from three alternate sources ted in the towns of North Stonington, Preston, and Canterbury is included in a report mitted in June 1976 and is included herein as Appendix 2.5M.
structural backfill is processed at the borrow pit by means of passing the soil through a screen, uring that the maximum particle size and gradation meet the backfill specification uirements. For Category I structural fill, the gradation limits are:
U.S. Standard Cumulative Sieve Size Percent Passing 3 inches 100 3/4 inch 75 to 100 3/8 inch 65 to 90 No. 10 40 to 60 No. 40 15 to 35 No. 100 0 to 20 No. 200 0 to 15 fficient of Uniformity, Cu = D60/D10 10.
structural backfill was compacted to 95 percent of the maximum dry density determined from Modified Proctor Test, ASTM D1557, Method D. Moisture content was maintained within
ontinuing program of testing, inspection, and documentation was in effect during construction nsure satisfactory placement of backfill. Category I structural backfill was tested every 500 ic yards for conformance to the specified gradation limits prior to being allowed into the struction area. In addition, the maximum density was determined by ASTM D1557, Method for every 500 cubic yards of fill placed. Field density tests, using ASTM D1556, were ormed for each lift of fill, but not less than one test for every 500 cubic yards of fill placed.
ations of field density tests under the emergency generator enclosure and control building are wn in Figure 2.5.4-54, and the test results are summarized in Table 2.5.4-20. Cross-sections wing generalized subsurface profiles beneath these two structures are presented in ures 2.5.4-55 (Section J-J') and 2.5.4-56 (Section K-K').
ar strength of compacted backfill materials was determined from drained direct shear tests on ples compacted to 95 percent of maximum dry (ATMS D1557) density. Samples tested in the ct shear box contained only the minus No. 4 portion of the sample. For consistency, the imum density was also determined on the minus No. 4 portion of the sample. However, the imum density of the minus 3/4-inch fraction was tested in the field, and it can be assumed that maximum density of the minus No. 4 fraction would be less than the maximum density inable at the site on the whole sample. Consequently, testing the minus No. 4 fraction results alues of shear strength more conservative than would be expected for the whole soil sample. A parison of maximum densities for the minus No. 4 and minus 3/4-inch fractions for esentative samples from the major borrow areas used for Category I structural backfill is ented below:
@ 95%
d max d max d max @ 90%
(-3/4") (-#4) RX (-#4) d max (-#4)
Backfill Source (pcf) (pcf) (deg) (deg)
Romanella Pit (Sample 136.4 129.5 41.5 --
R)
Preston Pit 138.8 131.0 35.0 34.6 No. Stonington Pit 148.0 136.1 37.9 34.0 Canterbury Pit 140.0 131.6 39.4 34.0 Hathaway Pit (Waterford) 129.7 121.1 39.1 --
Ledyard Pit (Soneco) 132.6 122.9 41.4 --
strains:
2 2630 ( 2.17 - e ) 1/2 (2.5.4-1)
G max = ---------------------------------------- ( o )
1+e re:
Gmax = maximum shear modulus in psi e = void ratio o = effective octahedral stress in psi d ratio was calculated assuming full saturation and a water content equal to 12 percent, which esents the water content at full saturation for a density of 95 percent of maximum, based on moisture-density curve for Sample R in Appendix 2.5M. The octahedral stress was assumed e equal to two-thirds of the effective overburden stress for a particular depth. The maximum ar modulus at a depth of 10 feet, which corresponds to the midpoint of the backfill layer eath the emergency generator enclosure building, is 13,400 psi. A profile of Gmax vs effective fining pressure is plotted on Figure 2.5.4-42.
esonant column test was performed on a sample of the structural backfill compacted to 95 ent of a maximum dry density. The values of Gmax plotted on Figure 2.5.4-42 obtained from test are in agreement with the Hardin and Black (1968) equation. The low strain damping o was calculated to be 1.4 percent.
ng's modulus for static strain levels was obtained through an iterative process where a value ertical strain was used to obtain a reduction factor for the G value. The value of E was ulated using the equation:
E = 2G(1+u) (2.5.4-2) re:
u = Poisson's ratio strain level assumed was checked with the expected strain level caused by the structural ing, using the equation:
(2.5.4-3)
= ---------z E
e = Vertical strain z = Increase in vertical stress from structure load E = Calculated value of Young's Modulus the emergency diesel generator enclosure building, the calculated vertical strain was roximately 10-3, and Young's modulus at a depth of 10 feet was approximately 10,000 psi. The ile of E static vs effective confining pressure is also plotted on Figure 2.5.4-42.
kfill placed behind concrete walls is described in Section 2.5.4.10.3.
4.5.3 Extent of Dredging facilitate the flow of water into the service and circulating water pumphouse, an intake channel been dredged to the limits shown on Figure 2.5.4-41. Side and longitudinal slopes of the ke channel are designed at 10 and 5 percent, respectively. The beach slope varies from 20 to percent and is protected with heavy armor, as discussed in Section 2.5.5.1.
ings and laboratory testing in the beach area adjacent to the circulating and service water phouse indicate that the beach sands are generally moderately dense, with occasional thin es of less dense material. Liquefaction analyses of these sands, discussed in tion 2.5.4.8.3.2, indicate that a general liquefaction of the sand adjacent to the pumphouse is hly unlikely. If the looser zones do liquefy, the extent of the failure would be strictly local and ld not cause a massive soil movement into the dredged channel.
4.6 GROUNDWATER CONDITIONS undwater observations have been documented in previous reports (Ebasco 1966; Bechtel poration 1969). Water level readings in borehole piezometers were taken for the Millstone 3 study between 1971 and 1973. In addition, pressure testing of rock in three boreholes and ng installation of rock anchors in the turbine and service buildings was conducted to rmine the permeability of the rock mass. Also, temporary drains were installed in sections of containment excavation face and the inflow of water into all excavations was observed ughout construction. These observations form the design bases for groundwater at the site, as ussed below.
4.6.1 Design Basis for Groundwater undwater observations at the site prior to construction were made in piezometers installed in eral borings. Listings of the water elevations and dates of reading are presented in Table 2.5.4-Three borings, 303, 310, and 311, were continually monitored over a 2-year period. A plot of ation vs date for water levels in these boreholes is shown on Figure 2.5.4-38. As a result of e observations, a stabilized groundwater level contour map, based on the water levels
alized perched groundwater conditions probably exist because of the irregular distribution of tion till materials of varying gradation and porosity. It is also likely that shallow, ponded er exists in localized bedrock troughs. The prevalence of bedrock outcrops to the north and hwest of the site indicates that bedrock acts as a groundwater divide, isolating the soils of the of Millstone Point from soils further inland. Thus, groundwater recharge would primarily be to absorption of local precipitation, with probable migration of the waters to the immediately cent Long Island Sound. Little groundwater is present in the crystalline bedrock, and virtually f the groundwater movement is restricted to the soil overburden.
asurements taken during previous investigations (Bechtel Corporation 1969) showed average ux rates into test pits of about 8 gallons per hour, and it was concluded that both the ablation basal tills were relatively impervious. The ablation till soils are more pervious than the basal and occasionally exhibit partial stratification, including sporadic sand lenses. Thus, the upper ions of the soil transmits water more readily than the underlying dense basal tills.
structures are designed for the groundwater levels shown in Table 2.5.4-14 which are based groundwater contours plotted on Figure 2.5.4-37. No safety-related permanent dewatering em is required to lower groundwater levels. These groundwater contours represent average undwater elevations of the site prior to the start of construction. A comparison of groundwater tours with the top of basal till contours on Figure 2.5.4-36 verifies that the primary medium groundwater flow is the permeable surficial soil overlying the basal till. Recharge of the undwater occurs mainly from precipitation infiltrating through the surficial soils, and flowing ard Long Island Sound and the outwash deposits above the till.
struction of the plant results in large changes to the site geohydraulic conditions. Site grade been lowered to a uniform elevation of +24 feet from the original site grade which varied m elevation 26 feet to 30 feet. The major plant structures are founded at approximately ation 0 feet on blasted rock excavations and backfilled from subgrade level to the ground ace with fill materials of relatively high permeability. The backfilled zones under and around e structures and the circulating water intake pipelines provide a continuous hydraulic conduit groundwater flow from the plant area to Long Island Sound. Therefore, the average water ls prior to construction are not necessarily representative of post-construction groundwater ditions. Design groundwater levels used in plant design are shown in Table 2.5.4-14.
eepage diversion system, consisting of a series of underdrains and porous concrete, has been alled under and around several structures to minimize the amount of seepage into the ement of structures founded below the groundwater table. The quantity of seepage expected to iverted through the system is small, due to the low permeability of the basal till and rock at site. This system is not considered safety related because dewatering is not necessary to ensure stability of any structure. However, enough leakage occurs to require pumping for equipment ection. The containment and all other Category I structures are protected from groundwater ow by a waterproof membrane below the groundwater level. Water which penetrates or umvents the membrane is diverted to the Engineered Safety Features Building porous
ty-related pump (see Section 9.3.3 for details of this system.)
ter levels measured in borings taken at the site in early 1972 indicate a groundwater ometric surface with a 3-percent gradient generally sloping from northeast to southwest, as wn on Figure 2.5.4-37.
discussed in Section 2.4.2.2, Flood Design Considerations, the controlling event for flooding he Millstone 3 site is a storm surge resulting from the occurrence of the probable maximum icane (PMH). The maximum stillwater level resulting from hurricane surge was calculated to elevation 19.7 feet msl. As shown on Figure 2.4-9, the water level drops significantly with e, so that after 2 hours2.314815e-5 days <br />5.555556e-4 hours <br />3.306878e-6 weeks <br />7.61e-7 months <br /> the flood level is at elevation 17 feet and after 6 hours6.944444e-5 days <br />0.00167 hours <br />9.920635e-6 weeks <br />2.283e-6 months <br /> the surge level sides to elevation 10 feet. A continuous hydraulic connection would occur across the site from main structure area to the shorefront through the backfill placed around structures and the kfill placed in the circulating water pipeline trench. It can be expected that the maximum undwater level due to flooding would not exceed elevation 19.7 feet and would probably be because of head losses in the soil. According to Figure 2.4-9, the water level drops to 17 feet r 2 hours2.314815e-5 days <br />5.555556e-4 hours <br />3.306878e-6 weeks <br />7.61e-7 months <br />.
design groundwater levels for major safety-related structures shown on Table 2.5.4-14 are all al to or greater than elevation 19 feet with the exception of the hydrogen recombiner building, ch has a design groundwater level of 18 feet. However, founding grade is at elevation 20 feet this structure, which is founded on concrete fill placed directly on bedrock. Design criteria for d conditions are discussed in Section 3.4.
4.6.2 Groundwater Conditions During Construction ing construction, the inflow of water into the excavations was controlled by pumping from ps located outside of the building lines adjacent to structures. Most flow through the rburden was transported through the sand lenses. All water-softened material was removed replaced with a fill concrete working mat as described in Section 2.5.4.5.1. The rate of inflow sufficiently low to allow enough time to pour the concrete working mat without further ening of the till.
inage pipes were installed in the southwest face of the containment excavation in order to eve the hydrostatic pressure on the bedrock joint and foliation surfaces. Very little water was erved flowing through these pipes, indicating that the quantity of flow through the bedrock is ll and that the permeability of the rock is low.
ter pressure tests were performed in three boreholes prior to construction. These tests indicated the rock within the site area is generally massive with slight to moderate interconnected ting. A summary of the water pressure test data from the boreholes is included in Table 2.5.4-Additional pressure tests were performed prior to installation of rock anchors in the turbine service buildings. These tests further verified the low permeability of the rock mass.
4.7 RESPONSE OF SOIL AND ROCK TO DYNAMIC LOADING Seismic Category I structures and associated piping are founded either on bedrock, basal till, tructural backfill. Portions of the circulating water discharge tunnel are founded on ablation in the vicinity of the Millstone stack north of Millstone Unit 1. A listing of the founding strata all Category I structures is included in Table 2.5.4-14.
d crystalline bedrock forms the basement complex of the area. The overlying dense basal till sists of a hard, compact soil which has been heavily preloaded by continental ice. Static and amic properties of the basal till and bedrock are discussed in Sections 2.5.4.2.5 and 2.5.4.2.6, ectively. Static and dynamic properties for the compacted structural backfill are discussed in tion 2.5.4.5.2.
bedrock, basal till, ablation till, and structural backfill are stable materials under vibratory ion caused by the SSE. The basal till, ablation till, and structural backfill are not susceptible to efaction, as discussed in Section 2.5.4.8.
soil-structure interaction analyses for Seismic Category I structures founded on soil were ormed using the computer program PLAXLY-3. The nonlinear behavior of the subgrade was ounted for by use of the computer program SHAKE (LaPlante and Christian 1974) which was d to determine the strain-corrected soil properties. The subsurface material properties used in SSI analysis are discussed in Section 2.5.4.7.1. The method of SSI analysis and the results are ussed in Section 3.7.2.4.
response of buried piping to seismic loadings is discussed in Section 3.7.3.12.
shorefront west of the circulating and service water pumphouse consists of a structural fill beach and outwash and slope varying from 5H:1V to 10H:1V, protected by graded layers of or stone. A plan showing the extent of the shoreline protection system is presented on ure 2.5.4-41. A typical section is shown on Figure 2.5.5-1. Static and dynamic properties of beach sands are discussed in Section 2.5.4.2.2 and documented in the reports in endix 2.5F and 2.5G. The liquefaction potential of the beach and outwash sand is discussed in tion 2.5.4.8. The stability of the shoreline slopes under static and dynamic loading is discussed ection 2.5.5.2.
service water intake pipes, between the circulating and service water pumphouse and the n plant area, are embedded in a rectangular concrete encasement. Soils encountered in the eline excavation include beach and outwash sands, unclassified stream deposits, and ablation These soils were removed under the pipeline to dense basal till and replaced with Category I ctural backfill. The fill was placed at a 1:1 slope from the till surface to the base of the asement and compacted to the requirements outlined in Section 2.5.4.5.2. The sides of the asement were backfilled with nonstructural fill similar to the material used to backfill behind
4.7.1 Subsurface Material Properties Used in SSI Analysis subsurface profiles used in the soil-structure interaction analyses for the control building and emergency generator enclosure (EGE) are idealized, horizontal profiles based on subsurface lorations conducted at the site and described in Section 2.5.4.3. Both of these structures are nded on dense basal till overlying bedrock. The computer program SHAKE was used to rmine strain corrected values of shear modulus obtained from low strain values previously rmined from field testing, laboratory testing, or empirical formulae based on laboratory test
. The program iterates to obtain values of modulus that are compatible with strain levels uced in a particular soil layer by a specific earthquake. The strain levels normally induced by hquakes of magnitudes similar to the Millstone SSE are several orders of magnitude higher the low strain levels achieved during laboratory or field testing, resulting in a reduction in ar modulus when these properties are corrected for strain and input into PLAXLY-3.
soil-structure model used in the EGE analyses is shown on Figure 2.5.4-72. This idealized ile was selected to conservatively model the subsurface conditions under the EGE and in the
-field. The geologic profiles presented in Figures 2.5.4-55, 2.5.4-56, and 2.5.4-71 indicate the rock surface slopes from approximately elevation 0 feet at the east end of the structure to ast elevation -10 feet at the west end. In the north-south direction, the sloping evacuation face ween the control building and the south end of the EGE was backfilled with structural fill over basal till. The extent of structural fill is shown on Section J-J (Figure 2.5.4-55) and ure 2.5.4-54. Because the depth and extent of the structural fill under the EGE is limited, it assumed that the model used in the SHAKE analysis is sufficiently conservative to account local variations in the subgrade and their effect on structural response.
soil properties input into the SHAKE calculation are listed in Table 2.5.4-21A for the free-d model and Table 2.5.4-22B for the structure-effects model. Three earthquake time histories, m the Taft, Helena and Parkfield earthquakes, were normalized to the site SSE peak eleration value of.17g and input at bedrock. Shear modulus and damping iterations were ormed within the SHAKE program in accordance with the curves marked Resonant Column t on Figures 2.5.4-73 and 2.5.4-74. These curves were developed from empirical formulae resonant column tests performed on samples of compacted structural fill from the Millstone
. These test results are presented on Figure 2.5.4-42. The tests show good correlation with ves present by Seed and Idriss in the SW-AJA report (1972).
strain corrected values of shear modulus and damping in the free-field are presented in le 2.5.4-21A. The mean value for each layer was calculated and used to represent the vidual soil layer properties used in the PLAXLY model shown on Figures 3.7B-11 and 3.7B-The Millstone site artificial earthquake was input at bedrock and the soil was modeled as a te element mesh. The use of SHAKE to perform shear modulus and damping iterations cludes the need to iterate in the PLAXLY model. A discussion of the soil-structure interaction lysis is presented in Section 3.7B.2.4.
. Shear wave velocities were used to define soil stiffness. The low strain and strain-corrected properties for the free field case are listed in Table 2.5.4-23.
4.8 LIQUEFACTION POTENTIAL foundation materials beneath some of the Seismic Category I structures consist of limited ths of dense to very dense basal tills and/or compacted select granular backfill. These erials are not susceptible to liquefaction under earthquake motions as described in the owing sections.
4.8.1 Structural Backfill ed on studies of soils where liquefaction has been observed (Seed 1968, Lee and Fitton 1969, hida 1969), it is concluded that the structural backfill described in Section 2.5.4.5.2 in areas w the groundwater table is not susceptible to liquefaction, as discussed below.
- 1. A liquefiable soil is generally a uniform sand with a uniformity coefficient of not more than 10 (Kishida 1969). The structural backfill has a uniformity coefficient ranging from 25 to 50 (Figure 2.5.4-44).
- 2. A soil having a relative density of more than 75 percent is not likely to liquefy (Kishida 1966, 1969; Koizumi 1966; Lee and Seed 1967; Seed and Lee 1966).
Accordingly, compaction criteria of the structural backfill given in Section 2.5.4.5.2 have been designed to yield a relative density higher than 75 percent.
- 3. According to the envelope of most liquefiable soils given by Lee and Fitton (1969), which also contains the envelope given by Kishida (1969), the average particle size, D, of the most liquefiable soils envelope is between 0.02 and 0.7 mm, whereas the corresponding particle size of the structural backfill used is larger than 1.0 mm (Figure 2.5.4-44).
concluded, therefore, that the structural backfill compacted as outlined in Section 2.5.4.5.2 is susceptible to liquefaction during the SSE.
4.8.2 Basal Tills ed on the regional geologic history, the basal tills are very dense deposits consisting of well ded materials ranging in size from boulders to clay (Section 2.5.1.2.3). Figure 2.5.4-30 shows dation curves for the basal till specimens. These specimens were recovered by split spoon pling with a 1 3/8-inch inside-diameter sampler. The gradation curves show that the till is well ded with a uniformity coefficient of about 80 and many particles larger than 3/8 inch. Larger icles present in the till (1 3/8+ inches) could not be recovered by the split spoon. The actual dation curves for the samples would therefore be shifted to the left of those on the figures, lting in a still wider graded soil than shown.
elope from sand deposits which liquefied in Japan. Kishida also notes that a criterion for uefiable soils is that the uniformity coefficient is generally less than 10. The gradation curves the basal till do not satisfy this criterion and are not enclosed within the envelopes developed ither Lee and Fitton or Kishida.
onclusion, the well graded grain size characteristics and the high relative density of the basal preclude the possibility of liquefaction in terms of criteria developed by Kishida (1969) and and Fitton (1969).
4.8.3 Beach and Glacial Outwash Sands circulating and service water pumphouse is located on the shorefront of Long Island Sound, roximately 200 feet west of the Millstone 2 intake structure. The pumphouse is founded on rock; however, the intake channel and adjacent slopes consist of beach and glacial outwash d to approximately el -40 feet. Based on the results of grain size analyses of samples obtained m the pumphouse area, the beach and glacial outwash sands consist mostly of medium to fine silty sand with a few layers of gravelly sand. High concentrations of mica are found ughout the sands in this area.
grain size ranges for beach and outwash sands in the pumphouse area are shown on ure 2.5.4-57. Envelopes of the most liquefiable soils are also shown on these figures.
beach and outwash sands are saturated below sea level. Grain size analyses indicate that a efaction analysis of the sands should be performed to determine whether these sands could efy and slide into the intake channel, causing a potential blockage of the service water inlet es. The analyses described in Sections 2.5.4.8.3.1 and 2.5.4.8.3.2 show that the safety factor inst liquefaction for the beach and glacial outwash sands is greater than 1.1 for the site SSE 7g. Therefore, these sands would not liquefy as a result of the SSE.
4.8.3.1 Dynamic Response Analysis of Beach and Glacial Outwash Sands dynamic response analysis of the shorefront sand deposits has been evaluated to assess the ntial amplification or deamplification of ground motions applied to the bedrock surface. This luation was made using the SHAKE (LaPlante and Christian 1974) computer program for lysis of the vertical transmission of horizontal shear stresses induced by the SSE through a red system. This program treats the strain dependence of the shear modulus and damping ratio n iterative manner.
onservative, idealized profile was selected due to the variability of the rock surface and sisted of 40 feet of sand (the maximum sand thickness in the area) overlying 5 feet of basal till bedrock. The sand layer was divided into four layers, each 10 feet thick, and the till was lyzed as a single layer. Groundwater level was established at 10 feet below the ground surface, corresponds to the mean high water level in Niantic Bay. The shear moduli and damping os of the sand were obtained from tests discussed in Section 2.5.4.2.2 and described in detail in
dulus of the basal till and bedrock was determined from geophysical surveys described in tion 2.5.4.4.1.
values of shear modulus (G) and damping (D) used in the SHAKE analysis for each layer are:
Layer Depth (ft) Soil Type Gmax (ksf) Dmax (%)
1 0-10 Sand 600 1.8 2 10-20 Sand 1,250 1.8 3 20-30 Sand 1,600 1.8 4 30-40 Sand 1,800 1.8 5 40-45 Till 2,500 1.8 reduction of Gmax with strain was performed through a series of iterations, based on the tionship 1/2 G = 1000K 2 ( m ) (2.5.4-4) re:
K2 = a constant that varies with shear strain, developed from resonant column tests and plotted on Figure 2.5.4-45, m = the mean principal effective stress in psf, and G = the shear modulus at a particular shear strain in psf.
time history of the following earthquakes, normalized to the site SSE value of 0.17g, were ut at the bedrock surface:
- 1. 1965 Olympia Earthquake (S86W component)
- 2. 1935 Helena Earthquake (west component)
- 3. 1971 San Fernando Earthquake (Pacoima Dam, N74W component) s analysis indicated that the average maximum acceleration would be 0.27g at ground surface the free-field case of the beach prior to construction of the shoreline slopes. The average shear ss induced by the earthquake, assumed to be 0.65 of the peak value, was calculated to vary m 107 psf at a depth of 5 feet to 410 psf at a depth of 42.5 feet for an effective strain level of than 0.1 percent. Shear stresses vs depth are plotted on Figure 2.5.4-46.
cedures for liquefaction analyses, developed by Seed et al. (Seed and Lee 1966, Seed and ss 1967, 1971), require the following quantitative evaluations:
- 1. The magnitude of shear stresses induced at varying depths in the underlying sand due to earthquakes.
- 2. The resistance of the sands to liquefaction, which may be expressed as the cyclic shear stress necessary to cause initial liquefaction in the number of cycles estimated to occur in an earthquake of the intensity selected (also known as the significant number of cycles to cause liquefaction).
resistance of a soil to liquefaction is expressed as a factor of safety, equal to the ratio of the ar strength available to resist liquefaction to the shear stresses induced by the earthquake.
SSE at the site is based on an Intensity VI-VII earthquake which corresponds to a magnitude pproximately 5.3 using relationships developed by Gutenberg and Richter (1942). Based on ure 2.5.4-58 from Seed, Idriss, et al. (1975), the irregular shear stress time history of the SSE be represented by five equivalent cycles of loading.
shear stresses induced by the SSE were calculated using the SHAKE program, assuming a imum bedrock acceleration of 0.17g. A discussion of the analysis and results for the sand is uded in Section 2.5.4.8.3.1.
cyclic shear stress necessary to cause initial liquefaction, or the shear strength available to st liquefaction, was determined from cyclic triaxial tests conducted on undisturbed samples m borings in the vicinity of the pumphouse. The testing program and results are described in il in Appendix 2.5F.
shear stress necessary to cause initial liquefaction in the field, res, is calculated from the owing equation:
1 - 3 res = v ------------------ C r (2.5.4-5) 2 c cyc
v= Vertical effective stress 1 - 3
- = Cyclic stress ratio 2
c cyc Cr = Reduction factor to be applied to laboratory triaxial test data to obtain the stress conditions causing liquefaction in the field.
the beach and glacial outwash sands at the site, the factor of safety was calculated at depths of 15, 25, 35, and 40 feet below the ground surface. The cyclic stress ratio for a particular depth vertical effective stress was determined from Figure 2.5.4-47, a plot of cyclic stress ratio vs fining pressure. This plot is based on test data from Figure 21 of Appendix 2.5F, assuming efaction occurs at a strain of 10 percent double amplitude.
factors of safety for liquefaction, based on Cr equal to 0.60, are shown in Table 2.5.4-18. The imum factor of safety calculated from laboratory tests is 1.25 at a depth of 40 feet. This factor afety is sufficiently large considering the conservative assumptions included in the analysis.
additional method of assessing liquefaction potential can be developed by comparing standard etration resistance data from the vicinity of the pumphouse structure with standard penetration stance data from sites which have been subjected to earthquakes. This method, described in il below, also indicates there is no danger of liquefaction in the beach sands at the site.
empirical approach relating standard penetration resistance data (N values) to liquefaction ntial was proposed by Seed, Arango, and Chan (1975), who presented cyclic strengths based empirical data from sites which did and did not experience liquefaction during earthquakes.
o included were data from large-scale shake table tests by DeAlba, Chan, and Seed (1975) ch were corrected to account for effects of stress history and multidirectional shaking. Based hese data, Figure 6-1 of Seed et al. (1975) (included herein as Figure 2.5.4-48) presents lower nds of the cyclic stress ratios causing liquefaction versus the standard penetration resistances ands for magnitudes 5 to 6 and 7 to 7 1/2 earthquakes, corrected to an effective overburden sure of 1 ton per square foot (N1) based on the Gibbs and Holtz (1957) correlation of relative sity of sands to blow count and effective stress. A plot of N1 values vs effective stress used in method is the SPT blow count for borings P1 through P8 and I2, I3, I8, I9, and I10 is included igures 2.5.4-28 and 2.5.4-29. The mean value of corrected blow count for these borings was ulated as 20.0, which corresponds to a cyclic stress ratio of 0.278 for a magnitude 5 to 6 hquake, using Figure 2.5.4-48. When compared with the earthquake induced shear stresses ined from the SHAKE analysis described in Section 2.5.4.7, the minimum factor of safety inst liquefaction calculated by this method was 1.68 at a depth of 15 feet.
ery conservative factor of safety against liquefaction was also calculated using a cyclic stress o based on the mean corrected blow count less one standard deviation. An N1 value of 13.1 used to obtain a cyclic stress ratio of 0.185 from Figure 2.5.4-48. The minimum factor of
eptable, considering the fact that the mean value of N1, less one standard deviation, is well w the mean value originally used by Seed et al. in determining the curves in Figure 2.5.4-48.
additional conservatism in the analysis is the use of the magnitude 6.0 relationship for rmining the cyclic stress ratio. The SSE at the site is based on an Intensity VI-VII earthquake, ch corresponds to a magnitude of approximately 5.3, using relationships developed by enberg and Richter (1942).
factor of safety against liquefaction at various depths for each analysis is presented on ure 2.5.4-49. It can be concluded that liquefaction would not occur in the beach and glacial wash sands adjacent to the circulating and service water pumphouse, and that the shorefront is le against sliding failures due to liquefaction of the sand. The stability against sliding of the refront during the SSE is discussed in Section 2.5.5.2.
4.8.3.3 Liquefaction Analyses of Beach Area Sands using 2-Dimensional Dynamic Response Analysis uefaction analyses performed on the sands at the shorefront and discussed in tions 2.5.4.8.3.1 and 2.5.4.8.3.2 were based on the assumption that the subsurface conditions his area could be modeled conservatively as a 40-foot deep uniform sand layer overlying 5 feet asal till and bedrock. An additional analysis was performed in which the sloping bedrock and und surfaces to the west of the circulating and service water pumphouse were incorporated a 2-dimensional dynamic response model to determine earthquake-induced shear stresses.
section selected for the 2-dimensional dynamic response model is similar to the slope ility profile shown on Figure 2.5.5-4. The liquefaction potential of the saturated glacial wash sands was determined by comparing the induced effective shear stresses calculated from dynamic model with the dynamic shear strength of the sand available to resist initial efaction previously determined from corrected blowcount values obtained from standard etration tests performed on beach area borings.
computer program PLAXLY (Plane Strain Dynamic Finite Element Analysis of Soil-cture Systems) was used to calculate earthquake-induced shear stresses within the soil profile.
initial value of low strain shear modulus and damping, total unit weight, and Poisson's ratio he elements were assigned in accordance with the following table.
Depth Gmax Damping Unit Wt Poisson's Soil Type (ft) (ksf) (min) (pcf) Ratio Outwash Sand 0-10 600 0.02 123 0.49 10-20 1,250 0.02 123 0.49 20-30 1,500 0.02 119 0.49 30-40 1,800 0.02 119 0.49 Basal Till 20,160 0.02 145 0.40 Bedrock 216,000 165 0.40 Armor Stone 0-14 900 150
strain compatible shear moduli and damping ratios of the soil were determined through a es of iterations within the PLAXLY program. The time histories of the four earthquakes listed w were normalized to the site SSE peak acceleration of 0.17g and input at the rigid base of the del. These earthquake records were selected because they were recorded at rock sites or stiff sites and therefore would be expected to approximately match dynamic response at the lstone site.
Taft S69E 1952 Kern County Earthquake Helena N-S 1935 Montana Earthquake Pacoima Dam S16F 1971 San Fernando Earthquake Temblor N65W Parkfield Earthquake profile used in the analysis is shown on Figure 2.5.4-75.
uefaction potential was calculated at each element for the six sections shown on this figure.
results of the PLAXLY analysis and the calculated values of safety factor against liquefaction presented in Table 2.5.4-25.
blowcount data used in Sections 1 to 5 were obtained from onshore borings in the shorefront
. The blowcount data from boring I21 was used to represent soil conditions in Section 6 ause the borings indicate that the sands offshore are denser than the onshore sands. The amic shear strength of the sand was calculated by determining the corrected blowcount (N1) ccordance with methods established by Gibbs and Holtz (1957), in which the corrected wcount data are corrected for an effective overburden stress of 1 tsf. The N1 values are plotted h vertical effective stress on Figures 2.5.4-28 and 2.5.4-29. The mean value of N1 was ulated from these data and used to determine the cyclic stress ratio to resist initial liquefaction m the Seed, et al, (1975) curve presented on Figure 2.5.4-48. The curve for Magnitude 6 hquakes was used to obtain a nonliquefaction cyclic stress ratio of 0.27, which was used in the lyses performed on Sections 1 to 5. For Section 6, a mean N1 value of 28 was calculated and a ss ratio of 0.42 was used in the liquefaction analysis.
earthquake-induced shear stresses were computed by averaging the peak shear stress values ained for each of the four earthquakes at each element in the PLAXLY model. The effective ar stress was obtained by multiplying the average of the four peak values by a factor of two-ds. Seed and Idriss (1971) recommend multiplying the absolute maximum shear stress value factor of 0.65 to obtain the equivalent uniform cyclic shear stress. This value was compared h the dynamic shear strength of the soil at each element to obtain the safety factor against efaction.
results of the analyses, presented in Table 2.5.4-25, indicate that the safety factor for ments 1 to 5 are all greater than 1.25. Low safety factors were determined for Section 6, mainly ause of the low vertical effective stress near the surface of the intake channel at elevation -29
nomenon limited to the intake channel only. The post-earthquake slope stability analysis ented in Section 2.5.5.2.1 was reanalyzed to consider the effect of liquefaction of the sand in intake channel (Soil 7 on Figure 2.5.5-4) on stability of the shorefront slopes. The ulations show no change in the safety factor of the critical failure circle, indicating that the refront slopes would not fail in the event that the sand in the intake channel would liquefy.
an be concluded from these analyses that liquefaction of the shorefront slopes would not occur that liquefaction of the intake channel bottom would not affect the integrity of the shorefront es adjacent to the circulating and service water pumphouse or result in a condition that would e the service water system inoperable. The soil underlying the service water pipe encasement cent to the pumphouse is not susceptible to liquefaction.
servatively postulating that liquefaction could occur during the site SSE, a study was made to rmine whether sliding of the slope into the intake channel would cause blockage of the ice water intake pumps. Data from slides caused by liquefaction during the Alaskan thquake of 1964, (Seed, 1968) indicate that flow slides maintain a slope steeper than 5 percent.
uming that the saturated sand overlying basal till adjacent to the pumphouse liquefies and s toward the intake channel, with a final slope of 5 percent, then it can be shown that 7 feet of er remains available for suction below the pump intakes. Therefore, it can be concluded that n in the highly unlikely event that liquefaction of the glacial outwash sands were to occur, the t would have an adequate supply of water available for cooling of safety-related systems.
4.8.4 Ablation Till circulating water discharge tunnel extends 1,700 feet from the main plant area to the lstone quarry east of Millstone 1. For approximately 1,200 feet, the tunnel is founded on rock. However, in the vicinity of the ventilation stack north of Millstone 1, bedrock drops ply to a trough. The maximum thickness of the overburden in this trough is approximately 60
. Borings 402 through 412 were drilled in this area to determine the subsurface conditions. A s-section of the trough along the discharge tunnel is presented on Figure 2.5.4-51. The tion of the section is shown on Figure 2.5.4-31. In this area, which extends for approximately feet, the fill and alluvium overlying the ablation and basal tills were excavated and replaced h crushed stone and concrete fill to the base elevation of the discharge tunnel. Because the tion till is a sandy material below the groundwater table, the liquefaction potential was lyzed. The analysis described in Section 2.5.4.8.4.1 shows that liquefaction of the ablation till ot possible under the site SSE. The structural fill and basal till have been shown to be liquefiable in Sections 2.5.4.8.1 and 2.5.4.8.2, respectively.
4.8.4.1 Dynamic Response Analysis of Ablation Till dynamic response of the ablation till has been evaluated to determine earthquake-induced ar stresses caused by ground motions applied at the bedrock surface and amplified through the profile. This evaluation was made using the computer program SHAKE, similar to the lysis in Section 2.5.4.8.3.1.
ountered in boring 411, which encountered the deepest rock, and represents the most servative profile in the study area. The generalized soil profile (Figure 2.5.4-50) used in the lysis of the tunnel consisted of 5 feet of structural fill, 13 feet of ablation till, and 22 feet of al till. Groundwater level was established at 10 feet below the ground surface, elevation +4
, based upon the average groundwater levels measured in borings 407 and 411. (See ure 2.5.4-31 for locations). The shear moduli values of the soils were obtained from cross-tests described in Section 2.5.4.4.3. The values of shear modulus (G) and damping (D) for strain levels used in the SHAKE analysis for each layer are:
Layer Elevation (ft) Depth (ft) Soil Type Gmax (ksf) Dmax (%)
1 +14 to -8 0-22 Discharge Tunnel -- 0.5 2 -8 to -13 22-27 Structural Fill 1.93 x 103 0.5 3 -13 to -26 27-40 Ablation Till 1.30 x 103 0.5 4 -26 to -48 40-62 Basal Till 2.0 x 104 0.5 reduction of Gmax with strain was performed through a series of iterations similar to the hod described in Section 2.5.4.8.3.1 using the same earthquake records normalized to 0.17g.
s analysis indicated that the average maximum shear stress in the ablation till induced by the
, varied from 515 psf to 533 psf. The average shear stress is assumed to be 0.65 of the peak e.
4.8.4.2 Liquefaction Analysis of Ablation Till cedures used for liquefaction analysis of the ablation till were similar to the empirical roach described in Section 2.5.4.8.3.2.
ndard penetration resistance data (N1 values) were related to liquefaction potential in ordance with methods developed by Seed, Arango, and Chan (1975) and DeAlba, Chan, and d (1975). N1 values for the ablation till were obtained from borings taken at the discharge nel location (400 series) and samples of ablation till from the main plant borings (300 series).
rage corrected blow count values and average N1 values less one standard deviation (N1-):
Midpoint of Induced Layer Shear Shear Shear Elevation Stress Strength Strength (ft) (psf) Mean N1 (psf) F.S. N1- (psf) F.S.
-15.2 515 28.7 1,079 2.10 15.5 578 1.12
-19.5 531 28.7 1,218 2.29 15.5 652 1.23
-23.9 533 28.7 1,357 2.60 15.5 726 1.36 an be concluded, therefore, that the ablation till under the discharge tunnel is not susceptible to efaction, even considering the ultraconservative case of the shear strength calculated from the n corrected blow count less one standard deviation.
4.9 EARTHQUAKE DESIGN BASIS afe shutdown earthquake of 0.17g and a 1/2 SSE value of 0.09g in the horizontal direction and
-thirds of these values in the vertical direction, input at the bedrock surface, have been used as design bases for seismic loading at the site. The derivation of these values is described in tions 2.5.2.6 and 2.5.2.7.
structures founded on soils, amplification effects have been considered by means of a soil-cture interaction analysis using the computer program PLAXLY-3 described in detail in tion 3.7.2.4.
the liquefaction analysis of the beach sands adjacent to the circulating and service water phouse, the SSE value of 0.17g was input at the bedrock surface, and the average amplified und motion at the surface determined from the SHAKE program using three earthquake rds and described in Section 2.5.4.8.3.1 was calculated to be 0.27g. Consequently, a value of g was conservatively used for the entire soil column as the average seismic loading of reline slopes in the stability analysis described in Section 2.5.5.2.
4.10 STATIC STABILITY 4.10.1 Bearing Capacity le 2.5.4-14 summarizes the bearing pressures for mats or individual spread footings founded arious foundation materials.
selection of the bearing capacity values used in footing design were based on the bearing acity formulae (Terzaghi and Peck 1967, Vesic 1975) for an estimated angle of internal friction basal till equal to 40 degrees and for structural backfill equal to 34 degrees. The total unit
les varying from 35 to 41.5 degrees for the structural fill compacted to 95 percent of the imum modified Proctor density. Inputting the relevant soil parameters described above, and ng into account the effect of the groundwater table, the bearing capacity formula for square ings or mats on basal till reduces to:
qall = 1.9 D + 1.1 B (2.5.4-6) qall (max) = 12 ksf structural backfill:
qall = 0.9 D + 0.4 B (2.5.4-7) qall (max) = 8 ksf re:
qall = Allowable bearing capacity in ksf with a minimum safety factor = 3 D = Depth of embedment (feet)
B = Width of footing (feet) le 2.5.4-24, Bearing Capacity of Major Structures, presents a summary of the allowable ring capacity for the material beneath each structure. In all cases, the factor of safety is greater 3, which is the minimum required value.
ed on Teng (1962), the design bearing capacity of foundations on rock is commonly taken as to 1/8 of the crushing strength (factor of safety of 5 to 8). A value of 200 ksf was selected for maximum allowable bearing capacity of bedrock at the site. This corresponds to roximately 1/7 of the average unconfined compressive strength of approximately 1,440 ksf 000 psi) reported in Table 2.5.4-10. The 200 ksf value also corresponds to the presumptive ace bearing value given by the Connecticut Basic Building Code (1978) for massive talline rock, including granite and gneiss.
m Table 2.5.4-14, the maximum average foundation pressure for a structure on rock is 8 ksf.
s, the factor of safety against a bearing capacity failure is much greater than 3 for all structures nded on rock.
4.10.2 Settlement of Structures k and soil supported Seismic Category I structures experience only elastic displacements er the design loads. Analyses using linear elasticity principals, assuming rigid foundations, cate that the vertical settlements of structures founded on rock are very small under the design s, as shown by the summary included in Table 2.5.4-14.
71). The main steam valve, auxiliary, and engineered safety features buildings, founded on k, were analyzed using equations for rigid rectangular mats on a semi-infinite mass developed Whitman and Richart (1967). Structures founded totally or partly on soil, such as the control, rgency diesel generator enclosure, fuel, and waste disposal buildings, were analyzed using tions obtained by Sovinc (1969) for rigid rectangles on a finite layer. The settlement of the erlying rock layer was also estimated using the Whitman and Richart equations.
stic properties of the rock and basal till are discussed in Section 2.5.4.4.3. The elastic modulus for static strain levels was estimated equal to 10,000 psi, as discussed in Section 2.5.4.5.2.
le 2.5.4-14 indicates that the maximum estimated settlement within any one structure occurs he emergency generator enclosure building and is equal to 0.40 inch. Most of this settlement lts from the conservative assumption that the south footing of the EGE is founded on 9 feet of ctural fill. Maximum estimated differential settlement between adjacent structures occurs ween the control building and the emergency generator enclosure building, and is equal to ut 0.40 inch. The rate of these settlements would essentially be the same as the rate of loading ause of elastic nature of the bearing material.
4.10.3 Lateral Earth Pressures magnitude and distribution of lateral earth pressures is a function of the allowable yielding of wall, the backfill material characteristics, water pressure, surcharge loads from adjacent ctures, and, for seismically designed structures, the earthquake loading. The concrete ndation walls were conservatively assumed to be rigid, unyielding walls. Therefore, the fficient of earth pressure at rest, Ko, has been used in evaluating lateral loads on these walls.
mpaction specifications prohibited the use of heavy vibratory compactors within 5 feet of all crete structures. Light compactors were used when backfilling against structures in order to imize residual lateral stresses in the fill due to the applied compactive effort. For the backfill e site, a value of Ko = 0.5 was used.
kfill placed behind walls consisted of well graded sands and gravels compacted to 90 percent maximum density (ASTM D1557) to minimize the horizontal loads induced by high pactive stresses. Tests on similar soils, compacted to 90 percent of maximum dry density and orted in Section 2.5.2.5.2, resulted in friction angles in excess of 34 degrees.
amic loadings include pressures due to the soil mass, water, and surcharge, accelerated in the ical and horizontal directions. Methods of analysis are based on procedures proposed by nonobe (1929), Okabe (1926), and Seed and Whitman (1970) and are graphically depicted on ure 2.5.4-43.
4.11 DESIGN CRITERIA design criteria and minimum required factors of safety for bearing capacity, hydrostatic uplift oyancy), and sliding (against lateral pressures) are summarized below:
Terzaghi and Peck 1967).
- 2. Hydrostatic Uplift - The determination of the buoyant force, and the weight of the structures can be made relatively precisely. Therefore, a factor of safety against hydrostatic uplifting of 1.1 was determined to be adequate under the normal water levels shown on Figure 2.5.4-37, considering the dead load of the structure only.
For high water level conditions, the dead load of the structure plus equipment load is at least 1.1 times the buoyant force of water.
- 3. Sliding - Either of two methods has been used to determine the factor of safety against sliding. One method considers that only frictional forces at the base resist sliding. A minimum factor of safety of 1.1 is required. The second method includes base friction and the resisting force due to passive earth pressure. For this method, a minimum factor of safety of 2.0 is required.
iscussion of the bearing capacity analysis is included in Section 2.5.4.10.1. The loads used for rmining lateral pressures on structures are discussed in Section 2.5.4.10.3.
design limits for the foundations of all Category I structures are discussed in the structural eptance criteria in Section 3.8.5.5.
4.12 TECHNIQUES TO IMPROVE SUBSURFACE CONDITIONS Category I structures are founded on either high quality, intact rock, undisturbed basal till, or pacted, select granular fill. Therefore, no improvement of the founding material below any cture was required.
k dowels were installed around the periphery of the auxiliary building to provide stability ng seismic loading. These dowels consist of 2 1/4-inch diameter, grade 60 steel bar with 10 s of fusion bonded epoxy coating for double corrosion protection. The dowels were designed ct as a passive support system, with stressing occurring only during seismic loading. Six test els of varying lengths were loaded to the yield strength of the bar (240 kips) to verify design meters.
k anchors were installed in the turbine building to provide resistance to overturning due to ado loading. These anchors consisted of a 1 1/4-inch diameter, high strength steel bar athed in a corrugated PVC casing and fully grouted for double corrosion protection. Each hor was proof loaded to 150 kips and then the load was reduced to 125 kips for 24 hours2.777778e-4 days <br />0.00667 hours <br />3.968254e-5 weeks <br />9.132e-6 months <br />. The hor was subsequently locked off at a permanent load of 25 kips and encased in the concrete ndation mat.
k anchors were installed in the service building to provide resistance to uplift loads due to yant forces and seismic forces. These anchors consisted of 1 3/8-in diameter, high strength l bar sheathed in a corrugated PVC casing and fully grouted for double corrosion protection.
ultimate strength of each anchor is 237 kips, with a working load equal to 60 percent of the
ed off at a load of 40 kips, which corresponds to the hydrostatic uplift component of the hor design load. The remaining capacity of the anchor is mobilized during seismic loading.
porary rock bolts were installed in the southwest sector of the containment excavation face to vent potential sliding failures along the foliation planes. These bolts consisted of Grade 60 l, No. 11 reinforcing bars with a working load of 45 kips. Anchorage of the rock bolts was vided by Celtite polyester resin encapsulation.
ailed geologic mapping of bedrock surfaces at the site, described in detail in Section 2.5.4.11, tified certain preferred joint surfaces that may cause potential sliding planes with the tainment excavation face. As a result of these findings, a reinforced concrete ring beam was ed in the annular space between the excavation face and the containment exterior wall to ilize the wedges. The slope stability analysis for the containment excavation is discussed in il in Section 2.5.5.1. The structural analysis is discussed in Chapter 3.
4.13 STRUCTURE SETTLEMENT st of the Category I structures at the site are founded on sound bedrock. Predicted settlements d in Table 2.5.4-14 for these structures are very small. Settlement predictions for structures nded on basal till or structural backfill indicate that the maximum expected settlement is less 0.4 inch and that this settlement occurs over a relatively short period of time due to the elastic re of the subsurface materials. Settlement has been monitored for the control, fuel, waste osal, and emergency generator enclosure buildings during construction. A plan of the location he settlement monitoring benchmark locations is shown on Figure 2.5.4-59. Plots of observed y settlement versus time for these structures are presented in Figures 2.5.4-60 through 2.5.4-The records show no significant movement of any structure, although some heave has urred due to rebound from excavation. Settlement of these structures has been periodically sured, and it has been determined that there does not appear to be any significant movement he monitoring points. Records of these measurements are being maintained, in accordance h Procedure No. SP-CE-223, as permanent plant records.
4.14 CONSTRUCTION NOTES significant problems were encountered during construction that required extensive redesign of ctures. A small amount of basal till was excavated and replaced with structural backfill eath the control building due to inflow of groundwater during excavation. This occurrence is ussed in detail in Section 2.5.4.5.1.
concrete backfill in the annular space between the containment exterior wall and the avation face was modified because of data obtained from the geologic mapping program. The crete backfill was revised to be a reinforced concrete structural support to resist the potential ure of rock wedges subjected to seismic loading and maintain the isolation of the containment cture from external forces. This ring beam is discussed in detail in Sections 2.5.5.1 and 5.3.
4.1-1 Bechtel Corporation, 1969. Subsurface, Geophysical, and Groundwater Investigations. In: Preliminary Safety Analysis Report, Millstone Nuclear Power Station - Unit 2.
4.1-2 Butterfield, R. and Banjerjee, P. K. 1971. A Rigid Disc Embedded in an Elastic Half Space. Geotechnical Engineering, Vol 2, No. 1, p 35-52.
4.1-3 Dale, T. N. 1923. The Commercial Granites of New England. U.S. Geol. Survey, Bulletin 738, p 448.
4.1-4 Dale, T. N. and Gregory, H. E. 1911. The Granites of Connecticut. U.S. Geol.
Survey, Bulletin 484, p 109-122.
4.1-5 DeAlba, P.; Chan, C. K.; and Seed, H. B. 1975. Determination of Soil Liquefaction Characteristics by Large-Scale Laboratory Tests. Earthquake Engineering Research Center, Report No. EERC 75-14, University of California, Berkeley, Calif.
4.1-6 Ebasco Services, Inc. 1966. Geology and Seismology. In: Design Analysis Report.
Millstone Nuclear Power Station - Unit No. 1, Section 5.0.
4.1-7 Gibbs, H. J. and Holtz W. G. 1957. Research on Determining the Density of Sands by Spoon Penetration Testing. Proceedings, Fourth International Conference on Soil Mechanics and Foundation Engineering, Vol 1, London, England, p 35-39.
4.1-8 Gutenberg, B. and Richter, C. F. 1942. Earthquake Magnitude, Intensity, Energy, and Acceleration. Seismological Society of American Bulletin, Vol 32, p 163-191.
4.1-9 Hardin, B. O. and Black, W. L. 1968. Vibration Modulus of Normally Consolidated Clay. Journal of the Soil Mechanics and Foundation Division, ASCE, Vol 94, No.
SM2.
4.1-10 Hardin, B. O. and Richart, F. E., Jr. 1963. Elastic Wave Velocities in Granular Soils.
Journal of the Soil Mechanics and Foundation Engineering Division, ASCE, Vol 89, SM1.
4.1-11 Kishida, H. 1966. Damage to Reinforced Concrete Buildings in Niigata City with Special Reference to Foundation Engineering. Soils and Foundations, Vol VI, No. 1, p 71-88.
4.1-12 Kishida, H. 1969. Characteristics of Liquefied Sand during Mino-Owari, Tohnankai, and Fuklai Earthquakes. Soils and Foundations, Vol IX, No. 1, p 75-92.
4.1-13 Koizumi, Y. 1966. Change in Density of Sand Subsoil Caused by the Niigata Earthquake Records. Transactions of the Architectural Institute of Japan,
4.1-14 LaPlante, F. and Christian, J. T. 1974. Earthquake Response Analysis of Horizontally Layered Sites (SHAKE). Stone & Webster Engineering Corporation, Computer Department User's Manual, ST-211, Boston, Mass.
4.1-15 Lee K. L. and Fitton, J. A. 1969. Factors Affecting the Cyclic Loading Strength of Soil. In: Vibration Effects of Earthquakes on Soils and Foundations. ASTM STP450, American Society of Testing and Materials.
4.1-16 Lee, K. L. and Seed, H. B. 1967. Cyclic Stress Conditions Causing Liquefaction of Sand. Journal of the Soil Mechanics and Foundations Engineering Division. ASCE, Vol 93, No. SM1.
4.1-17 Mononobe, N. 1929. Earthquake-Proof Construction of Masonry Dams.
Proceedings, World Engineering Conference, Vol 9, p 275.
4.1-18 Niles, W. H. 1975-76. Geological Agency of Lateral Pressure Exhibited by Certain Movements of Rocks. Proc. Boston Soc. of Natural History, Vol 18, p 279.
4.1-19 Northeast Nuclear Energy Company (NNECo.) 1975. Geologic Mapping of Bedrock Surface. Millstone Nuclear Power Station-Unit 3, Docket No. 50-423, Waterford, Conn.
4.1-20 Northeast Nuclear Energy Company (NNECo.) 1976. Report on Small Fault in Warehouse No. 5 - Unit 2 Condensate Polishing Facility. Millstone Nuclear Power Station-Unit 3, Docket No. 50-423, Waterford, Conn.
4.1-21 Northeast Nuclear Energy Company (NNECo.) 1977. Fault in Demineralized and Refueling Water Tank Area. Millstone Nuclear Power Station-Unit 3, Docket No.
50-423, Waterford, Conn.
4.1-22 Okabe, S. 1926. General Theory of Earth Pressure. Journal, Japanese Society of Civil Engineers, Vol 12, No. 1.
4.1-23 Seed, H. B. 1968. Landslides during Earthquakes Due to Liquefaction. Journal of the Soil Mechanics and Foundation Engineering Division, ASCE, Vol 94, No. SM5.
4.1-24 Seed, H. B.; Arrango, I.; and Chan, C. K. 1975. Evaluation of Soil Liquefaction Potential during Earthquakes. Earthquake Engineering Research Center, Report No.
EERC 75-28, University of California, Berkeley, Calif.
4.1-25 Seed, H. B. and Idriss, I. M. 1967. Analysis of Soil Liquefaction: Niigata Earthquake. Journal of the Soil Mechanics and Foundation Engineering Division, ASCE, Vol 93, No. SM3.
ASCE, Vol 97, No. SM9.
4.1-27 Seed, H.B.; Idriss, I.M.; Makdisi, F. and Banjeree, N.R. 1975. Representation of Irregular Stress Time Histories by Equivalent Uniform Stress Series in Liquefaction Analysis. Earthquake Engineering Research Center, Report No. EERC 75-29; University of California, Berkeley, Calif.
4.1-28 Seed, H. B. and Lee, K. L. 1966. Liquefaction of Saturated Sands during Cyclic Loading. Journal of the Soil Mechanics and Foundation Engineering Division.
ASCE, Vol 92, No. SM6.
4.1-29 Seed, H. B. and Whitman, R. V. 1970. Design of Earth Retaining Structures for Dynamic Loading. Proceedings, ASCE Specialty Conference on Lateral Stresses and Design of Earth Retaining Structures, Cornell University, Ithaca, NY.
4.1-30 Shannon and Wilson Inc. and Agbabian-Jacobsen Associates. 1972. Soil Behavior Under Earthquake Loading Conditions. Report prepared by Union Carbide Corporation for U.S. Atomic Energy Commission.
4.1-31 Sovinc, I. 1969. Displacements and Inclinations of Rigid Footings Resting on a Limited Elastic Layer of Uniform Thickness. Proceedings of the Seventh International Conference on Soil Mechanics and Foundation Engineering, Vol 1, p 385-389.
4.1-32 State of Connecticut, Basic Building Code, Seventh Edition, 1978. Building Officials and Code Administrators International, Inc. Homewood, Illinois.
4.1-33 Teng, W.C. 1962. Foundation Design, Prentice Hall, Inc., Englewood Cliffs, New Jersey.
4.1-34 Terzaghi, K. and Peck, R. 1967. Soil Mechanics in Engineering Practice. Second Edition, John Wiley and Sons, Inc., New York, NY.
4.1-35 Vesic, A. S. 1975. Bearing Capacity of Shallow Foundations. In: Foundation Engineering Handbook. Wenterkon, H. F. and Fang, H. Y. (ed.) Van Nostrand-Reinhold, New York, NY.
4.1-36 Whitman, R. V. and Richart, F. E., Jr. 1967. Design Procedures for Dynamically Loaded Foundations. Journal of the Soil Mechanics and Foundation Engineering Division, ASCE, Vol 93, No. SM6.
TABLE 2.5.4-1 LIST OF JOINTS - FINAL GRADE FLOORS OF STRUCTURES CLICK HERE TO SEE TABLE 2.5.4-1
BLE 2.5.4-2 LIST OF FOLIATIONS - FINAL GRADE FLOORS OF STRUCTURES CLICK HERE TO SEE TABLE 2.5.4-2
TABLE 2.5.4-3 LIST OF SLICKENSIDES - FINAL GRADE FLOORS OF STRUCTURES CLICK HERE TO SEE TABLE 2.5.4-3
TABLE 2.5.4-4 LIST OF JOINTS - FINAL GRADE CONTAINMENT AND ENGINEERED SAFETY FEATURES BUILDING WALLS CLICK HERE TO SEE TABLE 2.5.4-4
TABLE 2.5.4-5 LIST OF FOLIATIONS - FINAL GRADE CONTAINMENT AND ENGINEERED SAFETY FEATURES BUILDING WALLS CLICK HERE TO SEE TABLE 2.5.4-5
ABLE 2.5.4-6 LIST OF SLICKENSIDES - FINAL GRADE CONTAINMENT AND ENGINEERED SAFETY FEATURES BUILDING WALLS CLICK HERE TO SEE TABLE 2.5.4-6
TABLE 2.5.4-7 LIST OF JOINTS - FINAL GRADE WALLS OF STRUCTURES CLICK HERE TO SEE TABLE 2.5.4-7
ABLE 2.5.4-8 LIST OF FOLIATIONS - FINAL GRADE WALLS OF STRUCTURES CLICK HERE TO SEE TABLE 2.5.4-8
TABLE 2.5.4-9 LIST OF SLICKENSIDES - FINAL GRADE WALLS OF STRUCTURES CLICK HERE TO SEE TABLE 2.5.4-9
TABLE 2.5.4-10 ROCK COMPRESSION TEST RESULTS CLICK HERE TO SEE TABLE 2.5.4-10
ABLE 2.5.4-11 DIRECT SHEAR TEST RESULTS FROM JOINT AND FOLIATING SURFACES CLICK HERE TO SEE TABLE 2.5.4-11
Boring No. Sample Depth (ft) Elev (ft msl) t (lb/cu ft) Water Content (%) d (lb/cu ft) Gs Su (ksf) f (de P3 UP4A2 32.0 -18.65 117.2 32.9 88.2 2.75 1.9 33 P4 UP1A2 5.0 -01.13 114.9 29.8 88.5 2.76 1.6 41 P7 UP3A2 29.6 -25.80 116.3 32.9 87.5 2.77 2.2 35 P8 UP1A 12.0 -01.48 122.1 12.3 108.7 - 1.3 27
- Table 3 in Appendix 2.5F gives data on dynamic properties.
TABLE 2.5.4-13 NATURAL WATER CONTENTS OF SPLIT SPOON SAMPLES CLICK HERE TO SEE TABLE 2.5.4-13
Maxi Average Calcu Static Average Thickness Design Sta Bearing Founding Founding Thickness Till Structural Dimensions of Groundwater Settle Structure Load (psf) Grade (ft) Material (ft) Fill (ft) Foundation (ft) Elevation (ft) (in Containment 7,480 -38.7 Rock - - 158 diameter 21 0.04 Main Steam 5,000 +9.0 Rock - - 70 x 60 19 0.01 Valve Auxiliary 4,860 -0.5 Rock - - 177 x 102 23 0.02 Engineered 3,050 -0.5 Rock - - 139 x 47 21 0.01 Safety Features Control 3,810 -0.5 Till 0 to 10 - 120 x 103 19 0.02 to Emergency 3,070 +9.0 Till 10 10 Strip 19 0.01 to Generator Enclosure (EGE)
Emergency 1,230 +1.5 Till 10 4 65 x 32 19 less th Generator Oil 0.01 Tank Emergency 1,500 +18.50 Structural 17 9.5 44 x 12 19 0.25 Generator Backfill Mats Refueling 4,000 +15.0 Rock - - Octagon 64 - less th Water Storage inside diameter 0.01 Tank
Maxi Average Calcu Static Average Thickness Design Sta Bearing Founding Founding Thickness Till Structural Dimensions of Groundwater Settle Structure Load (psf) Grade (ft) Material (ft) Fill (ft) Foundation (ft) Elevation (ft) (in Demineralize 4,000 +14.5 Rock - - Octagon 40 - less th d Water inside diameter 0.01 Storage Tank Fuel 4,500 +3.0 Rock - - 93 x 112 23 less th 0.01 Waste 4,500 +0.5 Till 2 to 8 114 x 48 23 0.02 Disposal (Liquid)
Waste 3,030 +19.5 Structural 23 7 114 x 38 23 0.25 Disposal Backfill (Solid)
Hydrogen 4,490 +20.0 Concrete Fill - - 56 x 50 18 less th Recombiner 0.01 NOTE: All foundations are structural mat except EGE which is strip footing and slab on grade.
ELEVATIONS *
(SEE BOTH LOGS (Appendix 2.5J) FOR EXACT VALUES)
(Changes made to this table in 1997 correct transcription errors associated with the original submission.)
Site Coordinates Surface Groundwater Elevation Depth Drilled Top of Rock Boring No. North/South East/West (ft) (ft) Elevation (ft) Elevation (ft) Dat 301 N1562 E505 28.4 36.0 12.4 20.4 12-15-71 302 N1458 E421 28.2 75.0 15.7 17.7 12-22-71 303 N1635 E335 29.3 30.0 19.3 22.8 12-18-71 304 N1533 E371 28.5 75.0 14.5 19.5 12-21-71 305 N1448 E347 26.2 125.0 7.2 17.0 12-21-71 306 N1383 E346 24.1 75.0 10.6 12.6 12-15-71 307 N1558 E258 28.4 37.5 18.4 19.2 12-13-71 308 N1446 E267 24.8 75.0 14.8 18.0 12-17-71 309 N1632 E167 27.3 48.0 -0.7 17.3 12-13-71 310 N1433 E185 25.6 45.0 1.6 17.0 12-24-71 311 N1232 E213 20.9 50.0 -9.1 8.6 12-21-71 312 N1544 E123 25.9 65.0 -20.1 15.9 12-29-71 313 N1388 E113 22.7 66.5 -23.8 15.9 12-28-71 314 N1628 E062 24.9 53.0 -8.1 15.9 12-14-71 315 N1433 E076 23.3 63.0 -19.7 14.2 12-23-71 316 N1270 W010 15.6 53.0 -17.4 6.4 12-30-71
(SEE BOTH LOGS (Appendix 2.5J) FOR EXACT VALUES)
(Changes made to this table in 1997 correct transcription errors associated with the original submission.)
Site Coordinates Surface Groundwater Elevation Depth Drilled Top of Rock Boring No. North/South East/West (ft) (ft) Elevation (ft) Elevation (ft) Dat 317 N1446 E322 25.3 74.5 15.8 16.6 12-27-71 318 N1415 E115 25.0 67.0 -22.5 16.3 01-12-72 319 N1708 E065 25.0 79.3 -10.2 16.6 11-11-72 320 N1705 E183 28.3 26.0 12.5 20.6 11-02-72 321 N1808 E174 27.6 71.1 -19.0 13.1 11-03-72 322 N1808 E264 30.3 35.0 6.3 22.2 10-31-72 323 N1708 E264 29.3 32.0 12.3 21.1 10-30-72 324 N1718 E364 30.1 51.3 -4.9 21.6 10-26-72 325 N1718 E476 30.1 47.0 -6.7 21.3 10-18-72 326 N1432 E476 27.0 85.3 17.9 16.5 10-10-72 327
- N1510 E456 28.7 111.3 20.8 20.8 11-02-72 328
- N1593 E384 29.1 144.3 10.5 16.4 10-12-72 329
- N1520 E369 28.1 120.7 13.2 15.3 10-26-72 330
- N1521 E309 28.2 106.0 19.4 20.8 11-15-72 331
- N1460 E375 27.4 112.5 9.0 18.3 11-07-72 I-1 N1720 W980 18.1 31.5 7.6 6.6 12-31-71 I-2 N1520 W450 9.3 68.5 -38.2 1.3 01-14-72
(SEE BOTH LOGS (Appendix 2.5J) FOR EXACT VALUES)
(Changes made to this table in 1997 correct transcription errors associated with the original submission.)
Site Coordinates Surface Groundwater Elevation Depth Drilled Top of Rock Boring No. North/South East/West (ft) (ft) Elevation (ft) Elevation (ft) Dat I-3 N1097 W370 0.6 48.4 -34.2 -1.2 11-17-72 I-4 N1260 W200 16.5 50.3 -9.0 1.1 11-15-72 I-5 N1470 W20 20.1 75.0 -22.9 7.8 11-08-72 I-6 N1529 W150 20.3 46.5 -11.2 6.3 09-10-73 I-7 N1388 W205 18.2 47.0 -13.9 3.2 09-06-73 I-8 N1257 W310 14.7 58.5 -28.8 0.6 09-07-73 I-8A N1258 W307 14.8 18.0 I-9 N1225 W538 7.4 55.5 -30.6 I-10 N1148 W409 5.0 60.9 -40.9 1.7 09-10-73 I-10A N1121 W368 5.4 19.2 I-11 N1073 W279 7.1 60.0 -10.9 1.1 09-12-73 I-12 N0989 W171 9.0 29.0 -5.0 2.5 09-14-73 I-14 N1020 W485 -4.5 36.8 -41.3 Offshore I-15 N0948 W394 -4.6 51.8 -40.6 Offshore I-19A N0837 W272 -12.0 22.4 -19.4 Offshore I-20 N0970 W692 -10.8 55.1 -52.9 Offshore I-21 N0902 W574 -11.0 61.0 -57.0 Offshore
(SEE BOTH LOGS (Appendix 2.5J) FOR EXACT VALUES)
(Changes made to this table in 1997 correct transcription errors associated with the original submission.)
Site Coordinates Surface Groundwater Elevation Depth Drilled Top of Rock Boring No. North/South East/West (ft) (ft) Elevation (ft) Elevation (ft) Dat I-22 N0814 W433 -20.0 45.3 -56.3 Offshore I-23 N0836 W766 -14.1 22.5 -27.1 Offshore I-24 N0708 W679 -17.7 38.2 -40.9 Offshore DT-1 N0980 E580 15.1 44.5 -9.4 5.0 12-30-71 DT-2 N0250 E1040 9.8 150.0 1.8 3.3 01-11-72 DT-3 S0690 E1360 10.6 150.0 3.6 3.6 01-07-72 401 N1303 E487 21.9 35.3 0.2 12.2 06-04-74 402 N0950 E840 16.4 28.0 -0.1 8.9 06-04-74 403 N0710 E925 12.8 60.7 -33.7 404 N0388 E924 11.7 36.2 -4.4 405 N0301 E720 17.7 42.6 9.9 13.0 05-29-74 406 N800 E924 14.03 44.5 -20.5 6.03 04-30-80 407 N625 E924 14.94 65.0 -45.1 4.94 04-30-80 408 N788.12 E879.83 14.36 45.0 -29.1 - -
409 N782.27 E870.02 14.54 39.0 -19.5 5.54 05-02-80 410 N775.21 E860.03 14.36 39.0 -19.6 - -
411 N519.2 E922.6 13.94 64.5 -47.6 3.44 05-08-80
(SEE BOTH LOGS (Appendix 2.5J) FOR EXACT VALUES)
(Changes made to this table in 1997 correct transcription errors associated with the original submission.)
Site Coordinates Surface Groundwater Elevation Depth Drilled Top of Rock Boring No. North/South East/West (ft) (ft) Elevation (ft) Elevation (ft) Dat 402 N463.6 E921.2 13.34 57.0 -35.2 - -
P-1 N1037 W313 1.4 28.7 -22.3 P-2 N1109 W415 3.0 48.8 -41.0 P-3 N1188 W383 13.4 58.6 -40.3 P-4 N1046 W270 3.9 19.5 -10.6 P-5 N1010 W218 5.6 16.0 -5.4 P-6 N1068 W207 17.7 28.0 -5.3 P-7 N1165 W495 3.8 43.0 -34.2 P-8 N1315 W705 10.5 45.5 -30.0 P-9 N1292 W825 7.2 12.5 -0.3 P-10 N1254 W770 4.4 21.7 -12.3 T-1 N1287 E428 21.1 31.5 -1.4 T-2 N1234 E465 19.2 31.0 -3.3 T-3 N1234 E428 19.8 38.0 -8.2 T-4 N1159 E388 16.9 38.7 -16.8 T-5 N1208 E308 17.9 36.0 -4.1 T-6 N1250 E309 20.2 27.0 -1.8
(SEE BOTH LOGS (Appendix 2.5J) FOR EXACT VALUES)
(Changes made to this table in 1997 correct transcription errors associated with the original submission.)
Site Coordinates Surface Groundwater Elevation Depth Drilled Top of Rock Boring No. North/South East/West (ft) (ft) Elevation (ft) Elevation (ft) Dat T-7 N1186 E220 20.6 42.0 -7.4 Q-1 N0097 E323 2.7 50.0 -36.3-42.8 Offshore Q-2 N0095 E294 3.3 55.0 -46.7 Offshore Q-3 N0009 E330 -5.0 33.0 -33.0 Offshore Q-4 N0002 E299 -5.8 42.0 -42.8 Offshore B-1 N1175 E061 19.3 34.5 -8.2 B-2 N1172 E60.5 19.8 50.0 -25.2 B-3 N1109 E152 14.8 33.5 -13.7 B-4 N1059 E152 16.9 41.0 -17.6 B-5 N1012 E120 16.6 21.5 0.5 B-6 N1066 E061 17.0 21.0 2.5 NOTE:
- 45 degree angle boring
TABLE 2.5.4-16
SUMMARY
OF WATER PRESSURE TEST DATA CLICK HERE TO SEE TABLE 2.5.4-16
TABLE 2.5.4-17 GROUNDWATER OBSERVATIONS CLICK HERE TO SEE TABLE 2.5.4-17
TABLE 2.5.4-18 FACTORS OF SAFETY AGAINST LIQUEFACTION OF BEACH SANDS CLICK HERE TO SEE TABLE 2.5.4-18
TABLE 2.5.4-19 IN-PLACE DENSITY TEST RESULTS ON CATEGORY I STRUCTURAL BACKFILL BENEATH THE SERVICE WATER INTAKE PIPE ENCASEMENT CLICK HERE TO SEE TABLE 2.5.4-19
TABLE 2.5.4-20 IN-PLACE DENSITY TEST RESULTS AT CONTROL AND EMERGENCY GENERATOR ENCLOSURE BUILDINGS CLICK HERE TO SEE TABLE 2.5.4-20
ANALYSIS Taft Helena Parkfield Top of Layer Elevatio G (2)
Layer n (ft) Soil Type t (pcf) Gmax (psi) G (1 ) (psi) G (2) (ksf) D (3) G (2) (ksf) D (3) (ksf) D 1 24 Fill 143 1.2 X 104 3810 613 0.087 457 0.140 576 0.10 2 15 Fill 143 1.63 X 104 5324 778 0.104 825 0.116 697 0.12 3 (4) 10 Basal till 145 1.4 x 105 1.28 x 105 18,800 0.014 18,715 0.014 17,968 0.01 4 0 Basal till 145 1.4 x 105 1.20 x 105 17,850 0.018 17,249 0.022 16,913 0.02 5 -10 Basal till 145 1.4 x 105 1.12 x 105 16,981 0.022 16,077 0.025 15,470 0.02 6 -20 Bedrock - 1.5 x 106 -
NOTE:
- 1. Gg = The average of the G's for the 3 earthquakes.
- 2. G = Strain corrected shear modulus (ksf).
- 3. D = Strain corrected damping ratio.
- 4. The structure is founded on the basal till at the top of Layer 3. Soil stiffness was modeled using shear modulus.
ABLE 2.5.4-22B EMERGENCY GENERATOR ENCLOSURE - SOIL PROPERTIES WITH STRUCTURE EFFECTS FROM SHAKE ANALYSIS CLICK HERE TO SEE TABLE 2.5.4-21B
ABLE 2.5.4-23 EMERGENCY GENERATOR ENCLOSURE - SOIL PROPERTIES WITH STRUCTURE EFFECTS FROM SHAKE ANALYSIS CLICK HERE TO SEE TABLE 2.5.4-22
Approximate Allowable 1 Dimensions of Approximate Approximate Ultimate Bearing Contact Area Foundation Foundation Static Bearing Capacity Facto Structure (ft) Depth (ft) Load (ksf) Capacity (ksf) (ksf) Saf Containment 158 diam 62.7 7.48 - 200 > 26 Main Steam Valve 70 x 60 15.0 5 - 200 > 40 Auxiliary 177 x 102 24.5 4.86 - 200 > 41 Engineered Safety Features 139 x 47 24.5 3.05 - 200 > 65 Control 120 x 103 24.5 3.81 488 12 > 128 Emergency Generator Enclosure 10 strip 15.0 3.07 128 12 > 41 Emergency Generator Oil Tank 65 x 32 22.5 1.23 102 8 > 83 Refueling Water Storage Tank Octagon 64 I.D. 9.0 4 - 200 > 50 Demineralized Water Storage Tank Octagon 40 I.D. 9.5 4 - 200 > 50 Fuel 93 x 112 21.0 4.5 - 200 > 44 Waste Disposal (Liquid) 114 x 48 23.5 4.5 288 12 > 64 Hydrogen Recombiner 56 x 50 4.0 4.49 - 200 > 44 Discharge Tunnel 17 wide 32.5 to 22.5 3.59 56 2 8 15 Circulating Water Pumphouse 142 x 84 46.0 4.55 - 200 > 44 NOTES:
- 1. See FSAR Section 2.5.4.10
- 2. Bearing capacity determined for structure on concrete fill over ablation till
BLE 2.5.4-25 RESULTS OF TWO-DIMENSIONAL LIQUEFACTION ANALYSIS OF BEACH AREA SANDS CLICK HERE TO SEE TABLE 2.5.4-24
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-1 GEOLOGIC MAP OF FINAL GRADE, SERVICE WATER LINE WALLS - EAST Revision 3306/30/20 MPS-3 FSAR 2.5.4-67
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-2 GEOLOGIC MAP OF FINAL GRADE, SERVICE WATER LINE WALLS - WEST Revision 3306/30/20 MPS-3 FSAR 2.5.4-68
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-3 GEOLOGIC MAP OF FINAL GRADE, SOUTH WALL OF DISCHARGE TUNNEL Revision 3306/30/20 MPS-3 FSAR 2.5.4-69
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-4 GEOLOGIC MAP OF FINAL GRADE, NORTH WALL OF DISCHARGE TUNNEL Revision 3306/30/20 MPS-3 FSAR 2.5.4-70
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-5 GEOLOGIC MAP OF FINAL GRADE, EAST WALL OF DISCHARGE TUNNEL Revision 3306/30/20 MPS-3 FSAR 2.5.4-71
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-6 GEOLOGIC MAP OF FINAL GRADE, FLOORS OF STRUCTURES Revision 3306/30/20 MPS-3 FSAR 2.5.4-72
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-7 GEOLOGIC MAP OF FINAL GRADE, SERVICE WATER LINE FLOOR - WEST Revision 3306/30/20 MPS-3 FSAR 2.5.4-73
Revision 3306/30/20 MPS-3 FSAR 2.5.4-74 Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-8 GEOLOGIC MAP OF FINAL GRADE, PUMPHOUSE FLOOR
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-9 GEOLOGIC MAP OF FINAL GRADE, SERVICE WATER LINE FLOOR - EAST Revision 3306/30/20 MPS-3 FSAR 2.5.4-75
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-10 GEOLOGIC MAP OF FINAL GRADE, SOUTHEAST QUADRANT OF CONTAINMENT WALLS Revision 3306/30/20 MPS-3 FSAR 2.5.4-76
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-11 GEOLOGIC MAP OF FINAL GRADE, SOUTHWEST QUADRANT OF CONTAINMENT WALLS Revision 3306/30/20 MPS-3 FSAR 2.5.4-77
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-12 GEOLOGIC MAP OF FINAL GRADE, NORTHWEST QUADRANT OF CONTAINMENT WALLS Revision 3306/30/20 MPS-3 FSAR 2.5.4-78
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-13 GEOLOGIC MAP OF FINAL GRADE, NORTHEAST QUADRANT OF CONTAINMENT WALLS Revision 3306/30/20 MPS-3 FSAR 2.5.4-79
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-14 GEOLOGIC MAP OF FINAL GRADE, ENGINEERED SAFETY FEATURES, BUILDING SUMP WALLS Revision 3306/30/20 MPS-3 FSAR 2.5.4-80
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-15 GEOLOGIC MAP OF FINAL GRADE, AUXILIARY BUILDING PIPE TUNNEL PIT WALLS Revision 3306/30/20 MPS-3 FSAR 2.5.4-81
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-16 GEOLOGIC MAP OF FINAL GRADE, NORTH WALL OF EXCAVATION Revision 3306/30/20 MPS-3 FSAR 2.5.4-82
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-17 GEOLOGIC MAP OF FINAL GRADE, NORTHEAST AND SOUTHEAST PUMPHOUSE WALLS Revision 3306/30/20 MPS-3 FSAR 2.5.4-83
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-18 GEOLOGIC MAP OF FINAL GRADE ENGINEERED SAFETY FEATURES BUILDING WALL Revision 3306/30/20 MPS-3 FSAR 2.5.4-84
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-19 GEOLOGIC MAP OF FINAL GRADE DISCHARGE TUNNEL FLOOR Revision 3306/30/20 MPS-3 FSAR 2.5.4-85
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-20 GEOLOGICAL MAP OF FINAL GRADE DISCHARGE TUNNEL FLOOR Revision 3306/30/20 MPS-3 FSAR 2.5.4-86
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-21 GEOLOGICAL MAP OF FINAL GRADE NORTH WALL OF DISCHARGE TUNNEL Revision 3306/30/20 MPS-3 FSAR 2.5.4-87
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-22 GEOLOGICAL MAP OF FINAL GRADE SOUTH WALL OF DISCHARGE TUNNEL Revision 3306/30/20 MPS-3 FSAR 2.5.4-88
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-23 GEOLOGIC MAP OF FINAL GRADE DISCHARGE TUNNEL FLOOR Revision 3306/30/20 MPS-3 FSAR 2.5.4-89
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-24 GEOLOGIC MAP OF FINAL GRADE DISCHARGE TUNNEL FLOOR Revision 3306/30/20 MPS-3 FSAR 2.5.4-90
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-25 GEOLOGIC MAP OF FINAL GRADE WEST WALL OF DISCHARGE TUNNEL Revision 3306/30/20 MPS-3 FSAR 2.5.4-91
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-26 GEOLOGIC MAP OF FINAL GRADE EAST WALL OF DISCHARGE TUNNEL Revision 3306/30/20 MPS-3 FSAR 2.5.4-92
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-27 GEOLOGIC MAP OF FINAL GRADE DISCHARGE WEIR ROCK FACE Revision 3306/30/20 MPS-3 FSAR 2.5.4-93
Revision 3306/30/20 MPS-3 FSAR 2.5.4-94 FIGURE 2.5.4-28 CORRECTED BLOW COUNT PLOT, PUMPHOUSE AREA SANDS, ONSHORE BORING COMPOSITE
Revision 3306/30/20 MPS-3 FSAR 2.5.4-95 FIGURE 2.5.4-29 CORRECTED BLOW COUNT PLOT, PUMPHOUSE AREA SANDS, BORINGS P1 TO P8 COMPOSITE
M FIGURE 2.5.4-30 GRAIN SIZE DISTRIBUTION CURVES (SHEET 1)
Revision 3306/30/20 MPS-3 FSAR 2.5.4-96
FIGURE 2.5.4-30 GRAIN SIZE DISTRIBUTION CURVES (SHEET 2)
Revision 3306/30/20 MPS-3 FSAR 2.5.4-97
FIGURE 2.5.4-30 GRAIN SIZE DISTRIBUTION CURVES (SHEET 3)
Revision 3306/30/20 MPS-3 FSAR 2.5.4-98
FIGURE 2.5.4-30 GRAIN SIZE DISTRIBUTION CURVES (SHEET 4)
Revision 3306/30/20 MPS-3 FSAR 2.5.4-99
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-31 BORING LOCATION PLAN Revision 3306/30/20 MPS-3 FSAR 2.5.4-100
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-32 PLOT PLAN SHOWING LOCATIONS OF THE BORINGS AND THE GEOLOGIC SECTIONS Revision 3306/30/20 MPS-3 FSAR 2.5.4-101
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-33 GEOLOGIC PROFILE, SECTIONS A-A', B-B' Revision 3306/30/20 MPS-3 FSAR 2.5.4-102
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-34 GEOLOGIC PROFILE, SECTIONS C-C', D-D', E-E' Revision 3306/30/20 MPS-3 FSAR 2.5.4-103
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-35 GEOLOGIC PROFILE, SECTIONS F-F" AND G-G' Revision 3306/30/20 MPS-3 FSAR 2.5.4-104
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-36 TOP OF BASAL TILL CONTOUR MAP Revision 3306/30/20 MPS-3 FSAR 2.5.4-105
Revision 3306/30/20 MPS-3 FSAR 2.5.4-106 Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-37 GROUNDWATER CONTOUR MAP
Revision 3306/30/20 MPS-3 FSAR 2.5.4-107 FIGURE 2.5.4-38 GROUNDWATER OBSERVATIONS IN BOREHOLES
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-39 BEDROCK SURFACE CONTOUR MAP Revision 3306/30/20 MPS-3 FSAR 2.5.4-108
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-40 GENERAL EXCAVATION PLAN Revision 3306/30/20 MPS-3 FSAR 2.5.4-109
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-41 SHOREFRONT AND DREDGING PLAN Revision 3306/30/20 MPS-3 FSAR 2.5.4-110
Revision 3306/30/20 MPS-3 FSAR 2.5.4-111 FIGURE 2.5.4-42 MODULUS VS EFFECTIVE CONFINING PRESSURE, STRUCTURAL FILL
FIGURE 2.5.4-43 LATERAL PRESSURE DISTRIBUTION Revision 3306/30/20 MPS-3 FSAR 2.5.4-112
FIGURE 2.5.4-44 GRADATION CURVES FOR CATEGORY I STRUCTURAL FILL Revision 3306/30/20 MPS-3 FSAR 2.5.4-113
FIGURE 2.5.4-45 K2 VS SHEAR STRAIN FOR BEACH SANDS Revision 3306/30/20 MPS-3 FSAR 2.5.4-114
Revision 3306/30/20 MPS-3 FSAR 2.5.4-115 FIGURE 2.5.4-46 EARTHQUAKE INDUCED SHEAR STRESSES IN BEACH SANDS
Revision 3306/30/20 MPS-3 FSAR 2.5.4-116 FIGURE 2.5.4-47 CYCLIC STRESS RATIO VS CONFINING PRESSURE FOR BEACH SANDS
Revision 3306/30/20 MPS-3 FSAR 2.5.4-117 FIGURE 2.5.4-48 CYCLIC STRESS RATIO VS PENETRATION RESISTANCE OF SAND
Revision 3306/30/20 MPS-3 FSAR 2.5.4-118 FIGURE 2.5.4-49 FACTOR OF SAFETY AGAINST LIQUEFACTION OF BEACH SANDS
Revision 3306/30/20 MPS-3 FSAR 2.5.4-119 FIGURE 2.5.4-50 IDEALIZED SOIL PROFILE LIQUEFACTION ANALYSIS OF ABLATION TILL UNDER DISCHARGE TUNNEL
FIGURE 2.5.4-51 GEOLOGIC PROFILE, SECTION H-H
Revision 3306/30/20 MPS-3 FSAR 2.5.4-120
FIGURE 2.5.4-52 GEOLOGIC PROFILE, SECTION I-I
Revision 3306/30/20 MPS-3 FSAR 2.5.4-121
FIGURE 2.5.4-53 LOCATION OF FIELD DENSITY TESTS - SERVICE WATER INTAKE LINE Revision 3306/30/20 MPS-3 FSAR 2.5.4-122
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-54 LOCATION OF FIELD DENSITY TEST - EMERGENCY GENERATOR ENCLOSURE AND CONTROL BUILDING Revision 3306/30/20 MPS-3 FSAR 2.5.4-123
FIGURE 2.5.4-55 GEOLOGIC PROFILE, SECTION J-J' Revision 3306/30/20 MPS-3 FSAR 2.5.4-124
FIGURE 2.5.4-56 GEOLOGIC PROFILE, SECTION K-K' Revision 3306/30/20 MPS-3 FSAR 2.5.4-125
FIGURE 2.5.4-57 GRAIN SIZE DISTRIBUTION CURVES - PUMPHOUSE AREA OUTWASH SANDS (SHEET 1)
Revision 3306/30/20 MPS-3 FSAR 2.5.4-126
FIGURE 2.5.4-57 GRAIN SIZE DISTRIBUTION CURVES - PUMPHOUSE AREA OUTWASH SANDS (SHEET 2)
Revision 3306/30/20 MPS-3 FSAR 2.5.4-127
Revision 3306/30/20 MPS-3 FSAR 2.5.4-128 FIGURE 2.5.4-58 EQUIVALENT NUMBERS OF UNIFORM STRESS CYCLES BASED ON STRONGEST COMPONENTS OF GROUND MOTION
Withhold under 10 CFR 2.390 (d) (1)
FIGURE 2.5.4-59 PLAN OF SETTLEMENT MONITORING BENCHMARK LOCATIONS Revision 3306/30/20 MPS-3 FSAR 2.5.4-129
FIGURE 2.5.4-60 CONTROL BUILDING SETTLEMENT (SHEET 1)
Revision 3306/30/20 MPS-3 FSAR 2.5.4-130
FIGURE 2.5.4-60 CONTROL BUILDING SETTLEMENT (SHEET 2)
Revision 3306/30/20 MPS-3 FSAR 2.5.4-131
FIGURE 2.5.4-61 EMERGENCY GENERATOR ENCLOSURE SETTLEMENT Revision 3306/30/20 MPS-3 FSAR 2.5.4-132
FIGURE 2.5.4-62 .SOLID WASTE BUILDING SETTLEMENT Revision 3306/30/20 MPS-3 FSAR 2.5.4-133
FIGURE 2.5.4-63 LIQUID WASTE BUILDING SETTLEMENT Revision 3306/30/20 MPS-3 FSAR 2.5.4-134
FIGURE 2.5.4-64 FUEL BUILDING SETTLEMENT Revision 3306/30/20 MPS-3 FSAR 2.5.4-135
FIGURE 2.5.4-65 GEOLOGIC PROFILE SECTION L-L' Revision 3306/30/20 MPS-3 FSAR 2.5.4-136
FIGURE 2.5.4-66 GEOLOGIC PROFILE SECTION M-M' Revision 3306/30/20 MPS-3 FSAR 2.5.4-137
FIGURE 2.5.4-67 GEOLOGIC PROFILE SECTION N-N' Revision 3306/30/20 MPS-3 FSAR 2.5.4-138
FIGURE 2.5.4-68 GEOLOGIC PROFILE SECTION O-O' Revision 3306/30/20 MPS-3 FSAR 2.5.4-139
FIGURE 2.5.4-69 GEOLOGIC PROFILE SECTION P-P' Revision 3306/30/20 MPS-3 FSAR 2.5.4-140
FIGURE 2.5.4-70 GEOLOGIC PROFILE SECTION Q-Q' Revision 3306/30/20 MPS-3 FSAR 2.5.4-141
FIGURE 2.5.4-71 GEOLOGIC PROFILE SECTION R-R' Revision 3306/30/20 MPS-3 FSAR 2.5.4-142
Revision 3306/30/20 MPS-3 FSAR 2.5.4-143 FIGURE 2.5.4-72 SOIL-STRUCTURE INTERACTION EMERGENCY GENERATOR ENCLOSURE
Revision 3306/30/20 MPS-3 FSAR 2.5.4-144 FIGURE 2.5.4-73 SHEAR MODULUS CURVE TYPE 2 SOIL (STRUCTURAL BACKFILL AND BASAL TILL)
Revision 3306/30/20 MPS-3 FSAR 2.5.4-145 FIGURE 2.5.4-74 DAMPING CURVE TYPE 2 SOIL (STRUCTURAL BACKFILL AND BASAL TILL)
FIGURE 2.5.4-75 SHOREFRONT PROFILE USED IN LIQUEFACTION ANALYSES Revision 3306/30/20 MPS-3 FSAR 2.5.4-146
topography in the plant area is generally flat with the final grade at el +24 feet in the major t area. Detailed analyses were performed to determine static and dynamic stability of two
-made slopes at the site. The beach and outwash sand and armor stone slope at the shoreline, cent to the pumphouse, was analyzed using circular failure surfaces according to methods eloped by Bishop (1955) and incorporated in the LEASE II (SWEC 1977) computer program.
analysis showed that the slope (Figure 2.5.5-1) was safe under an average amplified seismic ing of 0.25g and under static conditions.
vertical rock cut excavated for the containment structure, (Figure 2.5.5-2) was analyzed by ming failure planes developed along fully continuous joint and foliation surfaces. Methods eloped by Hendron et al. (1971) and incorporated into the computer program SWARS-2P EC 1974b) were used to analyze slope stability of this rock cut under both static and dynamic ing conditions. It was determined that the stability of these slopes was inadequate to maintain ation of the containment structure walls from the external load applied by the rock wedges. As sult, a continuous structural hoop or ring beam was constructed in the annular space between containment walls and the rock face. The purpose of this ring beam is to transfer any rock ing from the less stable areas where potential failure wedges may form to the more stable s elsewhere around the containment, maintaining isolation of the rock loads from the tainment exterior walls. Section 3.8.1.1 discusses the details of the ring beam design.
5.1 SLOPE CHARACTERISTICS 5.1.1 Shoreline Slope lan of the shoreline in the vicinity of the Millstone 3 pumphouse is shown on Figure 2.5.4-41.
he east of the pumphouse, a reinforced concrete seawall with post-tensioned rock anchors has n built between the pumphouse and the Millstone 2 intake structure to retain the earth and ect the structures from wave action. On the west side of the pumphouse, extending in a herly direction, is a reinforced concrete retaining wall keyed into rock. The purpose of this l is to protect the circulating and service water lines from being undermined due to wave on on the adjoining slope. To the west of the pumphouse, a variable slope has been cut in the ch and outwash sand to provide for a transition from the offshore intake channel at el -32 feet he pumphouse area site grade at el +14 feet. The slope varies from five horizontal to one ical immediately adjacent to the pumphouse, to ten horizontal to one vertical near Bay Point, western extent of the beach. Compacted backfill was placed in areas where additional fill was uired to meet these grade requirements.
ultilayer stone armor zone was placed on the slope for protection against wave action during probable maximum hurricane. The maximum significant wave height of 13 feet was used to gn the slope protection system. The techniques used are described in the U.S. Army Shore tection Manual (USA CERC 1975).
armor layer is designed as a 2-layer system. The weight of stone was obtained from the ation:
W = ---------------------------------------------
3 K D ( S r - 1 ) cot re:
Wr = unit weight of the stone = 160 pcf H = design wave height = 13 ft Sr = specific gravity of rock with respect to seawater = 2.5 q = angle of slope = 5:1 = 11.3° KD = stability coefficient = 3.5 (assuming 2 layers of randomly placed rough angular quarrystone, a breaking wave, and a structure trunk with 0 to 5 percent armor layer damage).
individual stone weight of 6,000 pounds was obtained for the primary armor stone cover layer.
ange of 0.75W to 1.25W was allowed in specifications. Stone sizes, based on cubic shapes, ged from 3.0 to 3.6 feet.
thickness of the 2 armor layers was calculated as 7.7 feet. A minimum still water level of ation -3 feet was assumed in calculating the bottom of the primary armor layer at elevation feet. A secondary rock protection layer, referred to as Type B material, was calculated to be pound stone. Two layers of secondary protection, varying in size from 300 pounds to 700 nds with 75 percent greater than 500 pounds, with a minimum thickness of 3.6 feet, were rmined as necessary underlayment for the primary armor stone layer. This secondary layer placed on a continuous filter fabric layer.
in situ beach sands and compacted fill layers are overlain by filter fabric, which prevents ration of finer materials into the rock protection layers. Figure 2.5.5-1 shows a detailed cross-ion of the slope protection system. Stone for the slope protection was obtained from bedrock viously blasted during excavation at the site and from offsite sources.
ings P1 through P10 and I1 through I12 were drilled onshore in the vicinity of the pumphouse.
P-series of borings included undisturbed sampling of the saturated beach sands using the erberg sampler. The location of these borings is shown on Figure 2.5.4-31. A geologic profile ss this area is shown on Figure 2.5.4-35. The depth of sand along the beach varies from zero ay Point and the area just west of the Millstone 2 intake structure, where exposed rock is ent, to a maximum of approximately 40 feet in the vicinity of the Millstone 3 pumphouse.
beach and outwash sand deposits overlie a thin layer of basal till, generally less than 5 feet k which covers the bedrock. The relative density of the beach sand determined from the bs-Holtz correlation of blow count data averages approximately 70 percent. The data points plotted on Figure 2.5.4-28. No extensive or continuous loose zones were detected in these ngs.
reactor containment building is founded on bedrock at approximately el -39 feet. Top of rock es from approximately el 0 feet to el 20 feet, as shown on the bedrock surface contour map ure 2.5.4-39). The excavation walls are vertical, with a 9-inch bench at el -17 feet. The avation of bedrock for the containment is described in detail in Section 2.5.4.5.1.
a result of detailed geologic mapping of the bedrock surface during construction, described in tion 2.5.4.1.1, additional preferred joint sets were noted beyond those previously reported in local geological literature (Goldsmith 1967). These joint sets have been interpreted from the eonet projection plots for top of bedrock mapping previously reported (NNECo. 1975) and ted from final excavation grade data on Figure 2.5.1-16. Cross sections through the critical ges are shown on Figure 2.5.5-2.
assumption that failure surfaces develop in the bedrock along joint and foliation surfaces is y conservative. For this to occur, joint and foliation surfaces must be at least as long as the zontal projection of the wedge failure surfaces and must extend from the rock surface to a imum elevation of -27 feet, which corresponds to the top of the containment mat. There is no evidence that this situation does occur around the containment structure, particularly with ect to the minor joint sets. However, for the purposes of analysis, the joint and foliation planes modeled as flat, smooth continuous surfaces.
ect shear tests were conducted on samples of both foliation and joints to determine the tional values for each plane. A direct shear device capable of developing the low normal es representative of field conditions and sensitive enough to measure the shear force was used est NX core samples of rock from the borings previously taken in the vicinity of the tainment structure. A description of the samples tested is shown in Table 1 of Appendix 2.5I.
peak and residual values are plotted on Figure 2 of Appendix 2.5I. For analysis, a friction le of 32 degrees was used for the foliation, equals 34 degrees for the predominant joint set 04E and equals 37 degrees for the minor joint sets. These values do not take into account added strength of the asperities, which was significant for the higher normal stresses.
5.2 DESIGN CRITERIA AND ANALYSIS 5.2.1 Shoreline Slope omputer program, LEASE II (Limiting Equilibrium Analysis of Slopes and Embankments)
EC 1977), was used to analyze the stability of the shoreline slope. This program, part of ICES egrated Civil Engineering System--VI M3, dated November 1969), is accepted and widely d by soil mechanics and foundation engineers for analyzing slope stability problems. This ion is an update of an earlier version and provides for making dynamic analyses. LEASE II is ently being run on an IBM-370 Operating System, Model 165 at the SWEC Computer Center.
ASE II calculates three different factors of safety. The methods of analysis include the plified Bishop method of slices, the Fellenius method of slices, and the Rankine wedge hod. The simplified Bishop method of analysis was used to compute factors of safety of the
itman and Bailey indicate that the error involved in the simplified Bishop method is usually than 5 percent and, therefore, recommend it be used for slope stability analyses.
shoreline slope shown on Figure 2.5.4-41 was analyzed for static, dynamic, and post-hquake conditions. The slope to the west of the circulation and service water pumphouse is l-shaped. It is 5H to 1V at the steepest portion and decreases to 10H to 1V adjacent to the ke channel. Conservative assumptions were made in constructing the analytical model for the e stability analysis. The end constraints of three dimensional geometry were ignored and the re slope was assumed to be 5H:1V. A section through the modeled slope together with a mary of the results of the slope stability analysis is presented on Figure 2.5.5-1. Factors of ty are defined as the available shear resistance along a postulated failure surface divided by maximum driving forces along that surface.
ure 2.5.4-37 shows a groundwater gradient from the main plant area toward Long Island nd. Groundwater levels in boring 316, which is approximately 250 feet from the pumphouse, ed between elevations 6.4 and 8 feet. Water levels of elevation +6 feet onshore and -6 feet hore were conservatively selected to maximize the destabilizing forces in the analysis. These ls represent approximately four times the normal tidal range, and elevation +6 feet esponds to an appropriate flood tide level at the site.
strength properties used in the stability analysis were selected on the basis of standard etration tests and of cyclic triaxial and consolidated undrained (CIU) triaxial tests on isturbed samples, as reported in Appendixes 2.5G and 2.5F, respectively. The effect of sible pore pressure buildup in the beach and outwash sands was accounted for in the stability lysis for the post-earthquake conditions.
tatic slope stability analysis was conducted using the assumptions described above together h strengths for the various slope materials as shown on Figure 2.5.5-1. The effective internal tion angles assigned to the beach and glacial outwash sand were selected on the basis of dard penetration tests and of the CIU triaxial tests on undisturbed samples. The CIU tests (see endix 2.5G and Figure 2.5.5-5) revealed effective internal friction angles of 33 to more than degrees for the samples tested. For internal friction, an angle of 34 degrees was used in the lysis. The minimum factor of safety against slope failure for the static case is 2.9, which is quate. The dynamic slope stability during the SSE was evaluated by using a pseudo-static roach and undrained shear strengths of the soils. Input horizontal and vertical accelerations of 5g and 0.17g were based on the average amplified accelerations described in tions 2.5.4.8.3.1 and 2.5.4.9. Acceleration directions were selected to maximize instability.
rained strength parameters for the beach and glacial outwash sands were derived based on rained triaxial compression test results reported in Appendix 2.5G. Stress paths and data from e tests indicate that during undrained loading the average A parameter at maximum obliquity 0.13. The values of A ranged from +0.33 to -0.16. An A parameter equal to 0.5 and an rnal friction angle of 34° were used to derive the undrained strengths of the beach and glacial wash sands. These values are considered to be conservative based on the in situ density and
l of each layer.
ults of the pseudo-dynamic stability analysis indicate that the minimum factor of safety inst slope failure is 0.9 for the assumed conditions. This result is very conservative because of assumptions made about slope geometry and end effects. An additional analysis was ormed on this slope considering only the horizontal component of seismic loading, and a ty factor of 1.16 was calculated for the same failure circle.
to the low factor of safety obtained, an analysis was performed to estimate the deformations ch could theoretically occur along the postulated failure surface during earthquake loading.
analysis is based on an approach presented by Newmark (1965) using the computer program ES (Seismically Induced Displacement of Embankments and Slopes, SWEC, 1979) which ulates the cumulative mono-directional sliding displacement of a rigid body shaken by an hquake. An input earthquake accelerogram is represented by a maximum 12,000 point time ory of acceleration. No motion is assumed to occur within the slope until the strength of the is exceeded; i.e., the limiting acceleration producing a safety factor of 1.0 is exceeded.
lytical equations governing rigid solution are then solved incrementally on the assumption the input acceleration varies linearly from point to point, and that the displacements are ulative throughout the duration of the earthquake. Each of the three earthquakes used to pute the dynamic response of the soil were used (Section 2.5.4.8.3). Their time histories were ed to the appropriate average amplified accelerations (i.e., vertical acceleration of 0.17g and zontal acceleration of 0.25g) described above. Results from each of these earthquakes cate maximum cumulative slope movements less than 0.1 inch. The limiting horizontal and ical acceleration used were 0.2g and 0.12g, respectively. These results indicate that if there is movement of the slope during the SSE, the movement would be negligible. There would be dverse effect to any safety related system component or structures.
ost-earthquake stability analysis was performed to quantify the effect of pore pressure erated by the earthquake. The magnitude of pore pressure buildup was estimated from results yclic triaxial tests (Appendix 2.5F) considering such factors as the number of equivalent les, cyclic shear stress levels, confining pressures, material density, and gradation. The pore sure buildup for 5 cycles of loading on samples which most closely represent the in situ dition and dynamic loading was between 40 and 60 percent of the effective confining sure. An estimated pore pressure buildup of 50 percent was used to evaluate the post-hquake slope stability. Therefore, the soil properties are the same as in the static case but the e pressures are increased during the LEASE analysis. Results of the post-earthquake analyses al that the minimum factor of safety against slope failure is 1.4. This is considered acceptable.
analysis of slope stability indicates that the shoreline slope is stable under static, dynamic, post-earthquake conditions.
ddition to the above analysis, where the shorefront slope was considered to consist of a orm deposit of outwash sand to el -40 feet, the actual subsurface conditions were modeled to rmine whether a more critical cross-section existed due to sloping bedrock conditions at the refront. The actual soil profiled in this area is shown on Figure 2.5.4-52. This condition was
sloping rock condition used in the slope stability analysis showing soil properties and slope metry is shown on Figure 2.5.5-4.
minimum safety factor for static loading conditions is 3.2 for a circular arc failure surface.
s compares to a safety factor of 2.9 for the circular failure surface in the analysis for a uniform th of sand to el -40 feet indicating that the sloping rock profile is less critical than the uniform d profile. Also, the dynamic analysis for the sloping rock profile is less critical than the orm sand profile because the magnitude of the dynamic forces is reduced due to less lification through the stiffer till and because of the shallower depth to bedrock. The minimum ty factor for dynamic loading conditions is comparable with the uniform sand profile when ilar dynamic forces are assumed.
rgenstern-Price wedge failure analyses were also performed to further investigate the sloping k profile. Safety factors for static loading conditions of 4.14 for shallow wedge and 3.54 for per wedge failure surfaces were calculated. These safety factors against slope failure are her than the circular arc safety factor of 3.2 and confirm the inherent conservatism of the ular arc failure analysis.
uefaction of the shoreline slopes was also investigated. The analyses (Section 2.5.4.8.3) show the beach sands would not liquefy when subjected to the SSE.
5.2.2 Containment Rock Cut o computer programs have been developed to evaluate field data and compute the stability of k slopes. JTPLOT(ST-212) (SWEC 1974a) is used to reduce data from joint and foliation eys and to prepare contoured stereographic plots, such as those on Figures 2.5.1-15 and 1-16. SWARS-2P (SWEC 1974b) is used to analyze the stability of tetrahedral rock wedges med by the intersections of joint and foliations surfaces with the vertical excavation face. The are input in geological notation and are converted internally to the format required for rock hanics calculations. All possible combinations of joints are automatically considered. Effects eismic loads, rock bolts, surcharges, point loads, and several types of piezometric loads are uded in the analysis. In designing a restraining hoop or ring beam, the forces required to ilize the sliding wedges are input into the program as hypothetical rock bolts, with the load ributed across the projected vertical area of the rock wedge. A minimum safety factor of 1.1 considered acceptable for determining required stabilizing forces.
he analysis, the surcharge loading from adjacent structures was accelerated in the vertical ction, and soil surcharge was accelerated both vertically and horizontally. Water pressure was applied to the rock wedge surfaces, on the assumption that the differential head acts directly he containment wall. However, the buoyant weight of the rock was used to account for the ence of groundwater. This assumption is considered conservative because the buoyant weight ctively reduces the resistive forces. Wedges smaller than 100 cubic feet were disregarded, on assumption that these wedges were formed by the intersection of two high angle joint sets, and e probably removed during blasting and scaling operations.
ibit a high density of fractures that could produce these wedges. Figure 2.5.5-7 is a tograph showing the rock surface commonly found in the area of the main steam valve ding.
forces applied by the rock wedges on the ring beam are shown on Figure 2.5.5-3. The imum forces act in the southwest quadrant, due to the effect of the weaker foliation planes ch dip into the excavation face in this area. Other areas of instability can be attributed to the h dip angles of the jointing, which are inherently unstable when subjected to seismic and harge loadings. The design of the structural support, or ring beam, which transfers this load und the excavation, maintaining the isolation of containment structure from these external s, is discussed in detail in Section 3.8.1.1.
5.3 LOGS OF BORINGS boring logs are included in Appendix 2.5J. No borings were taken in borrow areas for erials used onsite.
5.4 COMPACTED FILL ctural backfill used to raise the shoreline slopes to final design lines meets the requirements ined in Section 2.5.4.5.2.
5.5 REFERENCES
FOR SECTION 2.5.5 5.1-1 Bishop, A.W., 1955. The use of the Slip Circle in the Stability Analysis of Slopes.
Geotechnical, Vol V.
5.1-2 Goldsmith, R. 1967. Bedrock Geologic Map of Niantic Quadrangle, Penn. U.S.
Geological Survey Quadrangle Map GQ-575, Washington, D.C.
5.1-3 Hendron, A.J.; Cording, E.J.; Aiyer, A.K., 1971. Analytical and Graphical Methods for the Analysis of Slopes in Rock Masses. NCG Technical Report No. 36.
5.1-4 Northeast Nuclear Energy Co. (NNECo.), 1975. Geologic Mapping of Bedrock Surface. Millstone Nuclear Power Station - Unit 3, Docket No. 50-423.
5.1-5 Stone & Webster Engineering Corporation (SWEC), 1974a. Stereographic Projection of Joints (JTPLOT). Computer Program ST-212. SWEC, Boston, Mass.
5.1-6 Stone & Webster Engineering Corporation (SWEC), 1974b. Analytical Method for Analysis of Stability of Rock Slopes; SWARS-2P. Computer Program ST-214.
SWEC, Boston, Mass.
Boston, Mass.
5.1-8 Stone & Webster Engineering Corporation (SWEC), 1979. Seismically Induced Displacements of Embankments and Slopes, Sides. Computer Program GT-009.
SWEC, Boston, Mass.
5.1-9 U.S. Army Coastal Engineering Research Center, 1975. Shore Protection Manual.
Department of the Army, Corps of Engineers, Washington, D.C. Vol II, Ch. 7.
5.1-10 Whitman, R.V. and Bailey, W.A., 1967. Use of Computers for Slope Stability Analyses. Journal of Soil Mechanics and Foundations Division, ASCE. Vol 93, SM4.
5.1-11 Whitman, R.V. and Moore, P.J., 1963. Thoughts Concerning the Mechanics of Slope Stability Analyses. Proceedings, Second Pan American Conference on Soil Mechanics and Foundation Engineering.
EXCAVATION BUILDING embankments or dams have been constructed at the Millstone site.
APPENDIX 2.5A- AGE OF TILL AT MILLSTONE POINT D.W. Caldwell, PhD
D. W. Caldwell, Ph.d.
Till at Millstone Point till exposures at Millstone are inadequate for the purpose of establishing their time of osition. The best exposures exist along an embankment some 200 feet long and from 10 to 15 high just to the west of the switchyard northeast of the plant. Part of this exposure has about 5 of artificial fill, covered by six inches of concrete. Beneath the fill, a grey, compact, clay -
h till is exposed. This till has few large stones and the matrix is closely jointed. At the northern of the embankment, the till is less compact and less jointed. The complete exposure is equate to determine the relationship between the two exposures of till, that is, whether one till nitely underlies the other or whether a single till simply changes in its texture and pactness from one exposure to another.
ause of the rarity of two-till exposures in New England, it is most probable that a single body ill, changing in its physical character from one place to the next, exists at the Millstone site. A ll likelihood exists that there are two separate bodies of till at Millstone, a fact which can only stablished by further and more extensive excavation.
Two-till Problem in New England several localities in southern New England there is evidence of two tills, believed to be of erent ages (Schafer and Hartshorn, 1965; Pessl and Schafer, 1968). The older till is usually pact, contains much silt and clay and has closely spaced jointing. Oxidation in the older till ges to depths of 10 or more feet. Drumlins are generally composed of the older till.
younger till is generally less compact, is sandier and does not contain numerous joints or ated structures. Oxidation is less than 3 feet and is usually absent altogether.
Age of Older Till hree localities in New England organic material beneath the lower till has been dated. At New ron, Maine, wood embedded in weathered sand, lying between two tills, has a radiocarbon age ore than 44,000 years B.P. (W-910; Caldwell, 1959). Wood at the base of an exposure of till allingford, Connecticut, has been dated as more than 40,000 years B.P. (Y-451). Peat overlain rumlin sediments in Worcester, Massachusetts, has been dated as more than 38,000 years B.P.
647, L-380). All of these listed C-14 ages of the older till are (or were at the time of analysis) ond the range of carbon-14 analysis, although they are probably post-Sangamon in age, that is, than about 100,000 years old. Without deposits of organic material at other till localities in w England, it is not possible to relate them to these three till sites where dating has been sible.
emplacement of the younger till is more firmly established than that of the older till, but her has been dated as completely and as consistently as they have been in the mid-west.
last (Wisconsin) ice sheet covered all of New England and its terminal position is marked by moraines on Long Island (Ronkonkoma moraine) and the Islands (Nantucket moraine on thas Vineyard and Nantucket). The advance of this ice sheet is dated by organic material, er than the advance, incorporated in till at Harvard, Massachusetts (21,200 +/-1,000 years B.P.;
544), and shells in drift on outer Cape Cod (20,700 +/-2,000 years B.P.). On Marthas Vineyard t material in clay (15,300 +/-800 years B.P.; W-1187) is overlain by till from a slight readvance he ice margin.
Marthas Vineyard moraine can be traced by underwater topography to Block Island and to g Island and the Ronkonkoma moraine, about 50 miles south of Millstone Point. The chart
. 1) presents the spatial and temporal relationship of the advance of the Wisconsin ice over the lstone area about 18,000 years ago. This is the approximate age of the upper part of the till at lstone (assuming there are two tills) or all the till (assuming there is but one till).
retreat of the Wisconsin ice from its terminus at the Ronkonkoma - Block Island moraine is d by numerous C-14 ages of organic material overlying the upper till. At Rogers Lake, Lyme, necticut, basal peat overlying last till is 14,240 +/-240 years B.P. (Y-950/51). Other dates of ilar material in Connecticut and Massachusetts are consistent with the Rogers Lake date.
t (1953, 1958) describes a short readvance of the ice to Middletown, Connecticut, before, but bably not long before, about 13,000 years ago. Following this readvance the ice melted dly up the Connecticut River Valley and Glacial Lake Hitchcock was formed in the valley by ck dam at Rocky Hill, Connecticut, north of Middletown. This lake was drained between 10 and 10,650 years B.P. (Y-253 and Y-251; Flint, 1956).
Conclusions last till deposition at Millstone Point occurred about 18,000 years ago. The area remained
- covered until about 14,000 years ago. If older till exists at Millstone, it may be equivalent to pre-Wisconsin-post-Sangamon till deposited more than 40,000 years ago. A more complete osure of the Millstone till would be required in order to establish the presence or absence of tills at the site.
erences for Appendix 2.5A A-1 Caldwell, D.W. 1959. Glacial Lake and Glacial Marine in the Farmington Area, Main.
Main Geological Survey Spec. Geol. Stud. 3.
A-2 Flint, R.F. 1953. Probable Wisconsin Substages and late Wisconsin Events in Northeastern United States. G.S.A. Bul. V 64, p 897-919.
A-4 Two Tills in Southern Connecticut. 1958. G.S.A. Bul. V 72, p 1687-1692.
A-5 Schafer, P. and Pessl, F. 1968. Two-till Problem in Naugatuck-Torrington Area, Western Connecticut. New England Intercol. Geol. Conf. Guidebook.
A-6 Schafer, P. and Hartshorn, J. 1965. The Quaternary of New England. In: The Quaternary of the United States, Wright, H.E. (Ed) Princeton Univ. Press.
CLICK HERE TO SEE APPENDIX 2.5B CLICK HERE TO SEE APPENDIX 2.5C CLICK HERE TO SEE APPENDIX 2.5D CLICK HERE TO SEE APPENDIX 2.5E CLICK HERE TO SEE APPENDIX 2.5F CLICK HERE TO SEE APPENDIX 2.5G CLICK HERE TO SEE APPENDIX 2.5H CLICK HERE TO SEE APPENDIX 2.5I CLICK HERE TO SEE APPENDIX 2.5J CLICK HERE TO SEE APPENDIX 2.5K CLICK HERE TO SEE APPENDIX 2.5L CLICK HERE TO SEE APPENDIX 2.5M