ML22278A209

Rev. 21 to Updated Final Safety Analysis Report, Chapter 2, Sections 2.4 Thru 5, Hydrologic Engineering

ML22278A209
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Site:	Limerick
Issue date:	09/19/2022
From:	Constellation Energy Generation
To:	Office of Nuclear Reactor Regulation
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LGS UFSAR 2.4 HYDROLOGIC ENGINEERING 2.4.1 HYDROLOGIC DESCRIPTION 2.4.1.1 Site and Facilities LGS is located in southeastern Pennsylvania on the Schuylkill River, about 1.7 miles southeast of the limits of the Borough of Pottstown and about 20.7 miles northwest of the Philadelphia city limits. The Schuylkill River passes through the site and separates the western portion, which is located in East Coventry Township, Chester County, from the eastern portion, which is partly in Limerick Township and partly in Lower Pottsgrove Township, both in Montgomery County, Pennsylvania. All of the major plant structures are located in Limerick Township. The location of the site and major plant structures with respect to the surrounding topography is shown in Figure 2.4-1.

The natural ground elevations vary from 110 feet MSL at the Schuylkill River to 280 feet MSL at the highest site elevation. The access elevations of major safety-related structures are listed in Table 2.4-1. Safety-related structures are shown in Figure 3.8-58. Because the topography of the site is such that storm water will flow away from safety-related equipment and structures, site grading is not considered safety-related.

Schuylkill River flooding and flooding that is due to the PMP falling on the plant area are discussed in Sections 2.4.2 and 2.4.3.

2.4.1.2 Hydrosphere 2.4.1.2.1 Hydrologic Characteristics The plant site is located on the east bank of the Schuylkill River at latitude 40 13' 3" N, longitude 75 35' 15" W, approximately 5.5 river miles downstream from Pottstown, Pennsylvania. The drainage area at this point is 1168 square miles, with river bed elevations ranging from about 105 feet MSL at the site to about 1750 feet MSL near the headwaters.

The watershed of the Schuylkill River (Figure 2.4-2) lies entirely in southeastern Pennsylvania. The basin is about 80 miles long by 25 miles wide and encompasses an area of 1909 square miles above its confluence with the Delaware River at Philadelphia. The principal towns and cities along the course of the river are Pottsville, Reading, Pottstown, Phoenixville, Norristown, Conshohocken, and Philadelphia.

To determine the discharge characteristics at the site, it is assumed that the flows at Pottstown and at the site are identical, since the drainage area at the site is less than 2% greater than that at the Pottstown gauging station.

The dominant hydrologic feature in the area is the Schuylkill River, which furnishes part of the project's water requirements and receives its effluents. The principal uses of the Schuylkill River are municipal and industrial water supply. The river is also used for recreational fishing and boating.

In the vicinity of the site, the Schuylkill is a meandering stream of quite gentle slope (2-21/2 feet/mile),

flanked by floodplains that consist of about 10% built-up areas, 30% thick forest growth, and 60%

cultivated or fallow fields.

CHAPTER 02 2.4-1 REV. 14, SEPTEMBER 2008

LGS UFSAR In about 1947, a massive cleanup project was started on the Schuylkill to remove the silt and debris deposited in the channel by many years of coal mining operations upstream. Physical features (stilling basins, dredged material, impounding basins, and initial channel dredging) were essentially completed by the end of 1950. Maintenance dredging will continue on an as-needed basis whenever the desilting pools become one-half full.

As for future developments along this portion of the river, no additional construction is planned downstream of the plant site, other than the project's own cooling water pumping station. Because of the width of the river at this point, the new pumping station would have no significant effect on the flood levels. Some channel improvement is planned for a 2 mile stretch of the river at Pottstown for small craft navigability and flood crest reduction, but this will have no effect on the flood levels at the plant site.

The extreme instantaneous and average daily flows of record at the Pottstown gauging station are:

Flow (cfs) Date Minimum (instantaneous) 87 August 13, 1930 Average 1,793 October 1926-September 1969 Maximum (instantaneous) 95,900 June 1972 As discussed in Section 2.4.3, the PMF on the Schuylkill River is conservatively estimated at 500,000 cfs, using Appendix B of Regulatory Guide 1.59 (Rev 2). The corresponding stage is estimated at el 174'. If it is also assumed that Ontelaunee Dam would fail under a maximum probable flood condition, the resulting stage would be increased to el 181', without wind.

2.4.1.2.2 Existing and Proposed Water-Control Structures Figure 2.4-3 shows the location of 23 small dams upstream of LGS. Heights, volumes, and drainage areas are given in Table 2.4-2. None of these dams are close enough and large enough to threaten LGS in event of failure.

There are three significant water-control structures upstream of LGS that are either in existence or in planning at the present time. They are the Ontelaunee, Blue Marsh, and Maiden Creek dams. Their locations are also given in Figure 2.4-3, and their general design characteristics are summarized in Table 2.4-3 and discussed in detail in Section 2.4.4. Ontelaunee Dam is owned by the city of Reading and is used for water supply and recreation. Blue Marsh Dam is a federally owned structure constructed (1979) by the U.S. Army Corps of Engineers. Plans for construction of Maiden Creek Dam, an authorized U.S. Army Corps of Engineers project, have been indefinitely deferred and the project is considered inactive at this time.

In Section 2.4.4, it is shown that a flood wave resulting from the failure of these three dams would result in a maximum transient water level of 201 feet, if the simultaneous crest of the Schuylkill River's SPF is assumed. The 1% wave from a 40 mph wind would have a run-up of less than 6.4 feet, so that the maximum instantaneous water surface from the combined event would theoretically be el 207'. However, conservative approaches used throughout the estimation procedure ensure that, in fact, the actual maximum water surface that is due to such an event would certainly be less CHAPTER 02 2.4-2 REV. 14, SEPTEMBER 2008

LGS UFSAR than el 207'. About 3.3 miles downstream of the plant site lies Vincent Dam - an older, free overflow, rock-filled timber crib structure about 12 feet high, with a crest elevation of 103.5 feet. The dam is used now for water supply (1.5 million gallons per day) by Home Water Company and is also used as a sediment trap and recreation pool. Since the thalweg of the Schuylkill is at el 102' to el 104' in the vicinity of the pumping station, there may be a small backwater effect at LGS. However, the Schuylkill River pumping station is not a safety-related facility, so that loss of tail-water due to a failure of Vincent Dam does not impair safety-related water supplies.

2.4.1.2.3 Surface Water Users Whose Intakes Could Be Adversely Affected by the Accidental Release of Contaminants Figure 2.4-2 shows the location of surface water users on the Schuylkill River between the site and the river mouth in Philadelphia. Domestic users, as classified by the Delaware River Basin Commission (Reference 2.4-1), are listed in Table 2.4-4, together with 1971 actual use and consumptive use details of individual entitlements. Similar information is presented in Table 2.4-5 for industrial users. Groundwater users are discussed in Section 2.4.13.2.

2.4.2 FLOODS 2.4.2.1 Flood History Flood history is discussed in detail in Section 2.4.3.5.2. Table 2.4-6 shows peak recorded flows at several stations in the Schuylkill Basin. Historically, the greatest recorded flood at Pottstown was due to Hurricane Agnes in 1972; however, according to Philadelphia records, there may have been greater floods before the beginning of the Pottstown record. None of the historic floods on the Schuylkill River (Section 2.4.3.5.2) was caused by ice jams or landslides. Surges, seiches, and tsunamis are not relevant to the LGS site. The consequences of the failures of upstream dams are discussed in Section 2.4.4.

2.4.2.2 Flood Design Considerations The DBFL with respect to the Schuylkill River is conservatively estimated at el 207'. This stage is derived from an SPF, combined with the wave crests from three simultaneous dam breaks and the 1% wave run-up generated by a 40 mph wind. Without the wave, the maximum level is estimated at el 201'. The three dams are Ontelaunee, Blue Marsh, and Maiden Creek (in the early planning stage). The derivation of this flood is discussed in Section 2.4.4.

The lowest grade level entrance to any safety-related structure is at el 217', which is 10 feet above the DBFL. Therefore, Schuylkill River floods cannot affect any of the safety-related facilities.

The Schuylkill River PMF is conservatively estimated at 500,000 cfs, based on Appendix B of Regulatory Guide 1.59 (Rev 2). When combined with a simultaneous dam break flood wave due to a PMF-induced failure of Ontelaunee Dam, the highest stage obtained at LGS was el 181' (without wind-wave). This is well below the stage obtained from the multiple dam break, as given above. The Schuylkill River PMF is discussed in detail in Section 2.4.3.

The water surface (el 201') resulting from the dam break analysis is transient. It is estimated that the time required for the dam break flood wave at the river cross-section at LGS to rise above and subside back to the SPF elevation (152 feet) is approximately 7.5 hours5.787037e-5 days <br />0.00139 hours <br />8.267196e-6 weeks <br />1.9025e-6 months <br />. The length of time for which the flood wave will be above el 177' and el 195' would be about 4.5 and 2 hours2.314815e-5 days <br />5.555556e-4 hours <br />3.306878e-6 weeks <br />7.61e-7 months <br />, respectively.

CHAPTER 02 2.4-3 REV. 14, SEPTEMBER 2008

LGS UFSAR The shortest distance from the el 201' (dam break flood elevation without wave run-up) contour to the nearest safety-related structure inland (diesel oil storage tanks) is about 126 feet. The foundation of these structures is at el 194'. For flood waters to reach these structures, percolation through the embankment would have to occur. An analysis of the percolation was made using the following equation, assuming that Darcy's law is valid and that flow is one-dimensional:

b 2b + z.b = nb (EQ.2.4-1) 2 x x x k t where:

b(x,t) = saturated thickness b = average saturated thickness k = hydraulic conductivity n = porosity z(x) = elevation of base of porous medium Subject to the following conditions:

b(0,t) = Ho b(x,0) = 0 where:

Ho = the available head at x = 0 It can be shown that the solution to the above equation can be written in the form:

b Ho e2cx erfc x erfc x c c t

2 2 t t 2 t (EQ. 2.4-2)

CHAPTER 02 2.4-4 REV. 14, SEPTEMBER 2008

LGS UFSAR where:

kb 2

n a

c 2b a = slope of the base of the porous medium The base of the porous medium (base of embankment material) was assumed to rise linearly from el 170' at the toe of the embankment to el 194' at the tanks. The results of the analysis indicate that, if a conservative permeability of 5x104 ft/yr is assumed and if the flood crest remains at el 202' (Section 2.4.3 shows maximum flood elevation of 201 feet) for a duration of two hours, groundwater would not be above el 194' anywhere within 80 feet of the tanks. Therefore the dam break flood wave would not affect hydrostatic pressures on the foundations of safety-related structures.

2.4.2.3 Effects of Local Intense Precipitation The latest updated estimate (1976) of point (10 square miles) PMP was obtained from the NOAA (Reference 2.4-2). In the updating, which was subsequent to Hurricane Agnes, relatively small increases were made compared to pre-Agnes PMP estimates from Reference 2.4-3. The updated values are given in Table 2.4-7.

As explained later, the impact of the onsite PMP was analyzed by dividing the site area into smaller subareas. For this purpose, the 72 hour8.333333e-4 days <br />0.02 hours <br />1.190476e-4 weeks <br />2.7396e-5 months <br /> PMP was divided into 6 hour6.944444e-5 days <br />0.00167 hours <br />9.920635e-6 weeks <br />2.283e-6 months <br /> increments using the distribution given in Reference 2.4-2. The 6 hour6.944444e-5 days <br />0.00167 hours <br />9.920635e-6 weeks <br />2.283e-6 months <br /> PMP was divided into 1 hour1.157407e-5 days <br />2.777778e-4 hours <br />1.653439e-6 weeks <br />3.805e-7 months <br /> increments using the U.S. Army Corps of Engineers distribution for standard project storms (Reference 2.4-4). These distributions are shown in Table 2.4-7. The 1 hour1.157407e-5 days <br />2.777778e-4 hours <br />1.653439e-6 weeks <br />3.805e-7 months <br /> PMP was further subdivided into 5, 10, and 15 minute increments using the distribution given in Reference 2.4-29.

To estimate the peak run-off from each subarea, the rational formula was conservatively used:

Q = CIA (EQ. 2.4-3) where:

Q = peak flow rate in cfs C = run-off coefficient I = rain fall intensity in inches/hour corresponding to a duration equal to the time of concentration, and A = area in acres.

The run-off coefficient, C, was conservatively assumed to be 1. Kirpich's formula (Reference 2.4-34) was used to estimate the times of concentration for different subareas.

CHAPTER 02 2.4-5 REV. 14, SEPTEMBER 2008

LGS UFSAR To obtain conservative results, the following assumptions were used in the analysis:

a. The site drainage system was assumed to be completely blocked.

b. It was assumed that storage due to ponding did not attenuate the outflow rate from any of these subareas.

c. Rating curves for flow over embankments or roadways were developed using the following equation for critical flow, based on the principle of minimum specific energy (Reference 2.4-33):

Q= (gA3/ T)1/2 (EQ. 2.4-4) where:

Q = discharge in cfs g = gravitational constant; 32.2 ft/sec2 A = cross-sectional area, ft2 T = top width of section, ft

= energy coefficient, conservatively assumed to be 1.38, based on the ratio of broad-crested to sharp-crested weir coefficients

d. For simplicity, the rating curve for CP-1 was developed using the following weir equation (Reference 2.4-33)

Q= CLH3/2 (EQ. 2.4-5) where:

Q = discharge in cfs L = crest length, 35 feet H = height of water above crest, ft C = weir coefficient, conservatively assumed to be 2.63 As shown in Figures 2.4-4 and 2.4-5, the plant site is divided into three main functional areas:

a. The turbine-reactor complex area, at el 217'

b. The cooling tower area, at el 257' to el 265'

c. The spray pond area CHAPTER 02 2.4-6 REV. 14, SEPTEMBER 2008

LGS UFSAR The cooling tower area is the highest of the three main functional areas and is located to the north of the turbine-reactor complex area. The spray pond is located north of the cooling tower area but is separated from it by approximately 300 feet. The intervening land rises (Figure 2.4-5) to a hilltop at el 285', about 800 feet east of the north-south plant centerline (E4000). Run-off from this hilltop would be directed partly to the cooling tower area; partly to the spray pond area; partly to Possum Hollow Run; partly to the turbine-reactor complex area; and partly to Sanatoga Creek. Sanatoga Creek is not shown on Figures 2.4-4 or 2.4-5. It is located north of the spray pond area and drains to the Schuylkill River upstream of the plant site. Figure 2.4-11 shows the confluence of Sanatoga Creek with the Schuylkill River near Sanatoga.

Run-off from the three main functional areas drains toward several low points, which in turn drain away from the site. Numerous local drains and small surface ditches have been provided in the site drainage system to facilitate the drainage of normal storm run-off. However, as noted above, for the investigation of onsite PMP, none of these drainage facilities is assumed to function, except for the open-channel portion of a ditch draining the cooling tower area. Otherwise, all flow is assumed to be surface flow, over land or over roadway. All drain pipes and culverts are assumed plugged.

The impact of a PMF in the Possum Hollow Run, which passes along the eastern and southern sides of the site area, is examined in Section 2.4.2.3.5. The impact of a PMF in the Sanatoga Creek is discussed in the following paragraph.

Sanatoga Creek drains an area of less than 10 square miles. With respect to the plant site, the nearest point on the creek is approximately 1400 feet upstream of its confluence with the Schuylkill River. At this location, the thalweg of the creek is approximately at el 127'. As shown in Figure 2.4-4, the spray pond is located mostly within the Sanatoga Creek Basin. The cooling towers are located on a ridge that rises in an ENE direction and separates them from the spray pond area. The same ridge forms the drainage boundary between Sanatoga Creek and Possum Hollow Run and isolates the turbine-reactor area from Sanatoga Creek. The lowest elevation in the vicinity of the spray pond is el 240', and the crest of the spray pond spillway is at el 251'. The SPF elevation of the Schuylkill River near its confluence with the Sanatoga Creek is estimated to be 155 feet. It is inconceivable that the water surface elevation in Sanatoga Creek, with backwater that is due to a concurrent SPF in the Schuylkill River, would rise higher than el 240', which is 113 feet above the creek thalweg and 85 feet above the SPF elevation of Schuylkill River. For reasons given in Section 2.4.3.5, it is considered unlikely that a PMF on the Schuylkill River would coincide with a PMF on the Sanatoga Creek. Therefore, it was concluded that a PMF in Sanatoga Creek would not endanger safety-related structures, and further detailed analysis was not considered necessary.

The following sections describe flow-routing assumptions, drainage areas, and the water surface elevations resulting from a PMP on the site area. Table 2.4-22 summarizes the drainage characteristics, probable maximum rainfall intensity for the critical duration and the resulting peak discharge for each of the subareas.

2.4.2.3.1 Drainage from Cooling Tower Area As shown in Figure 2.4-5, the cooling tower area and nearby natural topography are divided into three drainage areas, designated as DA-2, DA-3, and DA-4.

Area Designation Area (Acres) Peak Discharge (cfs)

DA-2 25.0 620 CHAPTER 02 2.4-7 REV. 14, SEPTEMBER 2008

LGS UFSAR DA-3 4.5 151 DA-4 20.8 550 2.4.2.3.1.1 Drainage Area DA-2 The run-off from DA-2 collects in part at the road junction at el 258.8', located at the northern central part of the cooling tower area. This point is designated as CP-1 (collection point-1) on detail 3 in Figure 2.4-6. From CP-1, the run-off drains westward along the roadway and the adjacent ditch and picks up run-off from the remainder of DA-2. Finally, the entire run-off from DA-2 passes through a low point in the roadway (CP-2) and enters a ravine that drains to the Schuylkill River (detail 4 in Figure 2.4-6). To the south of CP-2, the access roadway, which partly encircles the Unit 1 cooling tower, rises to a high point that forms the boundary between DA-2 and DA-3.

The most important aspect of drainage from DA-2 is the height to which water is ponded against the roadway fill that spans the distance between the two cooling towers on their south flank. This fill has a nominal top of el 264'. As long as run-off from DA-2 follows the route described and does not overtop this roadway fill, it need not be added to the flows entering the turbine-reactor complex area.

To estimate the maximum water depth against this embankment, it is conservatively assumed that all the run-off from DA-2 collects at CP-1 and passes through critical depth over a conservatively assumed effective flow width of 35 feet. Critical depth must occur at this point, since flow into the ditch downstream is supercritical. This is shown in section E-E' on detail 3 in Figure 2.4-6 which shows the assumed effective width and the discharge rating curve for CP-1. The discharge rating curve was developed using a weir coefficient of 2.63.

The water surface elevations at CP-1 and along the roadway embankment are shown on section A-A' in Figure 2.4-6. The effective width of 35 feet was selected arbitrarily to approximately correspond to the width of the main stream entering the ditch. However, flow would spread to a larger width of approximately 100 feet, as shown in detail 3 and section E-E' of Figure 2.4-6. Consequently, the maximum water surface elevation is less than the estimated value of el 262.7', which is based on an effective flow width of 35 feet. As shown on Section A-A' in Figure 2.4-6 and on Section C-C' in Figure 2.4-6, downstream of CP-1, the flow enters a ditch (supercritical flow with a normal depth of 4.4 feet) and then backs up over the roadway at CP-2 to a depth sufficient to pass the peak flow of 620 cfs.

To compute the resulting water surface elevation at CP-2, a rating curve was prepared for Section D-D' using the method based on the principle of minimum specific energy presented in Section 2.4.2.3.

It is shown in Figure 2.4-6. Using the D-D' rating curve, the water surface elevation at CP-2, corresponding to a discharge of 620 cfs, is estimated as el 245.5'. This is well below CP-1 (el 258.8'). Consideration of backwater effects demonstrates that the critical section at CP-1 is not submerged due to the flow depth at CP-2. A high point in the cooling tower access road just west of the Unit 1 cooling tower separates DA-2 from DA-3. The high point in the adjacent drainage ditch is at el 245.0'. Since the maximum water level at CP-2 was found to be el 245.5', some spillover from DA-2 to DA-3 would occur at this point. A rating curve was developed for the ditch section which acts as a control for any possible spill from DA-2 to DA-3. A water surface elevation of 245.5 feet resulted in a spill of less than 5 cfs. For purposes of estimating water levels this was considered to be negligible when compared to the DA-2 flood peak of 620 cfs and DA-3 flood peak of 151 cfs.

CHAPTER 02 2.4-8 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.2.3.1.2 Drainage Area DA-3 As shown in Figure 2.4-5, DA-3 comprises the southwest part of the cooling tower area. It drains generally into the western half of the turbine-reactor complex area, both down the roadway and down the south face of the slopes of the access roadway embankment. DA-3 includes natural ground and the roadway to the 220 kV switchyard, but not the switchyard itself, which is sloped generally south to drain toward the Schuylkill River. A peak flow of 151 cfs is generated from DA-3. The disposition of this discharge is discussed in Section 2.4.2.3.3 with the turbine-reactor complex area.

2.4.2.3.1.3 Drainage Area DA-4 Like DA-3, this area also drains directly to the turbine-reactor complex area on the eastern side of the plant centerline. However, it includes a relatively larger proportion of natural catchment, most of which is on the east of the cooling tower excavation area. The handling of the peak discharge from this area, 550 cfs, is discussed in Section 2.4.2.3.3.

2.4.2.3.2 Drainage from Spray Pond Area The spray pond drainage area, shown as DA-1 in Figure 2.4-5, includes the spray pond itself and the cut slope areas draining toward it, as well as two small pieces of natural topography that drain toward the cut slopes (one on the NW and the other on the SE perimeter). The total drainage area is 17.2 acres. Drainage is to the pond itself, which has a normal operating level of el 251' and a 60 foot wide spillway set at el 252' (shown as CP-5 on Figure 2.4-5). This spillway drains to a natural channel, across a road embankment, and then northward to Sanatoga Creek.

The spray pond spillway is designed to pass a routed PMF (48 hour5.555556e-4 days <br />0.0133 hours <br />7.936508e-5 weeks <br />1.8264e-5 months <br /> storm) preceded by an SPF assumed equal to one-half the PMF ordinates. No infiltration is assumed. The pond is assumed to be at normal water surface at the beginning of the storm. A 30 minute routing period is used, with a flood routing computer program adapted from the U. S. Army Corps of Engineers program for hydrograph combining and routing. Precipitation increments for the most critical six hours of the storm are arranged as follows:

Period Precipitation Period Precipitation (30 min) (% of 6 hr) (30 min) (% of 6 hr) 1 5 7 27 2 5 8 8 3 6 9 7 4 7 10 6 5 7 11 6 6 11 12 5 The remainder of the storm is distributed in accordance with Reference 2.4-4.

The pond is filled to a maximum of el 252.5' at the beginning of the 28th hour. The spillway discharges the surcharge above el 252' before the beginning of the second storm. The second flood peaks at a maximum mean inflow of 251 cfs, with a corresponding maximum outflow of 194 cfs and a maximum water surface of el 253'.

Further details for the pond are given in Section 2.4.8.

CHAPTER 02 2.4-9 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.2.3.3 Drainage from Power Plant Complex Area The finished floor elevation of relevant safety-related structures at the power plant complex area (power block) is el 217'. The plant coordinate line (E4000) divides the power block approximately in half, with the divide at el 217'. Drainage is generally away from the structures. The present site drainage condition was determined from the Limerick Site Master Plan, and various walkdowns.

Jersey barriers have been strategically placed along sections of the perimeter fence for security reasons. In the safety evaluation, the jersey barriers are assumed to block the drainage flow, and thus forming boundaries for the drainage areas. Some potential flow outlets along the south-west boundary of the site have been blocked, or partially blocked, by the jersey barriers. In addition, jersey barriers placed between the Technical Support Center and the Warehouse (see Figure 2.4-6) form a portion of the boundary separating the drainage areas DA-5 and DA-6.

The boundary separating drainage areas DA-5 and DA-6 runs along the jersey barriers, from the Technical Support Center to the Warehouse (see Figure 2.4-6). The boundary then runs along the southern edge of the Turbine Building, up to the centerline of the power block. The remaining southern portion of the boundary runs south along the centerline of the power block. For drainage area DA-5, the northern boundary is the plant access road which separates drainage area DA-3 and DA-5; and the western and southern boundaries are along the plant south-west security fence. For drainage area DA-6, the eastern boundary is the outside edge of the plant access road; the northern boundary is the service road which separates drainage areas DA-6 and DA-4; and the southern boundary is the service road adjacent to the southern side of the power plant complex.

The total drainage areas and peak flows from DA-5 and DA-6 were estimated to be as follows:

Area Designation Area (acres) Peak Flow (cfs)

DA-5 14.9 369 DA-6 11.6 242 The surface water flow paths around the power block shown on Figure 2.4-6 have been limited in order to provide additional space for lay-down or storage areas. The locations of the critical cross-sections and flow outlets are shown in Figure 2.4-6, while the widths and average bottom elevations for each of the cross-sections and flow outlets are summarized in Table 2.4-23.

Within drainage area DA-5, the important features are the yard and roads, the power plant structures, the refueling water tank dike, and various buildings (especially the ones erected along the security fence right-of-way on the southwest side of DA-5). The yard and road elevations encompassed by the security fence vary from about el 212 to el 217. The roads entering the power plant structures have crowns set at el 217 or lower. These are sloped to drain away from the structures.

The area assumed for DA-5 includes, conservatively, the roof area of the power block and the area bounded by the condensate storage tank and the resulting water tank dike. The tank dike crest is at el 223, approximately 7 feet above grade. The tank dike, therefore, would contain much more than the 40 inches of rainfall resulting from the 72 hour8.333333e-4 days <br />0.02 hours <br />1.190476e-4 weeks <br />2.7396e-5 months <br /> PMP.

The path of the surface water flow in DA-5 will begin with inflow from DA-3 and a portion of DA-4, through the northern boundary of DA-5. Run-off from DA-5, combined with the inflow, will flow along CHAPTER 02 2.4-10 REV. 14, SEPTEMBER 2008

LGS UFSAR the paths shown in Figure 2.4-6, west of the power block, and exit area DA-5 through the outlets along the perimeter fence on the south-west side of the power plant complex. In the safety evaluation, jersey barriers were assumed to block some the outlets along the south-west boundary of DA-5.

Within drainage area DA-6, important features are the yard and roads, and the parking lot around the Site Management Building, shown in Figure 2.4-6.

The surface water flow in DA-6 will begin with inflow from DA-4 along the northern boundary, as shown on Figure 2.4-6. The run-off from DA-6, combined with the inflow, will flow through the parking lot, and exit DA-6 through the outlet on the eastern side of the site, between the Site Management Building and the Warehouse and Procurement Building, as shown in Figure 2.4-6.

The rational formula was used to estimate peak run-off rates at collection points CP-3 and CP-4 (see Figure 2.4-4), which are 446 cfs and 555 cfs, respectively. Flooding calculations were performed assuming critical flow at the outlets, and a Mannings n equal to 0.022. These calculations yield maximum water surface elevations of 218.1 feet at CP-3, and 218.1 feet at CP-4. Backwater calculations performed for the flow path located north of the Turbine Building (shown on Figure 2.4-6) yielded a maximum water surface elevation of 218.6 feet along the northern edge of the Turbine Building.

The results of the peak run-off rates and peak water levels for the present condition at the power block are summarized in Table 2.4-8.

An engineering evaluation of the site drainage conditions determined that the only potentially adverse affect is that flood water due to the local intense precipitation (PMP) event or the postulated cooling tower basin failure (PCTBF) could enter the turbine enclosure for a limited period of time. At certain site locations, the PCTBF event results in higher flood levels than the PMP event, but the PMP event is the bounding event due to the longer duration of the event. Consequently, measures are taken to limit the possible flow of flood water from the turbine enclosure into the control enclosure where the safety-related chilled water system is located at el 200. These measures include controlling openings in the lower portion of the turbine enclosure walls, relying on flood control features within the turbine building, and by limiting the flow of water through the turbine enclosure doors by providing additional curbs and barriers inside the turbine enclosure and by administratively controlling the opening of certain doors. As a result of these measures, flooding into the control enclosure due to the present site drainage condition results in control enclosure flood level below the elevation of the safety-related chilled water system.

Furthermore, administrative controls have been implemented to ensure that future changes in site conditions, affecting flood water run-off, will receive engineering evaluation and approval prior to implementation.

CHAPTER 02 2.4-11 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.2.3.4 Roof Loads on Safety-Related Structures That Are Due to PMP Onsite In the previous analysis of surface drainage, it is assumed that all roof drainage overflows to the ground and then to the various control points. If all roof drains and scuppers are blocked, water could pond on the roofs of some safety-related structures to a depth controlled by the height of the roof parapets. The highest parapet on any safety-related structure is less than the maximum 24 hour2.777778e-4 days <br />0.00667 hours <br />3.968254e-5 weeks <br />9.132e-6 months <br /> PMP of 34.4 inches. Assuming that some accumulation and overflowing occur, the maximum water depth could equal the height of the parapet plus a small amount of head providing flow over the parapet. The design roof load due to PMP for all the safety-related structures (Section 3.2) is equivalent to this maximum water depth.

2.4.2.3.5 PMF in Possum Hollow Run Possum Hollow Run has a drainage area of 1.3 square miles. It rises approximately 2.5 miles northeast of the site and flows southwesterly, entering the Schuylkill River through a gorge along the south side of LGS.

To assess the flood hazard to the LGS site posed by Possum Hollow Run PMF, the PMF is assumed to occur in Possum Hollow Run at the same time that an SPF is occurring in the Schuylkill River. The Schuylkill River SPF is assumed to be 50% of the PMF, or 250,000 cfs, which results in a Schuylkill River stage of el 152'.

It is unlikely that a PMF on the Schuylkill River would be coincident with the PMF on the Possum Hollow Run. A PMF on the Possum Hollow Run (drainage area = 1.3 square miles) is caused by a local intense thunderstorm, while a PMF on the Schuylkill River (drainage area = 1170 square miles) is due to a basin-wide PMP storm system whose center lies well upstream of LGS. The two storms would have different characteristics, and the joint probability that they produce peak runoffs at LGS at the same time is very low.

Using the slope-area method, a rating curve was developed for the Possum Hollow Run. For this purpose, a typical cross-section is taken at the point where the el 152' contour crosses the stream.

The bed slope of the Possum Hollow Run is 0.02. Based on field inspections and comparison with photographs of streams with known (n) values (References 2.4-6 and 2.4-33), a Manning's (n) value of 0.05 is assumed.

A PMF hydrograph was developed for the Possum Hollow Run using the procedure outlined in Reference 2.4-6. The 6 hour6.944444e-5 days <br />0.00167 hours <br />9.920635e-6 weeks <br />2.283e-6 months <br /> PMP (Table 2.4-7) is divided into one-half hour increments, following the distribution used for the spray pond (Section 2.4.2.3.2). The resulting hydrograph peak is 3840 cfs, and the base is 15.5 hours5.787037e-5 days <br />0.00139 hours <br />8.267196e-6 weeks <br />1.9025e-6 months <br />.

Using the rating curve developed for the Possum Hollow Run, the water surface elevation corresponding to a PMF discharge of 3840 cfs is estimated to be 159 ft. The corresponding normal depth, velocity, and Froude number are 7.0 ft, 8.6 ft/sec, and 0.76, respectively.

As stated earlier, the SPF elevation in the Schuylkill River is 152 feet MSL. The bed elevation of the Possum Hollow Run at its confluence with the Schuylkill River is 105 feet MSL. Therefore, at the time of the SPF, the flood water of the Schuylkill River enters the Possum Hollow Run up to a point where its thalweg is at el 152'. This point is about 2400 feet upstream of the confluence of the Possum Hollow Run with the Schuylkill River. The PMF on the Possum Hollow Run is superimposed on the SPF in the Schuylkill River. It is conservatively assumed that in this backwater reach of 2400 CHAPTER 02 2.4-12 REV. 14, SEPTEMBER 2008

LGS UFSAR feet the cross-sectional area of the Possum Hollow Run below el 152' is ineffective, and the entire PMF flow passes through the remaining cross-sectional area. This results in a water surface elevation of 159 feet at a section 2400 feet upstream of the confluence. This flood level is about 57 feet below the plant bench. Upstream of this section, the Possum Hollow Run is separated from the plant by high ground that is more than 60 feet above the bed of the Possum Hollow Run. It is not credible that the PMF of the Possum Hollow Run (3840 cfs) causes a water depth of 60 feet or more.

Therefore, it is concluded that the PMF in the Possum Hollow Run coincident with the SPF in the Schuylkill River would not flood any safety-related structure at LGS.

2.4.2.3.6 Summary of Results - Local Intense Precipitation Table 2.4-8 summarizes the results of local intense precipitation investigations at the associated collection points and drainage areas. These investigations show that water levels are either below the access elevations of the safety-related structures or evaluated to be acceptable as shown in Section 2.4.2.3.3.

The Possum Hollow PMF is treated apart from the onsite local intense precipitation. A maximum PMF discharge of 3840 cfs was calculated. An SPF occurring simultaneously in the Schuylkill River produces a backwater effect, but conservative estimates indicate that the water surface elevation would be well below the plant grade el 216.5'.

2.4.3 PROBABLE MAXIMUM FLOOD ON STREAMS AND RIVERS The PSAR included a PMF developed by the U.S. Army Corps of Engineers for Pottstown, adjusted to the site by the ratio of drainage areas. This PMF was estimated at 356,000 cfs, with a stage at the plant site of 158 feet. After the PSAR was prepared, Hurricane Agnes occurred (1972). In 1975 the PSAR was update to reflect the effect of Hurricane Agnes. At that time there was no reason to change the PMF. But later the NWS revised the values of PMP for the basin. Another significant change occurred with respect to upstream dams. Construction began on Blue Marsh Dam, and Maiden Creek Dam became an authorized project, whereas at the time of the PSAR, construction on Blue Marsh had not started, and Maiden Creek was not expected to be authorized. These changes combined to require a new flood analysis.

Since the LGS PSAR was prepared, Regulatory Guide 1.59 (Rev 1 and Rev 2) were issued, giving the option of using either detailed flood routing studies (Appendix A) or enveloping maps for determining peak PMF flows (Appendix B). The latter method is simpler, but more conservative, giving a value of 500,000 cfs as compared with the PSAR estimate of 356,000 cfs (both before adjustment for a dam break). The Appendix B method was selected for this report because it is conservative and indicates a "dry-site" condition during the PMF, with a large margin of safety.

PMFs were computed for the Schuylkill River at LGS and for Possum Hollow Run. The computations show that the maximum resulting stage in the Schuylkill River is el 181'. The maximum stage that is due to a PMF in Possum Hollow Run was not computed, but was found to be less than el 186' at a point nearly due east of the turbine-reactor area complex. The PMF for the Schuylkill River is covered in this section. The PMF for Possum Hollow Run is discussed in Section 2.4.2.3.5.

CHAPTER 02 2.4-13 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.3.1 Probable Maximum Precipitation This section is not applicable because the PMF is estimated in accordance with Appendix B of Regulatory Guide 1.59.

2.4.3.2 Precipitation Losses This section is not applicable because the PMF is estimated in accordance with Appendix B of Regulatory Guide 1.59.

2.4.3.3 Run-Off and Stream Course Models This section is not applicable because the PMF is estimated in accordance with Appendix B of Regulatory Guide 1.59.

2.4.3.4 Probable Maximum Flood Flow From Appendix B of Regulatory Guide 1.59, the PMF for the Schuylkill River at the LGS site, corresponding to a drainage area of 1170 square miles, is 500,000 cfs.

The design flood for Ontelaunee Dam is 41,000 cfs. The PMF inflow to Maiden Creek Dam (which may be built at the headwater of Lake Ontelaunee) has been estimated by the U.S. Army Corps of Engineers at 118,000 cfs, with a spillway peak outflow at 92,000 cfs.

Blue Marsh Dam was designed by the U.S. Army Corps of Engineers to pass a PMF without failure.

Under current authorization plans, the U.S. Army Corps of Engineers plans to design the Maiden Creek Dam spillway for the PMF. Both of these projects attenuate their PMF inflow peaks considerably. For Blue Marsh, the attenuation is from 128,600 to 74,800 cfs, and for Maiden Creek, it is from 118,000 to 92,000 cfs. In the analysis of flooding that is due to a dam break, it is conservatively assumed that no attenuation occurred during the PMF passages through either Blue Marsh or Maiden Creek Dam.

The flood stage at the LGS site for a 500,000 cfs PMF peak is el 174', not accounting for a hypothesized failure of Ontelaunee Dam (Section 2.4.3.5). To assess stages produced by the flood wave caused by the PMF-induced failure of Ontelaunee Dam, the results of a study of dam failure permutations are used. The method of analysis used in this study is the same as described in Section 2.4.4.2 (References 2.4-23 and 2.4-24). One of the permutations in this study postulates the failure of Blue Marsh Dam superimposed on the SPF in the Schuylkill River. This results in a water surface elevation of 177 feet at LGS. The flood wave caused by the failure of Ontelaunee Dam is superimposed on the above condition, resulting in a water surface elevation of 184 feet at LGS.

Thus, the rise in water surface elevation that is due to the failure of Ontelaunee Dam is 7 feet.

The assumption is made that a failure of Ontelaunee Dam also produces an incremental increase in stages of 7 feet at LGS. Thus, the water surface elevation for the PMF plus a failure of Ontelaunee, is estimated as el 174' + 7' = 181'. This elevation (el 181') is 19.7 feet lower than the water surface elevation (el 201') caused by the hypothesized seismically induced failure of three major dams upstream of LGS (Section 2.4.4.2).

Therefore, the latter elevation governs plant safety, and a more refined analysis for the water surface elevation caused by the PMF combined with the failure of Ontelaunee Dam is not warranted.

CHAPTER 02 2.4-14 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.3.5 Water Level Determinations A discharge rating curve for the Schuylkill River near the project site (Figure 2.4-7) was developed using observed flood levels, computer backwater studies, and slope-area methods. The procedure that was used is discussed in the following sections.

2.4.3.5.1 Data Availability Stream flow data are available in USGS publications (References 2.4-7 through 2.4-11). Within the basin and upstream of the plant site, 22 gauging stations have been operated, with 11 presently active. The stations are listed in Table 2.4-9, and their locations are indicated on Figure 2.4-8.

Additional data are contained in References 2.4-12, 2.4-13, and 2.4-14. Data on river and flood profiles are available from the Commonwealth of Pennsylvania, U.S. Army Corps of Engineers, and USGS, in addition to a special high-water mark survey commissioned by the licensee in 1972 (Figure 2.4-9).

2.4.3.5.2 Historical Floods Flood-producing storms in this area are normally associated with tropical disturbances. Although flooding from snowmelt occurs annually, snowmelt run-off usually has not been associated with major historic floods. Peak stages and discharges published by the USGS and the U.S. Army Corps of Engineers for the major historic floods are given in Table 2.4-6 for several stations on the Schuylkill River. At Pottstown, the 1902 flood, with a peak discharge of 53,900 cfs, was the highest known until June, 1972. However, the Reading and Philadelphia data indicate that the 1902 flood was very likely exceeded in 1850 and 1869 and may have been exceeded in 1757 and 1839.

In June, 1972, Hurricane Agnes produced the flood of record on many Pennsylvania streams. The flow at Pottstown has been evaluated as 95,900 cfs by the USGS (Reference 2.4-15). Figure 2.4-10 shows the flood frequency curve for the Schuylkill River at Pottstown. This curve is based on composite regional flood discharge relationships given in Reference 2.4-16. It is not expected that the 1972 flood alters these regional relationships.

2.4.3.5.3 Water Level Determinations up to 356,000 cfs Table 2.4-10 gives the values from which the rating curve, shown in Figure 2.4-7, was drawn. For flows above 20,000 cfs, bridge clogging is assumed (Section 2.4.3.5.3.5).

After completion of the studies that resulted in the rating curve shown in Figure 2.4-7, Hurricane Agnes produced the flood of record at Pottstown, 95,900 cfs. Some seven hours before the flood's peak, an oil lagoon at Pottstown was overtopped by the flood waters, producing a slick along the river that left oil marks for a considerable distance downstream (Reference 2.4-17).

Figure 2.4-9 shows the results of a special survey commissioned by the licensee in July, 1972 to determine high-water marks between Sanatoga (1.4 miles upstream from the plant) and Cromby (8.6 miles downstream). All readings were taken along the east side of the Schuylkill River and reflect the top of the oil marks where they were visible.

CHAPTER 02 2.4-15 REV. 14, SEPTEMBER 2008

LGS UFSAR If an upper envelope of that portion of the profile near the site is taken as indicative of the actual high water, it appears that the 1972 flood rose to about el 131'. In addition, there appears to have been almost no clogging of the bridges during this flood.

The rating curve (Figure 2.4-7) indicates a water level at about el 134.3' for a flow of 95,900 cfs. This is 3.3 feet higher than el 131' given by the flood level survey shown in Figure 2.4-9. The assumption of bridge clogging (Section 2.4.3.5.3.5) accounts for about 2 feet of the difference. The rating curve is apparently conservative when used for estimating water level for a given discharge.

2.4.3.5.3.1 Methods of Computation The geometry of the channel near the site is taken from a survey made in 1969. Normal uniform flows are assumed for the low flows, and an approximate roughness is obtained using the average water surface slope shown in Reference 2.4-18. These low-stage computations were checked by field observations in December, 1969.

The flood levels for flows from 20,000 to 356,000 cfs are obtained using an adaptation of the U.S.

Army Corps of Engineers Standard Step Backwater Program (Reference 2.4-19). Flood levels for flows above 356,000 cfs are discussed in Section 2.4.3.5.4. The program employs a computing method similar to method 1 in Reference 2.4-20. Whenever data are insufficient, conservative estimates are made. The computations consider the 14.1 miles of river between Pottstown (5.5 miles upstream from the site) and the village of Cromby (8.6 miles below the site).

2.4.3.5.3.2 Topographic Data Figure 2.4-11 shows the locations of cross-sections used in the backwater studies. Data are obtained from four sources:

a. Commonwealth of Pennsylvania cross-sections taken in 1967 for the vicinity of Pottstown (river stations 3680+00 through 3810+00)

b. U.S. Army Corps of Engineers cross-sections taken in 1969 in the same reach to supplement the commonwealth sections

c. USGS topographic sheets revised in 1968, Pottstown and Phoenixville 7.5 minute quadrangles (10 foot contours) for the sections downstream of river station 3810+00

d. Specific surveys made for these studies in 1969 and 1970 by the licensee of all bridges between Pottstown and Vincent Dam and of the river bottom close to the site The only modifications made to existing topographic data consist of assuming that impounding basins along the river are full of dredged material. For the sections taken from the topographic maps, approximate bottom elevations are developed using information contained in the 1950 report to the Commonwealth of Pennsylvania by the Schuylkill River project engineers (Reference 2.4-18).

The surveyed cross-sections indicate that the main channel could be represented as having a horizontal bottom two feet above the thalweg elevation. This approximation is extended to the main channel portions of the unsurveyed sections, if they are assumed to be rectangular in section with their base elevations two feet above the thalwegs shown in the 1950 report.

CHAPTER 02 2.4-16 REV. 14, SEPTEMBER 2008

LGS UFSAR Lengths of reach for both the main and overbank sections are scaled from the USGS topographic maps.

USGS aerial photographs taken in mid-1968 are used to determine floodplain character for roughness evaluation. Floodplain use culture undoubtedly changes from time to time and place to place, but it is assumed that the net effect of such changes is to keep the floodplain roughness approximately constant.

2.4.3.5.3.3 Selection of Observed Flood Profile Roughness coefficients for a natural stream are best determined by a trial-and-error process of matching observed high-water profiles with those obtained from backwater computations using various roughness coefficients.

Flood profile information is not available for the 53,900 cfs flood of February, 1902.

Several flood marks are available from the USGS and the U.S. Army Corps of Engineers (Reference 2.4-21) for the 50,800 cfs flood of May, 1942 but were collected before the river restoration project was completed in 1950. Changes to the river, the installation of the impounding basins and dredging, have been sufficient to significantly alter the hydraulic characteristics of the river and floodplain.

The largest flood between 1950 and the time that the rating curve was computed occurred in August, 1955, with a peak of 42,300 cfs. The flood level at the USGS gauge on the Hanover Street Bridge in Pottstown was obtained from Reference 2.4-10, and other flood marks from Reference 2.4-21.

These data are as follows:

OBSERVED MAXIMUM WATER LEVELS, 1955 Location Flood Level (ft)

Pottstown Hanover St. Bridge Station 135.84 S. Pottstown Madison (Keim) St. Bridge 133.55 Linfield Highway Bridge 120.03 Spring City Highway Bridge 105.97 See Section 2.4.3.5.3 for a discussion of the 1972 flood.

2.4.3.5.3.4 Derivation of Manning's (n)

In 1969, the Philadelphia District Office, U.S. Army Corps of Engineers, performed backwater studies in connection with a proposed channel improvement project near Pottstown. The following values of Manning's (n) were apparently used by the U.S. Army Corps of Engineering in their 1969 study:

Natural main channel (before improvement) 0.042 Built-up areas 0.062 Fields (cultivated and fallow) 0.062 CHAPTER 02 2.4-17 REV. 14, SEPTEMBER 2008

LGS UFSAR Wooded areas 0.078 Aerial photographs support the values 0.062 and 0.078 as reasonable for overbank roughness, considering the types of culture along the river. Because of lack of sufficient definition of observed profiles, it was decided to adopt those values for overbank flows and to use main channel roughness coefficients that result in a surface profile matching the observed 1955 flood levels.

Conversation with a witness to the 1955 flood at Pottstown revealed that only small flotsam was involved and that most probably no clogging occurred at any of the bridges. Thus, the bridge openings were considered clear in the calculations performed to evaluate (n). Table 2.4-11 compares the computed and observed values of water levels and gives the (n) values developed for each reach.

The increase in computed roughness going downstream could be caused by actual discharge increases along the 111/2 mile reach, or it could reflect the attenuation of the flood wave as it moved downstream. These, however, are probably of minor importance, and the increase in (n) is probably due to an increasing lack of definition of the physical description of the channel.

It should be noted that a weighted mean of the calculated values agrees well with that used by the U.S. Army Corps of Engineers (0.041 versus 0.042). In terms of flood elevations at the site, it is conservative to use large roughness factors downstream of the project site.

Additional conservatism is introduced by assuming the (n) value to be constant for all flow stages.

2.4.3.5.3.5 Bridge Clogging The following assumptions are made regarding clogging at the various bridges: openings in handrails and trusses are assumed to be fully clogged; bridge openings flanked by woods are assumed 50% clogged; and main channel openings are taken as 20% clogged. In the 1955 flood, as noted above, the bridges did not become clogged with debris. However, in a flood such as the PMF, large debris would be common, and the bridges could be subject to clogging. Because the bridges are over a mile downstream and well submerged during extreme floods, the net effect of clogging is approximately a 2 foot increase in water level at the site.

2.4.3.5.3.6 Selection of Starting Water Surfaces To obtain proper convergence, it is necessary to extend the computations to Cromby, 8.6 miles downstream of the project site. At this downstream location, the slope-area method is used to obtain an approximate water level. A typical cross-section and the average slope observed during the 1955 flood are used for this. It is assumed that the true water surface at this location falls between the two elevations bounding a 25% and around the computed conveyance (AR2/3/n). Table 2.4-12 shows the convergences obtained in the various runs made in this study.

CHAPTER 02 2.4-18 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.3.5.3.7 Flood Discharges Studied The lower portion of the curve in Figure 2.4-7 was developed from several computed water levels.

The results in Table 2.4-10 have already been noted, but some discussion of the flood quantities is in order.

a. The average annual flood discharge, 21,000 cfs is based on 42 years of data at the USGS gauge at Pottstown.

b. The average annual flood at Pottstown, 28,000 cfs is computed from regional data presented by the USGS in Water Supply Paper 1672.

c. The second highest flow of record, 53,900 cfs is at Pottstown.

d. The 100 year flood, 99,000 cfs is computed from the regional data in Water Supply Paper 1672.

e. An arbitrary discharge used to obtain 200,000 cfs is better definition of the rating curve.

f. When the PSAR was prepared, 356,000 cfs was the original modified U.S. Army Corps of Engineers estimate of the PMF. In this report, the PMF is estimated at 500,000 cfs, using the more conservative estimating procedure given in Appendix B of Regulatory Guide 1.59. The calculation of levels for the 356,000 cfs flow with the backwater program provided an additional point for graphic extension of the rating curve.

Figure 2.4-12 shows the computed water surface profiles for these floods for the reach between Sanatoga Highway Bridge and Linfield Railroad Bridge. These two points are approximately 4500 feet upstream and 7500 feet downstream from the site, respectively.

2.4.3.5.4 Water Level Determination for Flows Above 356,000 cfs The discharge rating curve shown in Figure 2.4-7 is extended to above the 356,000 cfs flow (el 158')

using the approximate but conservative method described in this section.

The variation of friction slope with discharge in the backwater studies is plotted in Figure 2.4-13. The actual conveyance provided at the depth obtained in the backwater studies is estimated from the formula:

K = 1.486 (AR2/3)n-1 (EQ. 2.4-4) where:

K = conveyance A = flow area R = hydraulic mean depth n-1 = coefficient of roughness CHAPTER 02 2.4-19 REV. 14, SEPTEMBER 2008

LGS UFSAR The friction slope is computed from the formula:

Sf = (Q/K)2 EQ. 2.4-5) where:

Sf = the friction slope Q = the discharge K = the conveyance of the river at the LGS site.

This is done for each of the six discharges used to construct the rating curve at and below 356,000 cfs (Table 2.4-10)

In Figure 2.4-13, the points obtained in the backwater study are indicated with circles. As discharge increases from 21,000 to 28,000 cfs, there is essentially no change in friction slope. However, as discharge then changes to 53,000 cfs, there is a sharp decrease in friction slope; this continues to 99,000 cfs, which seems to be at or near a minimum friction slope; as discharge increases beyond 99,000 cfs, the friction slope increases. The minimum slope obtained here, however, is not the same as the minimum slope of uniform flow, critical slope. All the flows considered are well into the subcritical range, being M1-type backwater curves. Differentiating Equation 2.4-5 with respect to (Q),

the following equation is obtained:

s 2Q K K Q 3 (EQ. 2.4-6)

Q K K Q This equation shows that if the ratio Q/Q is larger (smaller) than K/K in a certain range of depth, there is a corresponding decrease (increase) in slope and an increase in discharge. Thus, the shape of slope-discharge relation (Figure 2.4-13) is a consequence of the backwater starting elevations and the consequent variation of conveyance with depth at LGS.

There appears to be a trend toward an asymptote of S = 0.00038 as (Q) increases beyond 356,000 cfs. However, without additional backwater computations to confirm this trend, its extrapolation would be questionable. To be conservative with respect to depth, a decreasing slope versus discharge relation was chosen instead of the apparent asymptotic value. The relationship chosen is the best straight-line fit-by-eye to the six points in Figure 2.4-13.

At extreme flood stages above el 158' at the LGS site, the left bank (facing downstream) curves away from the river alignment northeasterly, toward Possum Hollow Run. When the stage of the river rises above this level, this side of the cross-section is less effective in carrying discharge, because of local separation of flow in the bend. A conservative extension of the rating curve is made by neglecting a portion of the left side of the stream (looking downstream) in computing cross-sectional area. A plot of the LGS cross-section looking upstream is given in Figure 2.4-14 and shows the area used in computing conveyance above el 158'.

The slope-discharge relation of Figure 2.4-13 is used to determine the required conveyance and associated stages, as given in Table 2.4-13.

CHAPTER 02 2.4-20 REV. 14, SEPTEMBER 2008

LGS UFSAR The conservative nature of this extension is shown by comparing the conveyances required at 356,000 cfs. The backwater data for 356,000 cfs yield a friction slope of 0.000366, with a required conveyance (K) of 356,000 (0.000366)-1/2 = 18.6x106 cfs. The straight-line relation in Figure 2.4-13 indicates a slope of 0.00028, which gives a conveyance requirement of K = 356,000 (0.00028)-1/2 =

21.3x106 cfs. This is 14% higher than the conveyance actually required, based on the backwater study, and indicates a correspondingly higher stage at LGS. While the stage for 356,000 cfs is based on the backwater study (el 158'), this comparison demonstrates the conservative nature of the straight-line relation shown in Figure 2.4-13 when used in extending Figure 2.4-7 above el 158'.

The required stage given in Table 2.4-13 is determined from a stage conveyance curve, Stage = f(K),

in which the conveyance K is determined from Equation 2.4-4.

In this case, a weighted (n) value was computed for a water surface of el 135' based on 0.042 for the channel and 0.07 for the overbank area. The weighted value obtained is n = 0.063, and this value is used for all depths. For computing the area and hydraulic radius above el 158', the cross-section portion noted above on the left bank is omitted.

The final stage relation is shown in Figure 2.4-7. The conservative nature of the extension above el 158' (356,000 cfs) is clearly shown by the change in slope of the curve at that point. If the straight-line stage-discharge relation that prevails below el 158' is extrapolated, it gives a lower stage for the same discharge. The extended stage rating curve should give conservative estimates of stage at LGS for all discharges covered. This rating curve is not applicable to dam break waves. The stage and discharge during the passage of such waves are discussed in Section 2.4.4. The calculations for the above rating curve are based on the assumption that the river channel carries steady, gradually varied flow with the origin of the backwater curve at Cromby.

2.4.3.6 Coincident Wind-Wave Activity The water surface obtained with the PMF without wind is less than that obtained with the SPF and a multiple dam break, as discussed in Section 2.4.4. Since that case is more critical, the discussion of wind-wave activity is omitted here (see Section 2.4.4.3 for a discussion of wind-wave activity under the more adverse dam break case).

2.4.4 POTENTIAL DAM FAILURES, SEISMICALLY INDUCED The reservoirs in the Schuylkill Basin upstream of LGS can be classified as being either minor or major with respect to a seismically induced failure.

Table 2.4-2 is a list of minor dams that are either too small or too remote to cause significant flooding at LGS in the event of their seismic failure. The table indicates that these are all less than 100 feet high and less than 4000 acre-feet in volume. Except for three Schuylkill River dams, their drainage areas are all less than 16 square miles. Their locations are shown in Figure 2.4-3.

Table 2.4-3 lists three major dams whose seismic failure could, under certain circumstances, generate significant waves in the LGS reach of the Schuylkill River. These structures are discussed further, and their various failure permutations are considered here in detail.

Ontelaunee Dam has a storage volume of 11,900 acre-feet and a height of 52 feet. It is principally of concern because it is located immediately below the site of another and larger dam that is authorized for future construction by the U.S. Army Corps of Engineers - Maiden Creek Dam. Maiden Creek CHAPTER 02 2.4-21 REV. 14, SEPTEMBER 2008

LGS UFSAR Dam will be situated 5 miles upstream of Ontelaunee Dam and will have a volume of 114,000 acre-feet, of which 38,000 acre-feet will be dedicated to flood control storage; the maximum height will be 110 feet.

Blue Marsh Dam, completed in 1979, is about 35 miles upstream of LGS on Tulpehocken Creek.

This U.S. Army Corps of Engineers' project has a total storage of 50,000 acre-feet, of which a minimum of 22,900 acre-feet is dedicated for flood storage; the maximum height is 96 feet.

Figure 2.4-14 presents a location map and a schematic profile relating Ontelaunee, Maiden Creek, and Blue Marsh dams in elevation to LGS. All three dams have, or will have, uncontrolled open spillways.

Felix Dam is a recreation dam on the Schuylkill River in the city of Reading, which is downstream of the Maiden Creek confluence. The pool has a volume of 1470 acre-feet, and the dam is a 24 foot high rock-filled timber crib overflow structure, with a spillway el 237.5' MSL. This dam will be drowned out (or will be previously destroyed) by the SPF, and the flood wave resulting from its failure therefore need not be added to the flood waves from the failure of the Ontelaunee and Maiden Creek dams upstream.

Appendix A of Regulatory Guide 1.59 suggests that an acceptable combination of run-off floods and seismic events would be the SSE with a 25 year flood and the OBE with a standard project flood.

The OBE has a design acceleration of 0.075 g horizontal and 0.050 g vertical. The SSE has a design acceleration of 0.15 g horizontal and 0.10 g vertical.

On Blue Marsh Dam, the U.S. Army Corps of Engineers provides for horizontal design accelerations of 0.05 g for concrete structures and 0.10 g for the embankment; no provision is made for a vertical acceleration component. Plans for completing final design and for construction of Maiden Creek Dam, also a U.S. Army Corps of Engineers project, has been indefinitely deferred; the U.S. Army Corps of Engineers presumably would apply its Blue Marsh seismic design values to Maiden Creek Dam.

There is no indication that Ontelaunee Dam, an earth-fill and masonry spillway structure complete in 1934 and now owned by the City of Reading, has any specific design provision for seismic loadings.

The U.S. Army Corps of Engineers design accelerations do not clearly fall into either the OBE or the SSE design category. On the one hand, the embankment horizontal design acceleration exceeds the OBE horizontal design value, but on the other hand the concrete structure horizontal design value is less than the OBE. In addition, no provision is included for vertical design acceleration.

While the U.S. Army Corps of Engineers dam designs embody a considerable resistance to seismic failure, the amount of damage accompanying either an OBE or SSE is difficult to assess without a detailed seismic analysis of those structures. Therefore, for the purpose of this report, the two U.S.

Army Corps of Engineers dams and the Ontelaunee Dam are all considered as nonseismic Category I structures. Their total instantaneous failure is considered simultaneously with the LGS SPF, which thus corresponds to an OBE design failure condition. This simplification negates the need for study of the SSE combined with a 25 year flood, since that case is simply the same total seismic failure, but with a lesser coexisting flood.

CHAPTER 02 2.4-22 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.4.1 Dam Failure Permutations If a major seismic failure occurred at the future Maiden Creek Dam, a domino-type failure would be likely at Ontelaunee because of the limited spillway capacity. Moreover, these two dams will be approximately 5 miles apart, and Blue Marsh Dam on Tulpehocken Creek is only 6.7 miles from Ontelaunee; thus a seismic event severe enough to cause failure in any one of these could cause severe damage to the other two. Therefore, a multiple failure analysis is considered in which all three structures are considered to fail in such a way that their peaks arrive at LGS simultaneously.

The failure, seismic or otherwise, of any of the six Schuylkill River navigation dams downstream of LGS would not affect safety-related water supplies, since safety-related water supplies do not rely on the Schuylkill River.

Regulatory Guide 1.59 (Rev 2) specifies that the appropriate SPF at the dam should be coincident with the dam failure in the OBE case and that the flood control pool should be filled. This specification is equalled or exceeded in the analysis at all reservoirs, as indicated in Table 2.4-3.

The SPF peak flow for LGS of 250,000 cfs is estimated at 50% of the PMF, using the conservative criteria in Appendix B of Regulatory Guide 1.59 (Rev 2) for determining the PMF.

Landslides induced by seismic action could block the river, causing dam break type waves downstream when water pressure builds up and breaches the slide material. However, topographic considerations alone appear to preclude any hazard to the plant from landslide-induced waves. The Allegheny Mountains lie over 40 miles upstream from the site. Any landslides across the river in that area might temporarily back up water that could later breach the slide; however, by the time the flood wave appeared at LGS, it would be greatly attenuated. Below the Allegheny Mountains, in the Great Valley above Reading and the Triassic Lowland below Reading, the river's meander floodplain is typically 3000-5000 feet wide. Blockage of this wide channel by a landslide is not conceivable. The topography on both sides of the floodplain is generally gentle, but with some steeper hills occurring, particularly on the south side of the floodplain, between the site and Reading; however, a study of the USGS 1:24,000 maps indicates that, except at Reading, none of these appear to have the volume required to block the river.

At Reading, there are several steep hills on both sides of the river. Of these, the closest to the river and the steepest is Neversink Mountain, on the north side (Figure 2.4-14). The river is 200 feet wide and has no floodplain at this point. If Neversink Mountain were to collapse and block the river, with the slide subsequently breached by the river, water could back up to el 255' before spillage occurs over Poplar Neck, a long ridge on the south bank. The river bottom here is at about el 170';

therefore, the maximum depth of the water backup would be 85 feet, after which it would spill over Poplar Neck. The resulting wave could not produce a water surface elevation at LGS that would endanger the plant. The joint occurrence of a wave at LGS from a Neversink Mountain slide and a failure of one or more upstream dams is an event much less likely than is required by Regulatory Guide 1.59. The peak flow rate that is due to simultaneous failure of the three dams, Maiden Creek, Ontelaunee, and Blue Marsh, is conservatively estimated at 762,000 cfs. When superimposed on the SPF flow of 250,000 cfs, the combined event results in a total flow rate of 1,012,000 cfs.

2.4.4.2 Unsteady Flow Analysis of Potential Dam Failures The dam break problem has been discussed in the literature for various idealized cases. The most generalized and convenient approach developed to date is by Sakkas (Reference 2.4-23) as most recently described by Sakkas and Strelkoff (Reference 2.4-24). With this method, cognizance is CHAPTER 02 2.4-23 REV. 14, SEPTEMBER 2008

LGS UFSAR given to channel friction and volume of the flood wave. However, it was derived by dry-bed conditions and requires consideration of a uniform prismatic channel downstream.

In Reference 2.4-23, dimensionless graphs are given that are determined by numerical integration of the equations for unsteady flow, using the method of characteristics. A wide variety of combinations of dam height, channel slope, and cross-sections is analyzed so that the various graphs permit an analysis of many practical dam failure situations. Channel geometry is assumed as prismatic, and procedures are given in Reference 2.4-23 for deriving the constants characterizing the channel cross-section, yielding a relation between the cross-section width B and the depth Y. This relation is given by:

B CY m (EQ. 2.4-7) where:

B = the width at depth Y C and m = constants derived from the cross-section properties.

To determine the constants C and m, it is necessary to have the values of two pairs of B's and Y's (say Bo, Yo, and B1,Y1) of this representative cross-section. Thus paired B's and Y's were obtained by taking the geometric mean of paired b's and y's at three actual river cross-sections using:

b i 1

3 3 B i 1 1

3 3 Y

i1 y i and b1i 1

3 3 B1 i1 y 1i 1

3 3 Y1 i1 in which (i) denotes the individual of the three river cross-sections, at the hypothetical dam, the LGS site, and near Birdsboro, respectively. Because these paired b's and y's determine the (C) and (m) values and thus characterize the shape of the representative cross-section, they are called "paired characteristic breadth and depth dimensions".

Reference 2.4-23 gives dimensionless curves of the time of arrival of the wave front, the flood depth, and its time of arrival as a function of the distance downstream, empirical cross-section constants (C) and (m), and the initial Froude number (F) at the dam:

CHAPTER 02 2.4-24 REV. 14, SEPTEMBER 2008

LGS UFSAR m 1 V C C Y ,F 1

2 gY where:

2 1 1.486 R o S 3

=

2 Vo n

Ro = The hydraulic radius at the instant of failure Yo = the initial depth at the dam g = acceleration of gravity S = channel slope n = Mannings friction coefficient for channel downstream Symbols with bars over the top are dimensioned variables, and symbols without the bar are dimensionless, following Sakkas' notation.

For the assumptions specified, the results of the Sakkas procedure are considered reasonably accurate. However, adapting the general model to a particular situation and then superposing multiple dam failures requires careful consideration and interpretation to obtain conservative estimates of incremental and total flood depth. A set of simplified surrogate regimes is assumed whose cumulative effect on flood depth certainly is greater than the depth in the actual hydraulic regime that would prevail in such an event. This conservatism is justified only because of the LGS location high above the Schuylkill River.

To incorporate the prismatic channel assumption, it is necessary to adopt a single representative channel cross-section between the failed dam and the site. The river section at a dam is typically narrow, whereas the river floodplain below tends to be wider. In the Schuylkill River and its tributaries, the basic valley is U-shaped or V-shaped; i.e., there are no relatively wide floodplains in the reaches between the dams and LGS, and the main stream channel does not vary greatly in basic geometry, particularly under the initial condition of the SPF. Under these conditions an approximation of a constant prismatic section is considered realistic.

A single representative cross-section is defined by taking the geometric mean of two, paired characteristic breadth and depth dimensions at three cross-sections: one at the hypothetical dam described later; one at LGS; and one at a section of the Schuylkill River located about 1 mile upstream of the town of Birdsboro, or about 15 miles upstream of LGS. These three cross-sections are presented in Figures 2.4-21, 2.4-22 and 2.4-23 respectively. Figure 2.4-24 is the defined representative cross-section. The geometric mean section is then used for estimating the effect of a particular dam break.

The results given by the Sakkas procedure apply to a flood wave progressing down an initially dry-bed. To approximate this condition in the analysis, the channel slope and the geometry of the two downstream sections are set before a mean section is determined. They allow for the existence of the SPF and any stage increments from previous dam breaks. In effect, the peak of the SPF is "frozen" at 250,000 cfs at LGS for the time that the peak of a dam break flood wave is enroute to LGS. Conservatism is embedded in this approach, since an earthquake of multiple dam break failure CHAPTER 02 2.4-25 REV. 14, SEPTEMBER 2008

LGS UFSAR magnitude is more likely to be a single massive event dispensing flood waves from each dam at the same time, rather than timed so that their peaks would be additive at LGS, simultaneously with the crest of the SPF.

Since Maiden Creek and Ontelaunee dams are located on the same stream within 5 miles of each other, the SPF peak at Ontelaunee is estimated from that at Maiden Creek, using the ratio of drainage areas. This gives the SPF peak at Ontelaunee as 58,400 cfs. If it is conservatively assumed that there is no attenuation that is due to the storage in Ontelaunee reservoir above the spillway crest (el 294'), the SPF peak of 58,400 cfs discharges at a reservoir elevation of 304.2 feet.

It is estimated that, at this elevation, the volume of water impounded behind Ontelaunee Dam would be 29,000 acre-feet.

Based on the information available from the U.S. Army Corps of Engineers, the volumes of water stored in Maiden Creek and Blue Marsh reservoirs under SPF conditions are estimated as 130,000 and 76,700 acre-feet, respectively. Thus, the total storage of the three dams combined together is 235,700 acre-feet.

To obtain a conservative water surface elevation that is due to simultaneous failure of the three dams, it is assumed that the total volume of water (235,700 acre-feet) would be stored upstream of a hypothetical dam located near Reading, approximately 30 miles upstream of LGS and 5 miles downstream of Blue Marsh Dam. As the hypothetical dam is downstream of the three real dams, the channel storage assumed available for attenuation of a flood wave released from the hypothetical dam would be smaller than actual. This results in a higher, and therefore conservative, estimate of the water surface elevation at LGS.

The area-capacity characteristics of the hypothetical dam are assumed to be the same as at Blue Marsh Dam, which is the nearest upstream dam. Actually, the channel section at the location of the hypothetical dam would be wider than that at Blue Marsh, and so the above assumption provides conservative estimates of the water surface elevation corresponding to a given storage. By this assumption, the water surface elevation in the hypothetical reservoir for the combined storage of 235,700 acre-feet is estimated to be 333.5 feet. The river bed elevation at this location is 213 feet.

Thus the depth of impoundment of the hypothetical dam at the instant of the postulated failure was 120.5 feet.

Using a slope-area method and assuming uniform flow conditions, the water surface elevations corresponding to the SPF at Birdsboro and LGS are estimated to be 188 feet and 152 feet, respectively. To simulate pre-existing SPF conditions in the river, cross-sections at Birdsboro and LGS are assumed to be represented by the portions of the flow sections above the SPF elevations.

For the cross-section just downstream of the hypothetical dam, dry channel conditions are assumed.

With the Sakkas method, it is conservative to ignore any water in the channel immediately downstream of a failing dam because a deeper than actual reservoir storage results. As stated earlier, to use Sakkas' dimensionless charts, a representative prismatic section for the routing reach from the hypothetical dam to LGS is obtained by taking the geometric mean of the characteristic breadth and depth of the river sections at the hypothetical dam (above dry-bed), at Birdsboro (above SPF elevation), and at LGS (above SPF elevation).

To be conservative, the channel slope between the hypothetical dam and LGS is computed using the dry-bed elevation at the hypothetical dam site and the SPF elevation at LGS. The actual channel slope during SPF conditions is steeper than that used.

CHAPTER 02 2.4-26 REV. 14, SEPTEMBER 2008

LGS UFSAR To obtain a representative value of the coefficient of roughness, (n), for the river reach between the hypothetical dam and LGS, weighted average values are computed for the sections at Birdsboro and LGS using the equation (Reference 2.4-33):

AR 2

3 R

Ai i 2

n 2 3 (EQ. 2.4-8) i l ni where:

A = total flow area R = hydraulic mean depth for the entire section A1,R1 = flow area and hydraulic mean depth for the channel section A2,R2 = flow area and hydraulic mean depth for overbank flow n1,n2 = coefficients of roughness for the channel and overbank sections For the section at the hypothetical dam site, where dry-bed conditions are assumed, a value of n =

0.062 is used for the entire section. This value is recommended for floodplains with light-to-medium brush and trees (Reference 2.4-33) and conservatively represents the conditions at the hypothetical dam section. For the sections at Birdsboro and LGS, where the dam break flood wave is superimposed on the SPF, a value of 0.062 is used for the overbanks, and a conservative value of 0.03 is used for the interface between the water surface at SPF and the dam break flood wave. The arithmetic average of the (n) values for the three sections was assumed to be applicable for the entire reach.

Using Sakkas' curves (Reference 2.4-23), the maximum flood depth that is due to the dam break is determined at the downstream section at LGS. The calculated depth thus determined is a depth in the derived geometric mean cross-section. To relate the computed depth to the actual section, conveyance relations are used. Using the derived properties of the geometric section (C and m), the area is given by:

A = (CYm+1)(m+1)-1 (EQ. 2.4-9)

From this, the conveyance (Equation 2.4-4) required in both the geometric section and the natural section can be calculated. Using the stage conveyance characteristics of the natural river section above the SPF elevation at LGS, the water surface elevation that gives incremental conveyance (above SPF elevation) equal to that corresponding to the calculated depth in the geometric mean cross-section is computed. This results in a water surface elevation of 201 feet at LGS.

CHAPTER 02 2.4-27 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.4.3 Water Level at Plant Site 2.4.4.3.1 Maximum Computed Water Surface and Wave Effects This section describes the effect of wind-waves on the maximum water surface (el 201') estimated at LGS that is due to the dam break condition described in the previous section.

The dam break waves are transients and do not contain enough volume to cover the entire Schuylkill Valley above LGS to the computed maximum depth at LGS. However, for an approximate wind-wave analysis, the following conditions are assumed; the water surface at LGS is el 201', and wind velocity is 40 mph.

Fetch is based on a map study of a level pool upstream at el 201' At this water surface elevation, the ridges and the two bends between Pottstown and the site are submerged. This permits a roughly rectangular fetch of 31,000 feet average length and a width of 4,500 feet, on a bearing of about N 72 W. The centerline of the rectangle intersects a ridge about a mile west of the village of Stowe. It is conservative to assume a rectangular fetch area, which according to Reference 2.4-25 gives a fetch effectiveness ratio of 0.18 and an effective fetch of 2.65 miles. Using Reference 2.4-25, these parameters yield a significant wave HsV (33% frequency) of 3.8 feet, with a period of 3.9 seconds and a length of 78 feet. The maximum wave (1% frequency) is estimated as 1.67Hs = (1.67)(3.8) =

6.4 feet.

At the LGS site bench, the maximum water surface without wind action, el 201', intersects a sloped-fill surface to the windward (west) side of the plant (Figure 2.4-4). The fill surface has an average slope of 1 vertical to 2 horizontal. The toe of the slope varies between el 170' and el 200'. The top of the fill varies from el 213' at the northwest end (adjacent to the switchyard bench), to el 216' on the south side of the plant, where the top of the fill curves toward the east. From the top of this fill, the ground rises gently on a long flat slope to the roadway at el 215' to el 217', which encircles the west side of the turbine-reactor area complex. The roadways low part (el 215') is90-350 feet from the top of the slope. The fill material is broken rock and fines, with sizes up to 24-30 inches. For computing wave run-up, this material can be considered as graded riprap. To compute run-up, curves given in figures 7-19 of Reference 2.4-25 are used, which are for a 1:2 slope. However, the wave from the maximum fetch direction is not perpendicular to the 1:2 slope surface, but intersects it at an angle of about 64 from normal. This results in an effective slope normal to the wave of about 1:4.5, so that actual run-up is less than computed.

Run-up (R) from the significant (33%) wave and the maximum wave (1%) is calculated from Reference 2.4-25 as follows:

Wave Ho=Height (ft) Ho/gT2 R/Ho R (ft) Max el Run-up (ft)

Significant 3.8 0.0078 1.3 4.9 206 Maximum 6.4 0.013 1.0 6.4 207 Thus, the highest water surface elevation at LGS that is due to the most severe dam break permutation coincident with wave activity induced by a 40 mph wind would be 207 feet. This is 10 feet below the plant grade (el 217') and 8 feet below the roadway (el 215') .

CHAPTER 02 2.4-28 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.4.3.2 Recapitulation of Conservative Steps in Dam Break Analysis The computed water surface elevations (with and without waves) resulting from the dam break analysis are the result of a compounding of conservative assumptions.

The series of conservative steps used to compute the water surface are listed below.

a. The SPF is estimated at 50% of PMF, as determined from Appendix B of Regulatory Guide 1.59 (Rev 2). The PMF, as estimated by this procedure, has a peak of 500,000 cfs (unadjusted for dam breaks), whereas the peak PMF computed by conventional methods in the PSAR was 356,000 cfs, approximately 30% less. The SPF peak at 50% of the Appendix B PMF, or 250,000 cfs, is thus also conservative.

Conservatism is further confirmed by a published U.S. Army Corps of Engineers estimate (Reference 2.4-26) of 128,000 cfs for the SPF at Pottstown; adjusted to the site, this would be 131,000 cfs, 52% of the SPF given by Appendix B of Regulatory Guide 1.59 (Rev 2).

b. The effective cross-sectional area used at LGS omitted part of the left overbank area (looking downstream) to conservatively allow for local flow separation produced by the river bend.

c. The earth-fill dams should not fail instantaneously as is assumed.

d. A dry channel is assumed downstream of the hypothetical dam, resulting by a larger effective reservoir depth by using Reference 2.4-23.

e. The concept of the hypothetical dam implies seismic failures of the three real dams generated at different times, corresponding to three different travel times to LGS.

This is an improbable series of selectively destructive tremors. It is more likely that a single catastrophic tremor would be involved in a multiple dam failure, with the resulting flood waves arriving at LGS at different times.

f. The actual channel slope in the river reach from the hypothetical dam to LGS is steeper than the slope used in the analysis. This results in a higher water surface elevation at LGS, computed by Sakkas' procedure.

Since all of these assumptions are conservative, it is concluded that the maximum stages computed with and without waves are well above the stages that a precise analysis would indicate. It is finally concluded that the most severe seismic dam break permutation of the three dams, Blue Marsh, Ontelaunee, and Maiden Creek, would not endanger safety-related structures. The simplified analysis is justifiable because the plant area is high above the Schuylkill River.

2.4.5 PROBABLE MAXIMUM SURGE AND SEICHE FLOODING This section is not applicable to LGS.

2.4.6 PROBABLE MAXIMUM TSUNAMI FLOODING This section is not applicable to LGS.

CHAPTER 02 2.4-29 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.7 ICE EFFECTS This section is not applicable to LGS. Spray pond icing is discussed in Section 9.2.6.

2.4.8 COOLING WATER CANALS AND RESERVOIRS In this section, only the hydrologic engineering aspects of the spray pond components of the UHS are covered. There are no canals in the cooling water system. The spray pond serves as the UHS for the RHRSW system and the ESW system after a possible accident. The pumps of the two systems take water from the spray pond, and circulate it through coolers and heat exchangers. The warm water is returned to the spray pond through a network of spray nozzles. A complete description of the spray pond system is given in Section 9.2.6.

2.4.8.1 General Description of the Spray Pond The spray pond is located about 500 feet north of the cooling towers. The bottom of the pond is at el 241' and is composed of a 600x400 foot rectangular midsection, with a semicircle (radius = 200 feet) on each side. The spray pond system is common to both Units 1 and 2. The system consists of a spray pond, uncontrolled emergency spillway, pump structure complex, and associated piping and valves.

As shown in Figure 2.4-5, the spray pond is constructed by excavation only. The slopes of the excavation are 1:1 in rock and 4:1 in soils. Around most of the spray pond, random compacted fill about 3 feet deep overlies the original natural ground, with a bench at the soil-rock interface. The water surface area of the pond at el 251' is approximately 9.9 acres. An additional 7.3 acres of the surrounding area, including roads, cut surfaces, and natural terrain, drains towards the pond. Run-off onto the cut face is normally intercepted by a drainage ditch along the outside edge of a peripheral service road at el 255' and is directed to culverts that discharge to the pond. Along the north edge of the pond, the finished roadway is constructed to el 252' for a distance of 60 feet, with 9% slopes upward at either end to el 255'; this low portion of the roadway is designed to function as the crest of an uncontrolled emergency spillway. Spill is directed to a draw that drains northward into Sanatoga Creek.

2.4.8.2 Hydrologic Design Bases Derivation of the PMF that forms the basis for the hydrologic design of the spray pond and emergency spillway is discussed in Section 2.4.2.3.2. The emergency spillway is sized so that the design flood will be evacuated safely. The elevation of the roadway around the pond and the slope protection of the pond are dictated by anticipated wave action.

CHAPTER 02 2.4-30 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.8.2.1 Design Basis Flood Level To arrive at a conservative elevation that is due to severe floods, with coincident wind-wave activity in the spray pond, the following cases were analyzed for waves, freeboard, and slope protection, using the procedures given in Reference 2.4-25.

Still Pond Max Water Case Flood Surface el (ft) Maximum Wind (mph) 1 PMF 253.0 40 2 1/2 PMF (SPF) 252.5 90 Results are as follows:

Significant (33-1/3%) Maximum (1%) Maximum Wave Run-up Case Wave Height (ft) Wave Height (ft) ft el 1 0.8 1.3 1.1 254.1 2 2.0 3.3 2.4 254.9 Based on the above, the DBFL is set at el 253.9'. The roadway surface has been set at el 254', or 0.1 foot higher than the DBFL.

Using criteria in Reference 2.4-25, the minimum riprap stone requirements are: a minimum weight of 7 pounds, a 50 percentile weight of 30 pounds, and a maximum weight of 108 pounds (assuming a stone density of 165 lb/ft3). The gradation and design of the riprap for the spray pond soil slopes are discussed in Section 2.5.5. The riprap is capable of resisting the wave action and therefore protects the soil slopes. No protection is necessary for rock-cut slopes.

2.4.8.2.2 Safe Shutdown and Operating Basis Earthquakes According to Regulatory Guide 1.59, the higher of the following two alternative combinations of events is considered to be an adequate design basis for floods that are due to seismic failure of dams:

a. Alternative 1 - 25 year flood coincident with SSE and 2 year extreme wind speed from the critical direction and length of effective fetch

b. Alternative 2 - one-half PMF, coincident with OBE, and 2 year extreme wind speed from the critical direction and length of effective fetch When routed through the spray pond, the SPF yields a maximum water surface el 252.5'. The 25 year flood yields a maximum water surface el 251.8'. The starting water surface elevation in the spray pond for both these cases is assumed to be the normal operating water surface elevation (el 251'). In the latter case, the entire volume of the 25 year, 24 hour2.777778e-4 days <br />0.00667 hours <br />3.968254e-5 weeks <br />9.132e-6 months <br /> precipitation, without any losses over the contributing drainage area of 17.2 acres, is superimposed on the normal pond el 251'.

Because the flood is contained below the spillway crest (el 252'), flood routing computations are not required for this case.

CHAPTER 02 2.4-31 REV. 14, SEPTEMBER 2008

LGS UFSAR To estimate the height of waves due to earthquakes, the following equation, developed for determining the height of waves generated by a piston-type wave generator (Reference 2.4-31) is used:

2 sinh2 (2d)

H 2s L (2 d) sinh (2 d) cosh (2d)

L L L (EQ. 2.4-10) where:

H = wave height vertical distance between the wave crest and trough (feet) s = design displacement (amplitude) caused by the earthquake (feet) d = initial depth of water (feet)

L = wave length; a function of the period of the design basis earthquake (feet)

From the response spectra of the SSE and OBE, the design displacement (amplitude) and period are estimated to be 6.5 inches and 2.0 seconds for the SSE and 3.0 inches and 2.0 seconds for the OBE, respectively. Conservatively, assuming that a negative wave is generated at the opposite boundary of the spray pond and that the amplitudes of the positive and negative waves are in phase and additive, the maximum possible wave height would be 2H. This results in a maximum water surface elevation H above the still water. The computed values of H and the resulting water surface elevations are tabulated below:

Concurrent Earthquake Wave (ft) 2 Yr Wind Max Water Surface Earthquake Flood H 2H Effect (ft) el (ft)

OBE SPF 1.0 2.0 0.25 253.8 SSE 25 year 2.1 4.2 0.25 254.2 Because the spillway crest is at el 252', there would be some splash over the spillway for a short duration. However, water loss due to such splashes would be negligible compared to the total capacity of the spray pond and would not impair the safety-related water supply functions of the pond.

For the design of riprap protection, the wind-generated waves described in Section 2.4.8.2.1 are more critical than the short duration earthquake-induced waves.

For the spray-head pipe supports and the pump structure, hydrodynamic forces due to the OBE and SSE are computed. The forces on the pipe supports are computed using the virtual mass formula, F

= CmVa (Reference 2.4-30), in which (F) is the hydrodynamic force on a body of submerged volume (V) that is due to an acceleration (a), in a fluid of mass density (. Based on information available in Reference 2.4-30, a value of 1.5 is used for the coefficient (Cm). The resulting forces on the spray-CHAPTER 02 2.4-32 REV. 14, SEPTEMBER 2008

LGS UFSAR head pipe supports are 431 lbs and 890 lbs for the OBE and SSE, respectively; the forces are assumed to be applied at mid-depth, el 247.3' and el 247.5', respectively.

For the pump structure, hydrodynamic forces are computed both in the N-S and E-W directions. For the N-S direction, the TVA method (Reference 2.4-31) is used. For the E-W direction, where only part of the structure is exposed to water, the method given by Sarpkaya (Reference 2.4-32) is used.

The resulting forces and their points of application are given below:

Earthquake Direction Max Force (kips) Point of Application (ft)

SSE N-S 144.0 246.2 E-W 70.0 247.5 OBE N-S 67.0 246.0 E-W 34.0 247.5 These hydrodynamic forces on the pipe supports and the pump structure do not include the hydrostatic forces that are due to normal water depth, flood surcharge, or earth pressures, nor do they include dynamic forces that are due to waves.

2.4.8.3 Low Level Outlet Facilities Evacuation of the normal storage of the spray pond, if needed, is accomplished by using the ESW system and/or RHRSW system pumps to pump water to the cooling tower basins via the cooling tower spray pond intertie line down to the minimum operation level of the pumps, and by other means below that level.

2.4.9 CHANNEL DIVERSIONS This section is not applicable to LGS.

2.4.10 FLOODING PROTECTION REQUIREMENTS As discussed in Section 2.4.2.2, the safety-related structures and facilities are secure from flooding.

Hence, flooding protection requirements are not necessary.

2.4.11 LOW WATER CONSIDERATIONS Extreme low flow in streams does not affect the ability of any safety-related facilities to perform adequately, including the UHS, as discussed in Section 9.2.6.

Availability of nonsafety-related water supplies is governed by the Delaware River Basin Commission (DRBC) Docket Decision D-69-210CP. The DRBC also has exclusive jurisdiction over the necessity for and approval of compensating water storage capacity for the LGS. LGS shall operate its nonsafety-related water supply systems in accordance with the terms and conditions imposed by the DRBC.

CHAPTER 02 2.4-33 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4.12 DISPERSION, DILUTION, AND TRAVEL TIMES OF ACCIDENTAL RELEASES OF LIQUID EFFLUENTS IN SURFACE WATERS There are no credible accidents that can occur that result in an accidental liquid release of radioactive effluents to surface waters.

The term maximum permissible concentration (MPC) is used in various sections of the UFSAR. The term MPC is reflective of pre-1994 10CFR20, Apprendix B, Table II limits. These limits were used for the original licensing basis of the plant. Current liquid effluent releases are limited to ten-times the Effluent Concentration Limit (ECL) for each isotope specified in post-1994 10CFR20, Appendix B, Table 2, Column 2 (See Radioactive Effluent Release Controls Program required by the Technical Specifications).

The radwaste and reactor enclosures are seismic Category I structures. Each structure has a floor drain collection system to collect any potential liquid spills that may occur and to transfer the liquid to radwaste tanks. The radwaste tanks are located at the lowest elevation in the radwaste enclosure, which has a leak-tight membrane from the floor to el 174' MSL (grade is el 217' MSL).

Consequently, any effluents accidentally spilled inside these enclosures are contained and remain isolated from the surface water environment.

In addition, the release of radioactive liquids by the failure of an outside storage tank is not considered credible since the tanks are enclosed by retaining dikes (Section 9.2.7). These dikes are constructed of compacted fill. The dikes and the area between the tanks and the dikes are covered with asphalt pavement. The outside tanks that contain radioactive liquid are the Unit 1 and 2 CSTs, each having a capacity of 200,000 gallons, and the refueling water storage tank, with a capacity of 500,000 gallons.

The Unit 1 CST and the refueling water storage tank are surrounded by a retaining dike. The Unit 2 CST is also surrounded by a retaining dike (Section 9.2.7). These dikes are designed to contain 110% of the contents of the largest tank that they enclose. Therefore, even if the entire contents of the largest of these tanks is spilled, the effluents are contained within the diked area, and do not reach any source of surface water.

The retaining dikes surrounding the outside CSTs are designed in accordance with seismic Category IIA requirements (Section 3.2.1). However, an analysis to determine the potential for contamination of nearby surface waters assuming that both tanks and dikes fail is not necessary for the following reasons:

SRP 15.7.3 indicates that the technical specifications shall limit the contents of outside storage tanks to those levels that will not exceed 10CFR20 concentrations at the nearest potable water supply unless dikes are provided to limit the water released by a tank or associated component failure or overflow. LGS has provided a dike to limit these water releases.

Based on section C.1.p of Regulatory Guide 1.29, the CSTs and surrounding dikes have been classified as seismic Category II and IIA respectively because the airborne doses from a failure of these structures would not exceed 0.5 rem whole body or its equivalent to any part of the body (1.5 rem thyroid in this case). Thyroid doses of 2.1x10-3 rem have been calculated at the EAB, based on a 350,000 Ci/sec offgas release rate resulting in the tank inventories of Table 12.2-87, an iodine dose factor of 10, and a 5 percentile X/Q of 2.9x10-4.

Although the retaining dikes are not seismic Category I, they will retain their structural integrity during an SSE event as discussed in Section 2.5.5. In view of this, failure of both tanks and dikes need not CHAPTER 02 2.4-34 REV. 14, SEPTEMBER 2008

LGS UFSAR be assumed in determining the potential for contamination of nearby surface waters. An analysis is provided in Section 2.4.13 which assumes a radwaste tank failure only. The analysis indicates that significant periods of decay (over 600 years) will occur before any radioactivity is released via the groundwater pathway if the contents should be absorbed by the ground. Because the primary contributors to the normally expected CST activity are the iodines and several shorter-lived isotopes with half-lives of less than 1 day, and all other isotopes have expected concentrations less than the MPC levels for water in unrestricted areas, doses via any food or drinking water pathways will be well below 10CFR20 limits.

Accidental release of radioactive effluents into the groundwater environment is discussed in Section 2.4.13.3.

In view of the above, detailed analyses regarding the ability of the surface water environment to disperse, dilute, or concentrate accidentally released radioactive effluents are not required.

2.4.13 GROUNDWATER Investigation of regional and local groundwater conditions indicates that the construction and operation of the LGS has no adverse effects on the groundwater resources in the region and the site.

2.4.13.1 Description and Onsite Use The site is located in the Triassic Lowland Section of the Piedmont Physiographic Province. The region is characterized by rolling hills and long low ridges and is well drained by numerous tributaries to the Schuylkill River. Precipitation averages about 44 inches per year, and run-off averages about 21 inches per year. The balance represents mainly evaporation and transpiration losses, with minor infiltration.

Groundwater occurs in sedimentary rocks of Triassic-age. The Brunswick lithofacies, the aquifer underlying the LGS site, yield small-to-moderate quantities of water to wells.

The water supply for LGS plant operation is obtained from surface water, and no groundwater use is planned.

2.4.13.1.1 Groundwater Aquifer Groundwater in the region occurs in sedimentary rocks. The region is underlain by the Newark group of Triassic-age, which includes the Stockton Formation and the overlying Lockatong, Hammer Creek, and Brunswick lithofacies.

The strata are intruded by diabase dikes and sills. Although the other units provide some groundwater in the region, the Brunswick lithofacies are the only aquifer of significance at the LGS site.

The Stockton Formation, which crops out about five miles south of the site, is composed of interbedded sandstone, conglomerate, red shale, and red siltstone and is at too great a depth at the LGS site to be of hydraulic importance. The Hammer Creek conglomerates and sandstones stratigraphically overlie the Stockton in the region but are not present in the site area. The nearest occurrence of the Hammer Creek is a few miles west of the site, in Berks County. The Stockton is overlain south of the site by the Lockatong, a dark gray argillite. The areal distribution of these formations and lithofacies is shown in Figure 2.5-2.

CHAPTER 02 2.4-35 REV. 14, SEPTEMBER 2008

LGS UFSAR The Brunswick, the aquifer at the site, is composed of red shale, sandstone, and siltstone. Within the Triassic Lowland, as well as in the site region, the Brunswick is the most widespread source of groundwater. The Brunswick is interbedded with thin tongues of Lockatong, and changes in lithology are common.

The dip of the Brunswick and Lockatong strata is to the north and northwest at about 10 to 20.

Several broad synclines and anticlines trending about N 60 W are superimposed on the regional structure. The strata have been cut by many faults, most of which are relatively small. Joint systems are well developed and strike N 20 E to N 50 E and about N 50 W to N 90 W. The joints are mostly vertical, and in some strata are closely spaced.

The Brunswick is composed of very fine-grained rocks. The pore spaces within the rock matrix are very small and offer great resistance to the flow of groundwater. Therefore, permeability that is due to the primary porosity of the Brunswick is small.

Most of the groundwater movement within these rocks follows secondary openings that developed following the deposition of the beds. Some of these openings are fractures that parallel the bedding-planes. They are usually narrow and probably contribute little to the permeability. The most important openings are nearly vertical joint planes; they cross each other at various angles throughout the beds. These joints, where present, provide an interconnected series of channels through which groundwater can flow, giving the material a low to moderate permeability (Reference 2.4-36).

The number and width of secondary openings vary; consequently, the permeability differs from one bed to another. In a series of beds 100 feet thick, there may be only a few beds in which the secondary openings are well developed. These beds range in thickness from a few inches to a few feet; the average thickness is about 2 feet (Reference 2.4-36).

In the Brunswick, the water table is found at shallow depths. Wells that penetrate deep strata in the Brunswick may encounter water under confined conditions. The water table slopes from the topographic divides to discharge areas along streams and rivers. The Brunswick is generally a reliable source of small-to-moderate quantities of groundwater but, yields from wells that penetrate it vary widely because of lateral and vertical variations in lithology, the occurrence of lens-shaped beds, uneven spacing of joints and locally complex structure. Fault zones in the Triassic rocks have been found to be barriers to the flow of groundwater, and wells located near them generally have very low yields. The median yield of drilled municipal and industrial wells is about 110 gpm. Yields in excess of 300 gpm are rare and are obtained from wells that intersect a larger number of water-bearing zones. The median transmissivity of the Brunswick is 1100 gpd/ft (Reference 2.4-36).

Most groundwater in the Brunswick aquifer is a calcium bicarbonate-type, but some of the groundwater, with a total dissolved solids content in excess of 500 ppm, is a calcium sulfate-type (Reference 2.4-36). Water quality data for the Brunswick Lithofacies are presented in Table 2.4-14.

Most water from the aquifer is suitable for domestic and other purposes.

2.4.13.1.2 Aquifer Recharge and Discharge Recharge to the Brunswick occurs through the soil cover as precipitation percolates down to the water table. The water table generally follows the surface of the land, and groundwater flows from high to low topographic areas.

CHAPTER 02 2.4-36 REV. 14, SEPTEMBER 2008

LGS UFSAR Poor water quality and low yields in wells deeper than 600 feet indicate little groundwater movement below that depth. Most groundwater movement is prevalent only in the upper portion of the Brunswick, where the fracture density is greatest.

At the site, groundwater moves from the topographic high northeast of the plant structures down gradient to the north and the southwest, where it eventually discharges into the Schuylkill River and Possum Hollow Run (Figure 2.4-15).

2.4.13.1.3 Onsite Use of Groundwater Water for plant operation is supplied by surface water and is discussed in Section 2.4.

Three wells were developed : Well No. 1 is 310 feet deep, 6 inches in diameter, and yields 50 gpm; Well No. 3 is 585 feet deep, 7-7/8 inches in diameter, and yields 65 gpm; and Well No. 4 is 198 feet deep, 6 inches in diameter, and yields 65 gpm. Note that well No. 4 was decommissioned and is no longer in service for domestic water.

These wells will be utilized as permanent water sources as follows:

º Site well 1, source of chlorinated drinking water.

º Site well 3, water to batch plant water tank and 500,000 gallon backup fire water storage tank.

The operation of the wells has no permanent impact on the groundwater resources of the site area.

2.4.13.2 Sources Within 20 miles of the site, surface water sources supply most of the public and commercial demand.

The Schuylkill River and its tributaries are widely used. However, a number of municipalities and industries, together with many domestic users, utilize groundwater.

2.4.13.2.1 Regional Use of Groundwater The locations of public water supply systems within 20 miles that use groundwater are shown in Figure 2.4-16. An inventory of the supply systems is presented in Table 2.4-15. Public water systems supply more than 80% of the population in the region (Reference 2.4-37). Groundwater contributes less than 40% of the total amount of public water supplied.

Individual, private wells supply groundwater to slightly less than 20% of the population in the region and generally account for less than 10% of the total water usage (Reference 2.4-37).

2.4.13.2.2 Use of Groundwater in the Vicinity of the Site An inventory of local groundwater users is presented in Table 2.4-16. Locations of groundwater users within 1 mile of the site are shown on Figure 2.4-17.

Groundwater supplies several homes, farms, and commercial interests in the vicinity of the site.

Most of the wells are less than 200 feet deep and yield less than 100 gpm. The largest groundwater user, Publicker Industries, is 11/2 miles south of the site and uses 150,000 gpd from three wells in the Brunswick.

CHAPTER 02 2.4-37 REV. 14, SEPTEMBER 2008

LGS UFSAR Water wells in the site vicinity are either not in the same groundwater basin as the plant (hydrologically separate) or are hydraulically up-gradient of the plant. No wells to which groundwater beneath the plant discharges are present in the area.

2.4.13.2.3 Projected Future Use of Groundwater Population growth and relocation will create a corresponding need for larger supplies of water in the site region. Much of the increase in water usage will take place in Limerick Township and neighboring townships as suburban development creates new demands for groundwater (Reference 2.4-37). Based on the assumption that the usage in the region reflects the national average, water usage is expected to increase by approximately 1 gcd (gallon per capita per day) each year. By the year 1990, a 12% gcd increase, combined with more than a 30% increase in people served by public water systems, will increase the total amount of water use by approximately 50% (Reference 2.4-37).

The increase in the demand for water will be met mostly by public water supply systems, which are developing additional wells to meet part of the demand.

The number of domestic and private wells is not expected to increase significantly, because most new water users will purchase water from water supply companies. By 1990, more than 90% of the population in the region will be supplied by public water suppliers.

Large increases in the use of groundwater in the site area are precluded by the low permeability and limited area of the water-bearing units there.

LGS will have no effect on the future use of groundwater in the region. The plant uses groundwater from site wells 1, 3 and 4 for drinking water and for filling the 500,000 gallon backup fire water storage tank.. The site is hydrologically isolated from all public groundwater supplies and areas of extensive groundwater development.

2.4.13.2.4 Water Levels and Groundwater Movement Groundwater levels were measured in borings (observation wells) at various locations near the spray pond and power block areas. The observation wells were monitored at regular intervals from June 1973 through August 1975, and from November 1978 through May 1979, to determine the direction of groundwater movement and the elevation of the water table, and to establish structural design criteria for groundwater effects. These wells were subsequently monitored throughout the remaining plant construction period to determine the effect of plant construction on the groundwater movement and levels, and to verify/assure the adequacy of the design criteria used in the design of safety-related plant structures. Figure 2.4-18 shows hydrographs of the observation wells.

2.4.13.2.4.1 Observation Wells A network of 22 groundwater observation wells were installed at the site at various construction stages to determine the depth to water, and to measure groundwater level fluctuations. The locations of these wells are shown in Figure 2.4-15. The construction data (depth of hole, interval screened, etc.) are presented in Table 2.4-17.

Five observation wells were installed in the planned spray pond area in June 1973, and monthly water level readings commenced. In June 1974, four more wells were installed in the spray pond area and incorporated in the monitoring program. Monitoring was temporarily discontinued between August 1975 and November 1978 because of construction of the spray pond. Observation wells P2, SP20, and SP21 were destroyed by this construction.

CHAPTER 02 2.4-38 REV. 14, SEPTEMBER 2008

LGS UFSAR In November 1978, water level monitoring was resumed in the remaining six wells near the spray pond area. Furthermore, 10 more observation wells were installed in January 1979, and three more wells were installed in June 1981. Six of the 13 wells are located near the power block area, and the remaining seven wells are located near the spray pond area. Observation wells P1 and P4 were destroyed in 1981 by ongoing construction activities and grading. In addition, observation wells P9, P13, and P15 were abandoned in April 1988, June 1984, and December 1987, respectively.

The recorded water levels in the observation wells between early 1979 and early 1988 are shown on the hydrographs in Figure 2.4-18.

2.4.13.2.4.2 Groundwater Levels The depth to water in observation wells has ranged from 13-96 feet below ground surface. Recharge to the groundwater beneath the site is primarily from infiltration of precipitation on the ground surface.

During the period 1973 through 1975, precipitation was average for the region. Thus, groundwater levels during that period were probably near average.

The potentiometric surface determined from levels measured in May 1979, prior to spray pond construction, is shown in Figure 2.4-15. This indicates the water table east of the planned spray pond was at el 250' and decreased in elevation to less than 130 feet southwest of the power block.

Postconstruction water levels reveal the same general configuration, with localized variations probably due to site grading/surface and drainage installation.

Hydrographs of observation wells (Figure 2.4-18) indicate that water levels fluctuate seasonally, with the lowest levels occurring in the fall and early winter, and the highest levels occurring in the spring.

In the spray pond area, these records indicate a 17 foot maximum seasonal fluctuation (wells P5 and P6), with most well levels fluctuating less than 12 feet.

Neither average water levels nor water level fluctuations were significantly affected by the filling of the spray pond during October 1982 and November 1982. The only exception to this is the water level in well P17, which currently averages about 10 feet higher than it did prior to pond filling.

However, the water level fluctuation in well P17 remains as large as or larger than that recorded prior to pond filling, suggesting there is no hydraulic connection between well P17 and the water in the spray pond which is maintained at a constant level. In any event, measured seepage loss from the entire spray pond was calculated to be only 4.7 gpm (Reference 2.4-43).

Hydrographs of observation wells in the power block area (Figure 2.4-18) exhibit larger water level fluctuations than those in the spray pond area. These fluctuations were closely related to precipitation, as illustrated in Figure 2.4-19. Wells P12 and P15 do not show large fluctuations because the water table is below the screened intervals during low periods. When the water table drops below the screen, the level measured in the well is of the water remaining in the sump, a 5 foot length of casing below the screen. Thus, only the highest water levels reflect the water table (fluctuation peaks) at wells P12 and P15.

During construction, the observation wells in the power block area were adjacent to open trenches or to plant excavations backfilled with relatively permeable materials. Precipitation that collected in the open trenches or that infiltrated the permeable backfill provided abnormal amounts of recharge to the water table. When the open trenches were filled and asphalt or other relatively impermeable layer was placed over these backfilled areas, the water table fluctuation was reduced because of the reduction of infiltration of precipitation (Figure 2.4-18). The surfacing also resulted in large reductions CHAPTER 02 2.4-39 REV. 14, SEPTEMBER 2008

LGS UFSAR of average water levels in wells P13 and 14. Figure 2.4-15 shows potentiometric contours of the water table as of May 31, 1988.

2.4.13.2.4.3 Direction of Groundwater Flow Groundwater in the shale and siltstone strata immediately underlying the site flows in fractures and joints. Exploration at the site suggests the fractures are tight below a depth of 140 feet, indicating little or no movement of groundwater in these units below this depth.

The direction of flow is from topographically high areas to topographically low areas (Figure 2.4-15).

Groundwater beneath the plant flows southwesterly toward the Schuylkill River, eventually discharging to the river. North of the plant, a groundwater divide is present beneath the topographic ridge. Groundwater north of the divide flows northward, discharging to tributaries of the Schuylkill River. Because of the low permeability and poor hydraulic connection between beds in the bedrock, the influence of groundwater at the site on the regional hydrology is negligible.

Below the shale and siltstone strata described above, some discontinuous beds of sandstone yielded water to the wells used during construction. Pumping tests using these wells and closely spaced observation wells in the sandstone and the overlying siltstone-shale beds measured the hydraulic interconnection between these units. The results indicate that the connection ranges from very slight to negligible.

2.4.13.2.5 Aquifer Parameters A pumping test and other permeability tests were performed to evaluate the groundwater hydrology of the site.

For the pumping test, a 6 inch diameter test well, located near the center of the construction area, was installed to a depth of 185 feet. Changes in water level as a result of pumping were measured in 6 observation wells located at various distances (20 feet to 700 feet) from the test well. Using the time-drawdown data from the pumping test, and the Theis nonequilibrium formula, bedrock transmissivity (T) is computed to be 2250 gpd/ft. This value corresponds to an average permeability (K) of 11 gpd/ft2 (550 ft/yr).

Forty constant head tests were performed in auger holes and boreholes to estimate the permeability at the spray pond. Calculated permeabilities range from 4 ft/yr to 1247 ft/yr and are listed in Table 2.4-18. The average permeabilities for the overburden, the contact zone between the overburden and bedrock, and the bedrock are 3.5 ft/yr, 14 ft/yr, and 214 ft/yr, respectively. The tests indicate that the permeability of the overburden is lower than the bedrock's permeability. The greatest range of permeability values was determined from test data in bedrock. Because of the fractured nature of the bedrock, the permeability depends on the number of fractures the boring intercepts. More than 84% of the permeabilities measured in bedrock are less than 390 ft/yr.

2.4.13.2.6 Reversibility of Groundwater Flow Seepage from the spray pond may cause a groundwater mound beneath the pond, with minor, local reversals of flow direction. This would increase the groundwater flow to the north and to the Schuylkill River, but the general direction of flow would remain the same. Because the pond will be lined, these changes will be minor.

2.4.13.2.7 Water Quality CHAPTER 02 2.4-40 REV. 14, SEPTEMBER 2008

LGS UFSAR Chemical analyses of groundwater are available from wells that penetrate the Brunswick lithofacies in Montgomery County (Table 2.4-14). The median dissolved solids content of samples from these wells is 302 ppm, and the median hardness (as CaCO3) is 218 ppm. Groundwater from the Brunswick is largely of the calcium bicarbonate-type, although water samples having concentrations of dissolved solids greater than 500 ppm are of the calcium sulfate-type. Groundwater from the Brunswick is of good quality for most domestic and municipal uses, although it may need to be treated for hardness.

Water quality information is available from Well Nos. 1, 3, and 4 at the LGS site (Table 2.4-19).

Samples were taken from holding tanks rather than from open discharge at the wells, but are considered indicative of the water quality. The water is a calcium sulfate-type, with pH ranging from 7.5 to 8. The water is moderately hard, ranging from 134 ppm to 618 ppm as CaCO3, and contains total dissolved solids ranging from 199 ppm to 1052 ppm.

2.4.13.3 Accident Effects An accidental spill of radioactive effluents from the radwaste enclosure is hypothesized, and the effect of a spill is analyzed with regard to the nearest offsite source of potable water, the Schuylkill River, 800 feet from the radwaste enclosure. Potentiometric contours shown on Figure 2.4-15 indicate that, in the radwaste enclosure area, groundwater flows southwest toward the Schuylkill River. It is assumed that radionuclides from a spill would percolate directly to the water table in the rock and then move with the groundwater laterally toward the river. The radioactive effluents that would constitute a spill are assumed to originate from the waste sludge tank, which contains the largest source of radioactivity in the enclosure. For conservatism, radioactive effluents are postulated to immediately enter the water table below the radwaste enclosure, and concentrations of radionuclides at that point are assumed to be the same as those in the radwaste tanks. No credit is taken for ion exchange, dispersion, or radioactive decay during percolation to the water table.

In addition to the accidental spill of radioactive effluents from the radwaste enclosure hypothesized as the worst case spill, a feed water line break between the main condensers and the condensate filter demineralizer in the Unit 2 turbine enclosure could contaminate the groundwater supply to site well 4. Site well 4 is utilized as a source of chlorinated water at the site. Site well 4 is on the same side of the groundwater divide as the east side of the Unit 2 turbine enclosure therefore a radioactive spill in the Unit 2 turbine building could enter the ground directly beneath the building. This spill would then enter the groundwater and flow southeast of the groundwater divide approximately 600 feet to site well 4.

2.4.13.3.1 Groundwater Movement The rate at which the contaminants migrate is no greater than that of the groundwater in which it is carried. Groundwater travel time and direction of flow is controlled by permeability and effective porosity of the bedrock and by the hydraulic gradient of the water table. The average groundwater velocity and the travel time to the river may be calculated using a form of Darcy's Law:

CHAPTER 02 2.4-41 REV. 14, SEPTEMBER 2008

LGS UFSAR

= ki (EQ. 2.4-11) n tgw = xn (EQ. 2.4-12) ki where:

= average groundwater velocity (ft/yr) k = permeability (ft/yr) i = hydraulic gradient (ratio) n = effective porosity (ratio) tgw = travel time (yr) x = distance (ft)

A permeability of 390 ft/yr is selected as a conservative estimate of the effective permeability of the underlying materials, based on the 84th percentile of the permeability tests described in Section 2.4.13.2.5. No porosity measurements of site bedrock were performed, but the low permeability indicates that interconnected void spaces are few, resulting in a low effective porosity. An effective porosity of 5% is estimated for the bedrock.

The hydraulic gradient between the radwaste enclosure and the river is calculated as 0.031, based on the hydraulic head difference between the 130 foot elevation potentiometric contour just southwest of the radwaste enclosure (Figure 2.4-15) and the elevation of the river, 105 feet, divided by the distance between the river and the radwaste enclosure, 800 feet, which is assumed to be the flow path.

Using these values for the parameters in the above equations, the computed travel time for groundwater, traveling at an average velocity of 243 ft/yr, to flow from the radwaste enclosure to the river is 3.28 years.

2.4.13.3.2 Analytical Model for Radionuclide Migration The concentration of contaminants would be reduced during migration in the groundwater by the processes of ion exchange, dispersion, and radioactive decay. The total effect of these processes on the concentrations of the contaminants at the point where they would enter the Schuylkill River was determined by use of an analytical model (Reference 2.4-38). The model accounts for adsorption, one-dimensional longitudinal dispersion, and radioactive decay:

C/Co=1/2 exp(-t) erf x-uit -erf x-uit (EQ. 2.4-13) 2 Dm t 2 Dm t CHAPTER 02 2.4-42 REV. 14, SEPTEMBER 2008

LGS UFSAR where:

C = final concentration Co = initial concentration

= radioactive decay constant = 0.693/radioactive half-life t = time t' = (t-t) where t is the time necessary for the fluid to flow one pulse width (the length of spill volume) x = distance ui = average velocity of ion Dm = longitudinal dispersion constant 2

x t edt 2

erf = error function: erf(x) = o The waste sludge tank is assumed to be 80% full at the time of the spill. Thus, 10,240 gallons of radioactive effluent is assumed to be released. The spill is considered to enter directly into the groundwater. The volume of bedrock containing the effluent is assumed to be a cube. The length of one side of the cube would be 30 feet, based on an effective porosity of 5%. Of the radionuclides present in the waste sludge tank (Table 11.4-9), Sr-90 and Cs-137 are the most hazardous because of their long half-lives (29 years and 33 years, respectively) and their relatively high concentrations (2.2x10-3 c/ml). Because the concentration of Sr-90 is equal to Cs-137 and Sr-90 is adsorbed less than Cs-137, the movement of Sr-90 is analyzed as the worst case.

2.4.13.3.3 Radionuclide Ion Velocity and Travel Time The average velocity of Sr-90 in groundwater (Uion) is determined by the relationship (Reference 2.4-39):

Uw Ui (EQ. 2.4-14)

P K 1 b d n

CHAPTER 02 2.4-43 REV. 14, SEPTEMBER 2008

LGS UFSAR where:

Ui = average velocity of ion in groundwater (ft/yr)

Uw = average velocity of groundwater (ft/yr)

Kd = distribution coefficient (ion exchange factor)

= KSr Ca Ex_

CCa KSr Ca = equilibrium constant (ratio)

Ex = cation exchange capacity (mg/gm)

CCa = calcium (or Ca + Mg) ion concentration in ambient groundwater (meq/ml)

Pb = bulk density of the bedrock (g/ml) n = porosity (ratio)

Cation exchange capacities were measured for three bedrock samples from the site. The samples were ground into two different size fractions, and cation exchange capacities were determined for the two size fractions according to the Agriculture Handbook No. 60 Method (Reference 2.4-39). Cation exchange capacities for samples including particle sizes ranging from 0.001 mm to 1.39 mm averaged 0.5 meq/g. Samples of particle sizes ranging from 4.01 mm to 38 mm have an average cation exchange capacity of 0.05 meq/g.

A cation exchange capacity of 0.05 meq/g was selected as representative of the materials through which the groundwater moves. The cation exchange capacity of a ground rock sample is inversely related to the particle size. Although no fine particle sizes are included in the 4 mm to 38 mm size fraction, bedrock fractures at the site are partially filled with particle sizes less than 4 mm. Thus, it is probable that exchange capacities are higher than 0.05 meq/g at the site, making the value conservative.

A bulk density of 2.65 g/ml, obtained from the literature (Reference 2.4-40), was selected as representative of the rock at the site.

The equilibrium constant ( KSr Ca ) for strontium-calcium systems reported in the literature, ranges between 1.0 and 1.6, depending on the percentage of Sr and Ca of the total cation concentration in solution and the principal clay minerals (Reference 2.4-41). Because reduction in concentration is inversely proportional to the selectivity coefficient, a conservative value of 1.0 was chosen.

Concentrations of dissolved solids in ambient groundwater are given in Table 2.4-19. The highest concentration of Ca in the groundwater samples tested is 178 ppm as CaC03. This sample has a total hardness of 618 ppm as CaC03. The concentration of Mg as CaCo3 is equal to the total hardness minus Ca hardness as CaC03, or 440 ppm. These concentrations are equivalent, as divalent ions, to 0.004 meq/ml Ca++ and 0.009 meq/ml Mg++, which is a total of 0.013 meq/ml to compete with Sr-90 for exchange sites.

CHAPTER 02 2.4-44 REV. 14, SEPTEMBER 2008

LGS UFSAR Using these values for the parameters, and the above described relationships, a distribution coefficient (Kd) of 3.84 ml/gm is determined. According to the calculated velocity of the Sr-90 in groundwater, the ion travels at 1.2 ft/yr and takes 666.6 yrs to travel the 800 feet to the river.

2.4.13.3.4 Dispersivity (Dispersion Constant)

The dispersivity (Dm) can be expressed as a characteristic length. Attempts by researchers have been made to relate the median grain-size (d50) to dispersivity in porous media (Reference 2.4-42).

The jointed and fractured bedrock beneath the site approximates a porous medium over a large representative elementary volume. Thus, the distance between fractures, which in some places is on the order of meters, is a reasonable estimate of dispersivity. However, for conservatism in this analysis, a dispersivity of 4 mm (0.013 ft) was chosen, the grain-size fraction from which the cation exchange capacity was determined.

2.4.13.3.5 Results of Analysis All of the parameters used for the analysis were selected with a very high degree of conservatism.

The parameters are presented in Table 2.4-20. The solution, which accounts for ion adsorption, radioactive decay, and one-dimensional longitudinal dispersion, was obtained for the Sr-90 ion. As discussed previously, the movement of strontium in groundwater is faster than the movement of cesium. Thus, actual cesium concentrations at the river would be much less than those calculated for Sr-90. Other radionuclides in the radwaste tank have short half-lives in comparison to groundwater travel times, and they will decay below MPC levels before reaching the Schuylkill River.

As discussed at the beginning of this section, accidental releases are assumed to spill from the waste sludge tank at the radwaste enclosure directly into the groundwater and flow toward the Schuylkill River. Radionuclides in the contaminated groundwater are considered to be at the same concentration as in the waste sludge tank (Table 11.4-9). The travel time of groundwater from the radwaste enclosure to the river is 3.28 years. Considering the adsorption of the Sr-90 ion, the travel time is 666.6 years. The maximum concentration of Sr-90 and Cs-137 that will arrive at the river is less than 1.34x10-10 Ci/ml. This is 103 times below MPC concentrations for both Sr-90 and Cs-137.

Because of the conservative nature of the input parameters, actual concentrations would probably be lower.

From the analysis, it is evident that should a spill of radioactive effluents from the radwaste enclosure reach the groundwater beneath the LGS site, it would not constitute a health hazard to the potable sources of water in the vicinity. Radionuclides in such a spill would decay sufficiently below MPC concentrations before leaving the site.

For the radioactive spill in the Unit 2 turbine building entering the groundwater, based on a conservative analysis, the only isotope above the 10CFR20 maximum permissible concentration (mpc) released from a feedwater line break is tritium (3H). The mpc value for tritium (3H) is 3 x 10-3 C/gm. NUREG-0016 indicates that the expected Tritium concentration in the feedwater is 1 x 10-2 uC/gm. Tritium would not be retarded by absorption into the groundwater and since it has a long half life, radioactive decay would not have an appreciable effect. The equilibrium concentration of tritium at well 4 will be 1.5 x 10-3C/gm (below the 10CFR20 mpc level) one year after the accidental spill.

Using the conservative assumption of a continuous source to model the feedwater line break, concentrations at site well 4 are below the 10CFR20 mpc levels. Ample time exists between the accident and the time that concentrations reach the maximum levels. Therefore, site well 4 can be used following an accident. Furthermore, routine monitoring tests of the wells utilized for drinking CHAPTER 02 2.4-45 REV. 14, SEPTEMBER 2008

LGS UFSAR water assures that the continued use of site wells does not compromise the safety of the public or plant employees.

2.4.13.4 Monitoring and Safeguard Requirements Groundwater levels were measured in borings (observation wells) at various locations (Table 2.4-17) near the spray pond and power block areas. The observation wells were monitored at regular intervals from June 1973 through August 1975, and November 1978 through May 1979 to determine the direction of groundwater movement and the elevation of the water table, and to establish structural design criteria for groundwater effects (Sections 2.4.13.2.4 and 2.4.13.5). These wells were subsequently monitored throughout the remaining plant construction period to determine the effect of plant construction on groundwater movement and level, and to verify/assure the adequacy of the design criteria used for design of safety-related plant structures. The recorded water levels in the wells are shown in the hydrographs in Figure 2.4-18.

Figure 2.4-20 shows the corresponding monthly precipitation, and Table 2.4-21 lists the monthly highest water levels measured at the wells of Figure 2.4-15 during the 36 months of monitoring. Also listed in the table are the estimated maximum water levels and highest levels of record at each well, and the corresponding precipitation during the month of highest level. During the 36 month observation period, the highest observed water level was 4.8 feet below the estimated maximum water level at well P11, and the average difference between the estimated maximum water level and the observed maximum water level is 12 feet.

Water levels continued to be measured monthly in the observation wells between September 1979 and February 1982. Three additional wells were constructed in 1981 (wells P17, P18 and P19) and are shown on Figure 2.4-15. Wells P1 and P4 were destroyed during construction of the spray pond.

Monitoring of the wells, up to February 1982, covered a period of 66 months, beginning in June 1973.

The extended monitoring of the observation wells demonstrated that May is an annual period of high water levels. During the period of monitoring the water levels did not exceed the estimated maximum water levels except in well SP22. This occurred in May, 1981 when the water level reached el 247.4' which is 0.1 foot higher than the estimated maximum. This well is located between the cooling towers and the spray pond. A review of the hydrograph of well SP22 (Figure 2.4-18) shows a steady increase in water level beginning in June 1974, when it was installed, to about April 1980. Since April 1980 the level has fluctuated in a manner similar to that of other wells. None of the other observation wells experienced this steady increase in level. They have tended to maintain a steady level, or to slightly decline since installation.

The steady increase in water level at SP22 is attributed to the buildup of a localized groundwater mound that was caused by infiltration of water from construction activities within this area. As construction activities decrease and the surface drains in the area are lined, this localized groundwater mound is expected to slowly dissipate.

This water level monitoring has indicated no changes in the groundwater flow direction. The changes in potentiometric surface throughout major plant construction and site grading/surfacing are within the plant design criteria established in Section 2.4.13.5.

The groundwater monitoring program will be discontinued one year after all major site grading/surfacing is completed because further significant changes in the potentiometric surface are not expected. The network of groundwater observation wells will then be abandoned and sealed.

2.4.13.5 Design Bases for Subsurface Hydrostatic Loading CHAPTER 02 2.4-46 REV. 14, SEPTEMBER 2008

LGS UFSAR During the operating life of the plant, the underlying groundwater table is expected to be at approximately the same elevations as shown in Figure 2.4-15. As discussed in Section 2.4.13.2.4.2, fluctuations of the measured levels in most observation wells during the 36 months of monitoring have been less than 12 feet, and the maximum has been 17 feet. The levels indicated on Figure 2.4-15 are near the seasonally high levels of those fluctuations.

Although the period of monitoring covers only 36 months, it is sufficient to establish typical seasonal fluctuations and provide a basis for estimating what may be expected during the operating life of the plant. Thus, the maximum expected groundwater levels are conservatively estimated to be 15 feet above the levels shown in Figure 2.4-15 (these levels were measured in May, when levels are near their maximum). In applying this to the design bases for subsurface hydrostatic loading, a single groundwater level is not applicable to all structures, nor even to individual structures, because of the relatively steep gradient of the water table. It can be seen in Figure 2.4-15 that, from the northeast corner of the turbine building to the southwest corner of the radwaste enclosure, the water table drops relatively uniformly more than 50 feet in elevation. Therefore, for the design bases, two maximum expected water levels are provided for each principal structure: the highest (beneath the northeast corner) and the lowest (beneath the southwest corner).

The maximum expected water levels are:

Maximum expected water table el (ft)

Structure Northeast corner Southwest corner Turbine enclosure 205 175 Reactor enclosure 195 175 Radwaste enclosure 180 160 Groundwater contours prepared from water levels measured on May 31, 1988 indicate that the above water levels are conservative. These contours are shown in Figure 2.4-15.

2.4.14 TECHNICAL SPECIFICATION AND EMERGENCY OPERATION REQUIREMENTS The possibility of adverse hydrologically related events at the LGS site is precluded by the configuration of the site topography.

Consequently, there are no emergency protective measures designed to minimize the water associated impact of adverse hydrologically related events on safety-related facilities. In addition, there is no need for technical specifications for plant shutdown required by accidents resulting from these events. Further discussion is in Sections 2.4.1.1 and 2.4.2.2.

The UHS, as described in Sections 2.4.8 and 9.2.6, has been designed with appropriate consideration to adverse hydrologically related events.

2.4.15 REFERENCES 2.4-1 Letter and enclosures from the Delaware River Basin Commission to R.F. Kilmartin, Bechtel, (March 29, 1977).

2.4-2 Letter from John T. Riedel, Chief, Hydrometeorological Branch, Office of Hydrology, NWS, NOAA, U.S. Department of Commerce, to R.F. Kilmartin, Bechtel, (March 18, 1976).

CHAPTER 02 2.4-47 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4-3 J. F. Riedel, J. F. Appleby, and R. W. Schlaemer, "Seasonal Variation of the Probable Maximum Precipitation East of the 105th Meridian for Areas from 10 to 1000 Square Miles and Durations of 6, 12, 24, and 48 Hours", Hydrometeorological Report No. 33, Weather Bureau, U.S. Department of Commerce, Washington, D.C., (1956).

2.4-4 U.S. Army Corps of Engineers, "Standard Project Flood Determination, EM-1110-2-1411", Office of the Chief of Engineers, Washington, D.C., (1975).

2.4-5 Carl E. Kindsvater, "Discharge Characteristics of Embankment-Shaped Weirs", Water Supply Paper 1617-A, USGS, Washington, D.C., (1964).

2.4-6 U.S. Bureau of Reclamation, "Design of Small Dams", Denver, Colorado, (1974).

2.4-7 USGS, "Surface Water Supply of United States, Part 1-B", Annual Water Supply Paper series through the 1960 water year.

2.4-8 USGS, "Surface Water Records of Pennsylvania", annual publications, water years 1961-1964.

2.4-9 USGS, "Water Resources Data for Pennsylvania, Part 1, Surface Water Records",

annual publications, water years 1965-1969.

2.4-10 USGS, "Water Supply Paper 1302, Compilation of Surface Water Records through September 1950, Part 1-B", (1960).

2.4-11 USGS, "Water Supply Paper 1722, Compilation of Surface Water Records, October 1950 to September 1960, Part 1-B", (1964).

2.4-12 E. H. Bourquard and Associates, "Report on Water Resources Survey of Main Stem of Schuylkill River, Pennsylvania," Bulletin No. 4, Department of Forests and Waters, Commonwealth of Pennsylvania, Harrisburg, Pennsylvania, (March 1968).

2.4-13 J. E. Biesecker, et al., "Water Resources of the Schuylkill River Basin", Bulletin No. 3, Department of Forests and Waters, Commonwealth of Pennsylvania, Harrisburg, Pennsylvania, (May 1968).

2.4-14 W. F. Busch, and L. C. Shaw, "Pennsylvania Streamflow Characteristics: Low Flow Frequency and Flow Duration", Bulletin No. 1, Department of Forests and Waters, Commonwealth of Pennsylvania, Harrisburg, Pennsylvania, (April 1966).

2.4-15 USGS, "Water Resources Data for Pennsylvania, Part 1, Surface Water Records",

(1972).

2.4-16 R. H. Tice, "Magnitude and Frequency of Floods in the United States", Water Supply Paper 1672, Part 1-B, USGS, (1968).

2.4-17 EPA, Philadelphia Office, "Communication Regarding Schuylkill Oil Spill",

Philadelphia, Pennsylvania, (January 23, 1973).

2.4-18 Department of Forests and Waters, Commonwealth of Pennsylvania, Schuylkill River Project, "Plate No. 40, River and Flood Profiles, Location of Permanent and Temporary Dams and Impounding Basins", Harrisburg, Pennsylvania, (December 28, 1950).

CHAPTER 02 2.4-48 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4-19 U.S. Army Corps of Engineers, "Backwater -- Any Cross-Section", Hydrologic Engineering Center, (June 1967).

2.4-20 U.S. Army Corps of Engineers, "Backwater Curves in River Channels, EM-1110-2-1409", (September 1960).

2.4-21 U.S. Army Corps of Engineers, "Emergency Employment of Army and Other Resources, NAPDR-500-1-1", Philadelphia District Office, Philadelphia, Pennsylvania, (March 1970).

2.4-22 U.S. Army Corps of Engineers, "Land Acquisition Procedure for Blue Marsh Dam and Reservoir", Philadelphia District Office, Philadelphia, Pennsylvania, (January 1970).

2.4-23 John G. Sakkas, "Dimensionless Graphs of Floods from Ruptured Dams", Report prepared for the Hydrologic Engineering Center, U.S. Army Corps of Engineers, (January 1974).

2.4-24 John G. Sakkas and Theodor Strelkoff, "Dimensionless Solution of Dam Break Flood Waves", Journal of the Hydraulics Division, ASCE, 102, HY2, pp. 171-184, (February 1976).

2.4-25 U.S. Army Coastal Engineering Research Center, "Shore Protection Manual",

Government Printing Office, Washington, D.C., (1973).

2.4-26 U.S. Army Corps of Engineers, North Atlantic Division, "Hydrologic Study Tropical Storm Agnes", New York, New York, (December, 1975).

2.4-27 Letter with tentative updated estimates of PMP for project area from John T. Riedel, Chief, Hydrometeorological Branch, Office of Hydrology, NWS, NOAA, U.S.

Department of Commerce, to R.F. Kilmartin, Bechtel, (March 18, 1976).

2.4-28 U.S. Army Corps of Engineers, "Standard Project Flood Determination EM-1110-2-1411, 27", (March 1952, revised March 1965).

2.4-29 U.S. Weather Bureau, "Rainfall Frequency Atlas of the United States, Technical Paper No. 40", (1961).

2.4-30 R.L. Weigel, "Oceanographical Engineering", Prentice- Hall, Englewood Cliffs, New Jersey, (1964).

2.4-31 Tennessee Valley Authority, "The Kentucky Project, Report No. 13", (1951).

2.4-32 Turgut Sarpkaya, "Added Mass of Lenses and Parallel Plates", Journal of the Engineering Mechanics Division, EM3, ASCE, (June, 1960).

2.4-33 V.T. Chow, "Open-Channel Hydraulics", McGraw-Hill, New York, (1959).

2.4-34 V.T. Chow, "Handbook of Applied Hydrology", McGraw-Hill, New York, (1964).

2.4-35 J.E. Biesecker, J.B. Lescinsky, and C.R. Wood, "Water Resources of the Schuylkill River Basin", Water Resources Bulletin No. 3, Department of Forests and Waters, Harrisburg, Pennsylvania, (1968).

CHAPTER 02 2.4-49 REV. 14, SEPTEMBER 2008

LGS UFSAR 2.4-36 S.M. Longwill, and C.R. Wood, "Groundwater Resources of the Brunswick Formation in Montgomery and Berks Counties, Pennsylvania", Pennsylvania Geological Survey, 4th series, Bulletin W-22, (1965).

2.4-37 P. Hammer, "Water Service Plan", Montgomery County Planning Commission, Court House, Norristown, Pennsylvania, (1976).

2.4-38 D.B. Grove, "Method to Describe the Flow of Radioactivity in Groundwater", Sandia Laboratories Report SC-CR-70-6139, (December 1970).

2.4-39 U.S. Department of Agriculture, "Diagnosis and Improvement of Saline and Alkaline Soils", Agricultural Handbook No. 60, (February 1954).

2.4-40 S.P. Clark, "Handbook of Physical Constants", Geologic Society of America, Inc.,

New York, New York, (1966).

2.4-41 W.J. Kaufman, "An Appraisal of the Distribution Coefficient for Estimating Underground Movement of Radioisotopes", Hazleton-Nuclear Science Corporation, Palo Alto, California, (March 1963).

2.4-42 J.J. Fried, and M.A. Cambernous, "Dispersion in Porous Media," Advances in Hydroscience, 7, pp. 169-282, (1971).

2.4-43 Bechtel Power Corporation, "Spray Pond Seepage Test Report for Limerick Generating Station, Unit 1 and 2", PECo, Philadelphia, Pennsylvania, (October 1983).

CHAPTER 02 2.4-50 REV. 14, SEPTEMBER 2008

LGS UFSAR Table 2.4-1 ACCESS OPENINGS IN SAFETY-RELATED STRUCTURES NUMBER FLOOR OF ELEVATION STRUCTURE ACCESS OPENING OPENINGS (ft)

Reactor Personnel door to service 1 per unit 201 enclosure water pipe tunnel Personnel door to outside 1 per unit 217 Personnel door to radwaste 1 per unit 217 enclosure or turbine enclosure Equipment airlock 1 per unit 217 Railroad car airlock 1 common 217 Personnel door to turbine 1 per unit 269 enclosure Control Double doors to turbine 3 200 structure enclosure Double doors to turbine 3 217 enclosure Personnel doors to turbine 9 239 enclosure Personnel door to turbine 1 254 enclosure Personnel doors to turbine 3 269 enclosure Personnel door to turbine 1 290 enclosure Personnel doors to turbine 2 304 enclosure Double doors to turbine 304 enclosure CHAPTER 02 2.4-51 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-1 (Cont'd)

NUMBER FLOOR OF ELEVATION STRUCTURE ACCESS OPENING OPENINGS (ft)

Diesel Personnel doors to outside 2 per DG 217 generator cell enclosure Spray Personnel doors to outside 2 268 pond pump structure Roll-up doors to outside 2 268 CHAPTER 02 2.4-52 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-2 MINOR DAMS UPSTREAM OF LGS SITE(1)

DRAINAGE HEIGHT VOLUME AREA NAME STREAM (ft) (acre-ft) (mi2)

Still Creek L. Schuylkill - Trib. 48 767 8.5 PA-422 L. Schuylkill - Trib. 87 3,850 15.6 PA-422A L. Schuylkill - Trib. 55 925 3.1 PA-424 L. Schuylkill - Trib. 35 459 2.2 PA-423 L. Schuylkill - Trib. 98 1,965 13.1 Tamaqua - 1 L.Schuylkill - Trib. 28 123 2.7 Tamaqua - 2 L. Schuylkill - Trib. 38 954 1.7 PA-425 L. Schuylkill - Trib. 21 229 1.1 Minersville - 1 Schuylkill - Trib. 24 55 4.4 Minersville - 2 Schuylkill - Trib. 33 196 2.6 Crystal Schuylkill - Trib. 40 200 5.1 Indian Run Schuylkill - Trib. 14 15 2.4 Silver Creek Schuylkill - Trib. 47 712 1.1 Auburn Schuylkill River 16 1,900 157.0 Dear Lake Schuylkill - Trib. 9 55 13.6 Kernsville Schuylkill River 17 583 340.0 PA-476 Schuylkill - Trib. 38 63 0.5 PA-477 Schuylkill - Trib. 47 206 1.59 PA-478 Schuylkill - Trib. 51 664 1.39 Felix Schuylkill River 24 1,470 647.0 Bernhart Schuylkill - Trib. 30 129 2.6 Antietam Schuylkill - Trib. 60 310 5.4 Green Hills Schuylkill - Trib. 17 187 15.3 (1)

These data were compiled from Reference 2.4-12 and from a tabulation furnished by the Dam Safety Section, Division of Dams and Encroachments, Department of Environmental Resources, Commonwealth of Pennsylvania, P.O. Box 2063, Harrisburg, PA 17120, dated March 1, 1977.

CHAPTER 02 2.4-53 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-3 DAM FAILURE INVESTIGATION:

CHARACTERISTICS OF MAJOR UPSTREAM DAMS MAIDEN BLUE CHARACTERISTICS ONTELAUNEE CREEK MARSH Date constructed 1926-34 Authorized(2) 1979 Distance upstream LGS (mi) 37 42 35 Stream name Maiden Maiden Creek Creek Tulpehocken el at base (ft) 253 302 236 el spillway crest (ft) 294 394 307 el top dam (ft) 305 412 332 PMF - peak inflow (cfs) No Est. 118,000 128,600 PMF - peak outflow (cfs) No Est. 92,000 74,800 SPF - peak inflow (cfs) No Est. 49,000 54,300 SPF - peak outflow (cfs) No Est. 17,500 30,300 SPF - max water surface (ft) No Est. 397.5 317.5 Volume below spillway (acre-ft) 11,900 114,000 50,000 Spillway capacity (cfs) 41,000 92,000 74,000 Type of dam Earth fill and Earth and Random fill with masonry rock fill impervious core gravity spillway Assumed water surface elevation 304.2 397.5 317.5 at time of failure(1) (ft)

Volume at time of failure (acre-ft) 29,000 130,000 76,700 (1) Used in failure study. Equivalent to water surface required to pass floods greater than or equal to the standard project flood (2) Construction indefinitely deferred CHAPTER 02 2.4-54 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-4 DOMESTIC WATER USERS ON SCHUYLKILL RIVER DOWNSTREAM OF LGS SITE ENTITLEMENT Consumptive as Approximate (2

1971 Non- a Percentage of Location No. Water User(1 Use Total Comsumptive Comsumptive Total (River mi.)(3)

1. Philadelphia Water Dept., 5,145.250 7,843.200 7,451.040 392.160 5 10.21 (Belmont)

Belmont Water Treatment Plant (4) (4) (4) (4) 4)

2. Philadelphia Water Dept., 12.61 (Queen Lane)

Queen Lane Water Treatment Plant

3. Keystone Water Co., 258.333 510.720 459.648 51.072 10 24.30 Norristown Dist.

4. Philadelphia Suburban 11.650 608.000 528.960 79.040 13 34.40 Water Co.

5. Phoenixville Borough 144.827 212.800 191.520 21.280 10 35.50

6. Citizens Utility Home 35.649 60.800 54.720 6.080 10 45.70 Water Co.

(1)

See Figure 2.4-2 for locations of water users.

(2)

Water use is given in 106 gallons per month (1 month = 30.4 days).

(3)

Measured from the confluence of the Schuylkill River with the Delaware River. Source: DRBC, Trenton N.J., and EPA Region III, Philadelphia, PA.

(4)

Data included with Belmont Water Treatment Plant.

CHAPTER 02 2.4-55 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-5 INDUSTRIAL WATER USERS ON SCHUYLKILL RIVER DOWNSTREAM OF LGS SITE(4)

ENTITLEMENT Consumptive as River Distance 1971(2 Non- a Percentage of From Station (1

No. Water User Use Total Comsumptive Comsumptive Total ( mi.)(3)

7. Connelly Containers Inc., 1.160 1.751 1.576 0.175 10 34.4 Philadelphia Plant

8. Container Corp. of America, 247.333 328.320 321.754 6.566 2 33.0 Philadelphia Plant, Mill Div.

9. Nicolet Industries Inc., 1.700 13.133 12.520 0.613 4.67 30.0 Norristown Plant

10. PECo, W. Conshohocken Gas 26.250 55.750 55.192 0.558 1 27.2 Plant

11. National Gypsum Co. (Allentown 66.666 66.666 65.999 0.667 1 27.05 Portland-Cement Co.)

W. Conshohocken Plant

12. Lukens Steel Co.(5) 54.602 52.107 2.495 4.6 25.70

13. PECo, Barbadoes Generating Station Closed in 1996 Station

14. Synthane-Taylor Corp. 24.417 39.398 39.004 0.394 1 17.4

15. Phoenix Steel Corp. 250.000 509.490 407.592 101.898 20 12.4 (Phoenixville Plant)

16. PECo, Cromby Generating Station Closed in 2011 Station

17. Keystone Coke Co.(5) 65.362 49.506 15.856 24.3 26.10 (1)

These serial numbers are a continuation of those in Table 2.4-4.

(2)

See Figure 2.4-2 for locations of water users.

(3)

Water use is given in 106 gallons per month (1 month = 30.4 days).

(4)

Source: DRBC, Trenton N.J., and EPA Region III, Philadelphia, PA.

(5)

No entitlement, rather a fee is paid for water withdrawn from the river.

CHAPTER 02 2.4-56 REV. 18, SEPTEMBER 2016

LGS UFSAR Table 2.4-6 MAJOR FLOODS AT SELECTED STATIONS ON SCHUYLKILL RIVER(1)

BERNE READING POTTSTOWN PHILADELPHIA DA = 335 sq mi DA = 880 sq mi DA = 1147 sq mi DA = 1893 sq mi Gauge Ht Peak Q Gauge Ht Peak Q Gauge Ht Peak Q Gauge Ht Peak Q YEAR DATE (ft) (cfs) (ft) (cfs) (ft) (cfs) (ft) (cfs) 1972(2) 22-23 June 19.00 42,800 - - 29.97 95,900 14.65 103,000 1955 19 Aug 15.73 29,400 - - 17.98 42,300 14.32 90,100 1950 26 Nov 14.52 23,300 - - 17.90 42,000 14.32 89,800 1942 23 May 15.00 26,900 21.6(3) - 20.15 50,800 12.44 61,400 1933 24 Aug - - - - 19.20 47,800 14.70 98,200 1902 28 Feb - - 21.5 70,600 21.00 53,900 14.80 98,000 1869 4 Oct - - 21.6 71,200 - - 17.00 135,000 1850 2 Sept - - 23.0 80,000 - - 16.42 125,000 1839 26 Jan - - 13.9 33,300 - - 15.80 114,000 1757 15 July - - 15.0 37,200 - -

(1)

From References 2.4-21 and 2.4-22 (2)

From Reference 2.4-15 (3)

Estimated CHAPTER 02 2.4-57 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-7 PROBABLE MAXIMUM PRECIPITATION AT LGS SITE INITIAL 6 HOUR PMP DISTRIBUTION(1)

Incremental Fraction of Time Depth Hour Total (hrs) (in) Depth (%)

0-1 2.7 10 1-2 3.2 12 2-3 4.0 15 3-4 10.2 38 4-5 3.8 14 5-6 2.9 11 72 HOUR PMP DISTRIBUTION(2)

Accumulated Time PMP Depth (hrs) (in) 6 26.8 12 31.1 24 34.4 48 38.0 72 39.7 (1) 6 hour6.944444e-5 days <br />0.00167 hours <br />9.920635e-6 weeks <br />2.283e-6 months <br /> PMP distribution taken from Reference 2.4-4 (2)

PMP estimates taken from Reference 2.4-2 CHAPTER 02 2.4-58 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-8

SUMMARY

OF RUN-OFF FROM LOCAL INTENSE PRECIPITATION RUN-OFF BY DRAINAGE AREA Drainage Area Peak Discharge Routed to Area(1) (acres) (cfs) Collection Point(1)

DA-1 17.2 251(2) CP-5 DA-2 25.0 620 CP-1 and CP-2 DA-3 4.5 151 CP-3 DA-4 20.8 550 CP-3 and CP-4 DA-5 14.9 369 CP-3 DA-6 11.6 242 CP-4 RUN-OFF BY COLLECTION POINTS Maximum Maximum Collection Areas Discharge Energy Grade-Point(1) Drained(1) at CP (cfs) Line at CP CP-1 DA-2 530 262.7 CP-2 DA-2 620 245.5 CP-3 DA-3, 4,and 5 446 218.1 CP-4 DA-4 and 6 555 218.1 CP-5 DA-1 194 252.0 (1)

See Figures 2.4-4, 2.4-5, and 2.4-6 for locations of drainage areas and collection points.

(2)

Corresponds to maximum 30 minute mean inflow to the spray pond.

CHAPTER 02 2.4-59 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-9 SURFACE WATER GAUGING STATIONS UPSTREAM FROM LGS SITE PERIODS OF RECORD Daily or Annual RAINAGE Monthly Peak Low Flow LOCATION STATION AREA (Calendar (Water Measurements NO. STATION NAME NO. (sq mi) Year) Years) (Water Years) 1 Schuylkill River at Pottsville, Pa. 4675 53.4 1943- - -

2 West Branch Schuylkill River at Cressona, Pa. 4679.5 52.5 1964-1965 - -

3 Schuylkill River at Landingville, Pa. 4685 133 1947-1953; 1963-1965 - -

4 Schuylkill River at Auburn, Pa. 4690 160 1947-1951 - -

5 Still Creek Reservoir near Hometown, Pa. 4692 8.5 1933- - -

6 Little Schuylkill River at Tamaqua, Pa. 4695 42.9 1916-1919; - -

1919-7 Little Schuylkill River at Drehersville, Pa. 4700 122 1947-1951; - -

1963-1965 8 Schuylkill River at Berne, Pa. 4705 355 1947- - -

9 Maiden Creek near Lenhartsville, Pa. 4707 75.5 - - 1943; 1946-57 10 Maiden Creek Tributary at Lenhartsville, Pa. 4707.2 7.46 1965- 1962-1965 -

11 Sacony Creek at 4707.5 55.5 - - 1958 -

Virginville, Pa.

12 Mill Creek near Bernville, Pa. 4707.8 11.9 - - -

CHAPTER 02 2.4-60 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-9 (Cont'd)

PERIODS OF RECORD Daily or Annual RAINAGE Monthly Peak Low Flow LOCATION STATION AREA (Calendar (Water Measurements NO. STATION NAME NO. (sq mi) Year) Years) (Water Years) 13 Tulpehocken Creek at Bernville, Pa. 4708 84.8 - - 1943; 1946-57 14 Northmill Creek at Bernville, Pa. 4709 42.0 - - 1943; 1946-57 15 Tulpehocken Creek at Blue Marsh Dam site near Reading, Pa. 4709.6 175 1965- - -

16 Tulpehocken Creek near Reading, Pa. 4710 211 1950- - -

17 Schuylkill River at Reading, Pa. 4715 880 1914-1915; - -

1915-1919; 1919-1930 18 Allegheny Creek at Beckersville, Pa. 4716 11.3 - - 1946-1957 19 Monocacy Creek at Limekiln, Pa. 4717 6.68 - - 1946-1957 20 Pine Creek near Manatawny, Pa. 4718 15.6 - 1961- -

21 Manatawny Creek at Earlville, Pa. 4719 60.0 - - 1946-1957 22 Schuylkill River at Pottstown, Pa. 4720 1147 1926- - -

(1)

The locations are shown on Figure 2.4-8.

CHAPTER 02 2.4-61 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-10 OBSERVED AND ESTIMATED WATER SURFACE ELEVATIONS OF SCHUYLKILL RIVER AT LGS SITE WATER SURFACE ELEVATION DISCHARGE AT SITE OCCURRENCE (cfs) (ft)

Observed 12/5/69 566 105.3 Observed 12/18/69 1,610 106.5 Computed Floods Average Annual (Pottstown Records) 21,000 117.4 Average Annual (WSP 1672) 28,000 119.5 Maximum Observed Before 1972 53,900 125.7 100 Year (WSP 1672) 99,000 134.8 Additional Discharge for Rating Curve 200,000 145.0 Additional Discharge for Rating Curve 356,000 158.0 CHAPTER 02 2.4-62 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-11 DERIVED VALUES OF MANNING'S (n) FOR SCHUYLKILL RIVER MAIN CHANNEL WATER LEVEL APPROXIMATE MANNING'S REACH LOCATION Observed Computed (n) LENGTHS (mi)

Pottstown Hanover 135.84 135.8 St. Bridge 0.028 1 S. Pottstown 133.55 133.6 Madison St. Bridge 0.036 6 Linfield Highway 120.03 120.0 Bridge 0.048 4.7 Spring City Highway 105.97 106.0(1) - -

Bridge (1)

Assigned starting elevation for computations CHAPTER 02 2.4-63 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-12 COMPUTATION OF CONVERGENCE IN BACKWATER STUDY STARTING WATER COMPUTED WATER SURFACE SURFACE AT SITE DISCHARGE (cfs) -121/2% +121/2%(1) -121/2% +121/2% DIFF (ft) 21,000 95.8 97.6 117.4 117.4 -

28,000 97.8 100.0 119.5 119.5 -

53,900 103.9 106.8 125.7 125.7 -

99,000 111.4 115.3 134.7 134.9 0.2 200,000 123.7 128.9 144.8 145.7 0.9 356,000 137.8 144.5 157.3 159.2 1.9 AR 23 (1)

+121/2% refers to assumed channel conveyance n

compared with the computed requirement at the downstream end of study reach.

CHAPTER 02 2.4-64 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-13 CONVEYANCE, SLOPE, AND STAGE FOR SELECTED DISCHARGES ABOVE 356,000 cfs AT LGS SITE REQUIRED REQUIRED DISCHARGE FRICTION CONVEYANCE STAGE Q SLOPE K el=f(K)

(cfs) Sf (106 cfs) (ft) 450,000 0.00025 28.2 171 600,000 0.00023 39.9 183 750,000 0.00021 51.3 197 900,000 0.00020 63.6 207 1,100,000 0.00019 80.4 217 CHAPTER 02 2.4-65 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-14 CHEMICAL ANALYSIS OF GROUNDWATER IN THE BRUNSWICK LITHOFACIES IN MONTGOMERY COUNTY, PENNSYLVANIA(1)

DEPTH OF TOTAL DATE OF WELL SILICA TOTAL IRON MANGANESE CALCIUM MAGNESIUM SODIUM POTASSIUM BICARBONATE COLLECTION (ft) (SiO) (Fe) (Mn) (Ca) (Mg) (Na) (K) (HCO3) 9-25-25 350 18 0.06 - 47 17 9.4 2.1 194 9-28-25 388 32 0.05 - 36 15 11 1.8 173 2-21-52 387 21 0.01 - 24 20 6 1.0 150 5-10-62 219 30 0.37 0.00 57 18 13 0.7 171 4-07-53 205 13 0.26 - 57 28 13 1.0 242 2-07-62 373 17 0.07 0.00 45 5.4 12 1.0 134 2-08-62 202 22 0.04 0.00 55 23 14 1.0 256 4-09-62 600 20 0.44 0.02 116 51 22 0.8 163 3-02-61 300 22 0.05 0.00 39 8.3 10 1.0 120 3-01-61 450 19 0.26 0.01 47 9.0 14 1.0 179 4-09-62 210 24 0.00 0.17 49 12 12 0.8 128 3-02-61 916 28 3.9 0.04 180 32 27 1.0 180 4-21-49 100 20 0.17 - 52 13 11 1.4 198 2-28-61 500 16 0.38 0.03 30 8.2 45 0.5 173 3-01-62 312 28 0.02 0.03 90 36 19 1.8 162 2-27-61 300 17 0.21 0.06 71 30 30 1.5 217 3-01-61 300 24 0.06 0.04 39 14 8.3 1.0 178 3-01-61 500 28 0.70 0.38 59 17 15 1.0 252 3-01-61 300 32 0.02 0.05 126 28 19 2.5 158 2-05-62 123 23 0.14 0.00 70 21 12 1.5 236 2-08-62 80 23 0.05 0.00 44 34 12 1.0 298 2-08-62 100 22 0.20 0.00 55 20 14 2.2 202 2-08-62 81 33 0.07 0.00 36 17 11 3.0 152 2-05-62 157 21 1.6 0.00 54 36 17 1.5 268 4-09-62 133 19 3.0 0.12 53 15 15 1.5 211 9-30-25 111 23 4.9 - 45 11 12 3.0 183 9-30-25 110 25 0.15 - 48 24 12 1.2 234 CHAPTER 02 2.4-66 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-14 (Cont'd)

DISSOLVED HARDNESS AS SPECIFIC SOLIDS CaCO3 CONDUCTANCE DATE OF SULFATE CHLORIDE FLUORIDE NITRATE (residue CALCIUM, NON- (microhms/cm COLLECTION (SO4) (Cl) (F) (NO3) at 180oC) MAGNESIUM CARBONATE at 25oC) pH 9-25-25 23 13 - 7.5 232 187 28 - -

9-28-25 15 8 - 2.5 201 152 10 - -

2-21-52 22 5 - 0.4 - 142 19 321 6.4 5-10-62 69 16 0.1 4.9 317 216 76 457 6.6 4-07-53 58 18 0.0 9.9 327 257 59 555 7.3 2-07-62 26 10 0.1 13.0 200 135 25 313 7.5 2-08-62 19 17 0.1 7.7 285 232 22 480 7.6 4-09-62 370 11 0.1 11 732 500 366 959 7.3 3-02-61 24 7.4 0.1 20 192 132 33 295 7.7 3-01-61 12 9.3 0.0 18 214 155 8 351 7.8 4-09-62 69 5.8 0.1 13 252 172 67 378 6.8 3-02-61 420 18 0.2 2.8 805 581 433 1090 7.4 4-21-49 23 7.0 0.0 12 242 183 21 392 7.5 2-28-61 48 3.5 0.0 3.7 239 109 0 378 8.0 3-01-62 248 6.0 0.2 2.7 534 373 240 747 7.7 2-27-61 84 68 0.1 5.6 426 301 123 695 7.9 3-01-61 17 4.2 0.0 8.0 204 155 9 322 7.7 3-01-61 37 4.2 0.1 0.2 283 217 11 447 7.5 3-01-61 300 16 0.6 0.5 620 430 300 838 7.6 2-05-62 32 20 0.1 36 344 261 68 536 7.6 2-08-62 9.7 5.1 0.1 20 302 250 6 490 8.2 2-08-62 40 18 0.2 19 307 219 54 478 7.4 2-08-62 29 13 0.2 8.8 236 160 36 359 7.7 2-05-62 83 6.4 0.1 3.1 360 283 63 565 7.6 4-09-62 27 13 0.0 4.0 255 194 21 429 6.9 9-30-25 13 12 - 0.88 209 158 8 - -

9-30-25 6.4 19 - 14 252 218 26 - -

(1)

Results are in parts per million except as indicated.

CHAPTER 02 2.4-67 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-15 PUBLIC GROUNDWATER SUPPLIES SOURCE PRIMARY (1), SECONDARY (2)

AV PLANT ESTIMATED NUMBER OUTPUT POPULATION OF WATER COMPANY AREA SERVED STREAMS WELLS SPRINGS SOURCES (mgd) SERVED SERVICES Ambler Borough Ambler Borough; - 1 2 1 spring 1.82 15,310 4,410 Water Department (B) part of Upper 6 wells Dublin, Whitpain Lower Gwynedd, and Whitemarsh Townships Audubon Water Lower Providence - 1 - 10 wells .39 5,600 1,156 Company Township (B)

Borough of Borough of 2 - 1 Several .46 4,400 1,423 Boyertown Water Boyertown; springs, Department (A) part of Douglass Trout River, and Colebrookdale unnamed Townships tributary to Ironstone Creek, Ironstone Creek Citizens Utilities Royersford and - 2 - 1.46 7,830 3,059 Home Water Spring City Schuylkill River Company (B) Boroughs, part 2 wells of Limerick and Upper Providence Townships Collegeville-Trappe Collegeville and - 1 - 6 wells .26 4,870 977 Joint Water Works (B) Trappe Boroughs Douglasville Amity Township - 1 - 2 wells .02 450 100 Water Company (A)

Dublin Water Upper Dublin Company (B) Township - 1 - 3 wells .60 3,650 1,170 Borough of East East Greenville 1 2 - Perkiomen Greenville Water Borough Creek Department (B) 1 well .19 2,100 1,170 CHAPTER 02 2.4-68 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-15 (Cont'd)

SOURCE PRIMARY (1), SECONDARY (2)

AV PLANT ESTIMATED WATER OUTPUT POPULATION NUMBER OF COMPANY AREA SERVED STREAMS WELLS SPRINGS SOURCES (mgd) SERVED SERVICES Evansburg Water Evansburg Borough; - 1 - 1 well .02 120 38 Company (B) part of Lower Providence Township Graterford State Graterford - 1 - 8 wells .67 1,700 --

Correctional Institution (B)

Hatfield Water Hatfield Borough; - 1 - 7 wells .50 6,790 1,142 Department (B) part of Hatfield Township Horsham Water Horsham Township - 1 - 11 wells .74 7,850 2,497 Authority (B)

Lionville Water Co Uwchland Township (A) - 1 - 2 wells .01 150 45 Lyons Water Co Lyons Borough 2 wells, 2 (A) - 1 2 springs .03 590 143 Malvern (Water Malvern Borough - 1 2 .15 725 180 Dept.) 3 wells, (A) 2 springs Mount Penn Mount Penn 3 1 2 3 wells, .60 12,000 2,797 Municipal Borough, Lower springs, Water Authority Alsace and St. stream (A) Lawrence Townships Norristown State Norristown - 1 - 6 wells .36 1,600 --

Hospital (B)

North Penn Water Lansdale and - 1 - 36 wells 3.40 38,000 9,495 Authority (B) Souderton Bulk purchase Boroughs, from North Franconia,Hatfield Wales and Towamencin Townships; part of Worcester Township CHAPTER 02 2.4-69 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-15 (Cont'd)

SOURCE PRIMARY (1) SECONDARY (2)

AV PLANT ESTIMATED OUTPUT POPULATION NUMBER OF WATER COMPANY AREA SERVED STREAMS WELLS SPRINGS SOURCES (mgd) SERVED SERVICES North Wales Water North Wales Borough; - 1 - 19 wells 2.97 22,710 6,192 Authority (B) Upper & Lower Gwynedd Townships and Montgomery Township; bulk water to Lansdale Borough Pennhurst State East Vincent - 1 - 6 wells .40 3,900 --

School (A)

Township Pennsburg Water Co (A) Part of Upper - 1 2 2 wells .35 2,400 700 Hanover and Hereford Townships Perkasie Borough Perkasie Borough - 1 2 4 wells, .25 5,500 1,350 Authority (A) springs Philadelphia Suburban Conshohocken, 1 2 - Schuylkill River, 77.6 311,170 219,796 Water Company (B) Narberth and , Pickering Creek West Conshohocken Perkiomen Creek, Boroughs; 17 wells Abington, Cheltenham, Lower Merion, Plymouth, Spring Field, Upper Dublin, Upper Merion, Upper Moreland, Lower Moreland, Whitemarsh, Eastown, East Whiteland, Tredyffrin, Chester, Willistown and Radnor Townships CHAPTER 02 2.4-70 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-15 (Cont'd)

SOURCE PRIMARY (1), SECONDARY (2)

AV PLANT ESTIMATED OUTPUT POPULATION NUMBER OF WATER COMPANY AREA SERVED STREAMS WELLS SPRINGS SOURCES (mgd) SERVED SERVICES Red Hill Water Red Hill - - 1 spring .12 1,550 496 Authority (B) Borough Saint Gabriel's Lower Providence - 1 - 2 wells .02 -- --

Hall (B) Township Schwenksville Schwenksville - 1 - 3 wells .13 1,985 336 Borough Borough; part of Water Department Lower Frederick (B) and Perkiomen Townships Sellersville Sellersville - 2 1 Springs, .20 2,500 650 Water Department 4 wells (B)

Skippack Water Skippack - 1 - 8 wells .002 225 72 Company Township Telford Water Telford and part - 1 - 4 wells .02 - -

Authority (B) of Souderton Boroughs, part of Franconia Township Trumbauersville Trumbauersville - 1 - 2 wells .03 400 94 Water Department (A)

(A)Biesecker et al., 1968 (B)Hammer, 1976 CHAPTER 02 2.4-71 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-16 (1)

PRIVATE GROUNDWATER USERS IN THE VICINITY OF THE SITE OWNER'S NAME STATIC TOTAL CASING WATER DEPTH DIAM LENGTH LEVEL YIELD (ft) (in) (ft) (ft) (gpm) REMARKS Leroy Shaner 226 6 40+/- - - Pump Capacity 30 gpm Serves 104 trailer houses and apartments Leroy Shaner 300 6 40+/- 50+/- - Pump Capacity 18 gpm Tom & Mike Volpe 193 6 - - - Serves bar and grill Saratoga Inn 120 6 - - - Serves hotel bar and restaurant Mike Bar 138 6 - - - Single dwelling Pottstown Trap Rock 130 6 20+/- 40+/- 100 Pumped at 50 gpm 10 hr/day Quarry James Florig 260 6 20 80+/- -

Cliff Groff 20 - - - - Dug well George Kirlin 110 6 45 40+/- - Single dwelling and livestock Leonard Miller 109 6 20 50 6 Single dwelling William Soditus 87 - - - - Single dwelling Parker Ford Esso 38 - - - - Gasoline service station Sailors Groceries 100+/- - - - - Grocery store and single dwelling Eastern Warehouses, Inc 80 4 - - 40

" " " 140 6 60+/- - 120 Business (Warehouse and trucking)

" " " 100+/- 4 - - -

" " " 100+/- 4 - - - -

Wilmer Godshall 160 6 60 100+/- 60 Single dwelling Ray Miller 90 - - 15+/- - Single dwelling East Coventry Elem 202 6 69 72 25 500 elementary students and staff-School two wells E. Pennypacker - - - - - Spring in basement, supplied 30 head dairy cattle and dwelling.

Never dry in 50 years.

R. Elliot Supplies 80 head dairy cattle and 90 6 30+/- 5+/- - dwelling Swanson Service Co Shallow - - 13+/- - 10 employees N. J. Hedrick 125 6 - 75+/- - Single dwelling Publicker Industries 200 6 100+/- 50+/- 300 150,000 gpd average use from

" " three wells in warehousing and 160 6 100+/- 50+/- 200 bottling of distilled spirits

" " 160 6 100+/- 50+/- 120

" " - - - - - Four wells similar to above in same

" " - - - - - area are inactive or abandoned.

(1)

Inventory of wells performed by Dames and Moore, 1970 CHAPTER 02 2.4-72 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-17 OBSERVATION WELL CONSTRUCTION DATA ELEVATION (ft)

INTERVAL SCREENED BELOW OBS. MONTH Ground Reference DEPTH OF HOLE GROUND SURFACE WELL CONSTRUCTED Surface Point (ft) (ft)

P1 6/73 268.31 268.59 120.0 105.0 - 115 P2 6/73 264.93 266.03 63.0 47 - 57 P3 6/73 244.65 245.38 65.0 50 - 60 P4 6/73 256.45 257.03 75.0 60 - 70 P5 6/73 249.03 247.60 71.0 56 - 66 P6 1/79 275.6 278.0 70.0 55 - 65 P7 1/79 267.8 269.6 72.5 57.5 - 67.5 P8 1/79 253.9 255.9 65.0 50 - 60 P9 1/79 257.6 260.2 65.0 50 - 60 P11 1/79 215.7 218.8 60.0 45 - 55 P12 1/79 215.5 218.3 100.0 65 - 75 85 - 95 P13 1/79 216.8 219.7 100.0 65 - 75 85 - 95 P14 1/79 216.4 217.3 80.0 65 - 75 P15 1/79 216.1 218.5 90.0 75 - 85 P16 1/79 216.7 219.2 80.0 65 - 75 P17 6/81 262.56 262.52 115.0 98 - 108 P18 6/81 266.0 267.8 120.0 100 - 110 P19 6/81 253.22 255.64 99.0 61 - 71 SP20 6/74 244.52 245.24 65.0 50 - 60 SP21 6/74 244.63 247.90 60.0 45 - 55 SP22 6/74 263.85 265.73 62.2 47.2 - 57.2 SP23 6/74 274.67 276.03 63.5 48.5 - 58.5 CHAPTER 02 2.4-73 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-18 PERMEABILITY DATA Depth Interval Permeability Hole No. Type of Test Tested (ft.) Zone Tested (ft/yr)

PT1 Well Permeameter 4.38 - 7.55 Soil <1 PT2 " 7.75 - 12.17 Soil 3 PT3 " 16.67 - 22.67 Soil 3 PT4 " 12.00 - 16.16 Soil 9 PT5 " 21.52 - 26.94 Rock <1 PT6 " 19.37 - 23.25 Rock <1 PT7 " 6.37 - 12.67 Soil <1 SP17 " 20.97 - 21.77 Contact 21 SP18 " 19.10 - 21.10 Contact 17 SP19 " 14.70 - 16.45 Contact 3 P1 Packer 10 - 20 Rock 38

" 20 - 30 Rock 41

" 30 - 40 Rock 176

" 40 - 50 Rock 266

" 50 - 60 Rock 18

" 50 - 60 Rock 7

" 60 - 70 Rock 71

" 70 - 80 Rock 1

" 80 - 90 Rock 287

" 90 - 100 Rock <1

" 100 - 110 Rock 28

" 110 - 120 Rock 5 P2 " 14 - 22 Rock <1

" 22 - 32 Rock 41

" 32 - 42 Rock 86

" 42 - 52 Rock 119

" 52 - 62 Rock 50 P3 " 15 - 25 Rock 1081

" 25 - 35 Rock 476

" 35 - 45 Rock 459

" 45 - 55 Rock 371

" 55 - 65 Rock 191 P4 " 25 - 35 Rock 329

" 35 - 45 Rock 58

" 45 - 55 Rock 525

" 55 - 65 Rock 127

" 65 - 75 Rock 361 P5 " 21 - 31 Rock 1247

" 41 - 51 Rock 66

" 51 - 61 Rock 314

" 61 - 71 Rock 200 CHAPTER 02 2.4-74 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-19 CHEMICAL ANALYSIS OF GROUNDWATER FROM WELLS IN THE BRUNSWICK LITHOFACIES AT THE LGS SITE (Results in ppm except as indicated)

TOTAL DATE DEPTH OF CALCUIM NITRATE AMMONIA IRON TOTAL HARDNESS DISSOLVED Well No. COLLECTED WELL (ft) AS CaCO3 CHLORIDE SULFATE AS (N) AS (N) (TOTAL) ALKALINITY EDTSA SOLIDS pH 1 12/10/70 307 16.0 4.5 422 0.26 1.36 .0 120 540 988 7.7 3 12/10/70 508 177.9 6.0 461 0.38 2.24 .0 148 618 1052 7.9 4 08/12/71 198 32.0 3.0 77 1.0 ---- .04 116 134 202 8.0 CHAPTER 02 2.4-75 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-20 ACCIDENTAL SPILL ANALYSIS PARAMETERS PARAMETERS SYMBOL VALUE Hydraulic Conductivity k 390 ft/yr Hydraulic Gradient I 0.031 Effective Porosity n 0.05 Average Groundwater Velocity U 243 ft/yr Distance x 800 ft Travel Time of Groundwater t 3.28 yr Volume of spill 10,240 gal Width of spill 30 ft Initial Concentration Sr-90 Co 2.2x10-3 Ci/ml Ion Exchange Capacity Ex 0.05 mea/g Concentration of Divalent Ions Ca 0.013 mea/ml Equilibrium Constant K 1 Median Grain-Size d50 4 mm (0.013 ft)

Dispersivity (Dispersion constant) D 0.013 ft Distribution Coefficient Kd 3.84 ml/g Bulk Density Pb 2.65 gm/ml CHAPTER 02 2.4-76 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-21

SUMMARY

OF GROUNDWATER LEVEL MEASUREMENTS USED TO ESTABLISH DESIGN BASES FOR HYDROSTATIC LOADING Water Level Elevation (ft)

Precipitation during month Estimated Highest level Month of of highest level (c)

Observation maximum of Highest level of highest level of of 36 months (a) (b (c) (c)

Well level record 36 months 36 months (inches)

(d)

P1 239.2 228.4 226.4 3/79 2.91 (e)

P2 - 229.2 229.2 6/75 7.91 (i)

P3 242.4 229.6 229.6 2/75 3.96 (d)

P4 243.3 230.1 230.1 2/75 3.96 (i)

P5 239.1 229.6 229.6 2/75 3.96 (i)

P6 255.2 241.4 240.2 5/79 6.98 (i)

P7 237.3 222.3 222.3 5/79 6.98 (i)

P8 245.9 232.3 230.9 5/79 6.98 (h)

P9 257.2 242.8 242.5 3/79 2.91 (i)

P11 206.5 203.3 201.7 3/79 2.91 (i)

P12 138.3 127.6 127.6 2/79 6.27 (f)

P13 177.0 167.2 162.3 1/79 13.18 (i)

P14 193.0 178.0 178.0 5/79 6.98 (g)

P15 151.0 138.8 138.8 1/79 13.18 (i)

P16 169.2 161.1 161.8 3/79 2.91 (i)

P17 --- 283.3 --- --- ---

(i)

P18 --- 226.5 --- --- ---

(i)

P19 --- 224.5 --- --- ---

(e)

SP20 --- 230.5 230.5 2/75 3.96 (e)

SP21 --- 222.4 222.4 7/74 1.93 (i)

SP22 247.3 247.4 240.6 1/79 13.18 (i)

SP23 266.9 255.0 255.0 10/74 2.51 (a)

Determined by adding 15 feet to May 1979 measured level.

(b)

Measurements as of February 1982.

(c)

June 1973 to August 1975, November 1978, January 1979 to August 1979.

(d)

Well destroyed.

(e)

Destroyed prior to May 1979.

(f)

Abandoned in June 1984.

(g)

Abandoned in December 1987.

(h)

Abandoned in April 1988.

(i)

Monitoring to be discontinued one year after completion of all major site grading/surfacing.

CHAPTER 02 2.4-77 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-22 PARAMETERS USED IN RATIONAL FORMULA AND KIRPICH'S FORMULA Time of Concentration Area Slope Length Tc Rainfall Intensity Peak Flow Rate SUBAREA A S L Calculated Adopted I Q (acres) (ft) (min) (min) (in/hr) (cfs)

DA-1 17.2 * * * *

• 251 DA-2 25.0 0.00133 600 13.8 10 24.78 620 DA-3 4.5 0.0398 830 4.8 5 32.16 151 DA-4 20.8 0.019 1200 8.4 8.4 26.37 550 DA-5 14.9 0.0033 900 13.3 10 24.78 369 DA-6 11.6 0.00222 900 15.4 15 20.88 242 CP-1 20.0 0.00133 320 8.4 8.4 26.37 530 CP-2 25.0 0.00133 600 13.8 10 24.78 620 CP-3 12.8 0.0033 900 13.3 10 24.78 446 CP-4 35.6 0.0024 1250 19.2 20.0 18.27 555 CP-5 17.2 ** ** ** ** ** 194 Rational Formula Q = CIA Kirpich's Formula Tc = 0.0078 (L/S) ^ 0.77

• Rational method was not used for subarea DA-1.

• • Rational method was not used for collection point CP-5, flow is maximum 30 minute mean inflow to spray pond.

CHAPTER 02 2.4-78 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.4-23 DRAINAGE FLOW PATH CROSS-SECTIONAL DATA CROSS- OUTLET AVERAGE BOTTOM SECTION NUMBER WIDTH (ft) ELEVATION (ft)

A 7 37 215.00 B 40 215.40 C 6 17.5 215.00 E 60 216.40 F 5 29 215.00 G 22.75 215.00 H 50 216.65 I 62 215.80 II 4 41 215.00 K 3 52 215.00 L 28 215.80 M 2 15 214.90 N 13 216.40 O 1 12 216.20 P 11 216.40 Q 72 216.40 R 104 216.10 S 150 215.20 TT 200 216.20 V 45 215.50 Z 8 166 216.60 CHAPTER 02 2.4-79 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5 GEOLOGY AND SEISMOLOGY In accordance with the criteria provided in 10CFR100 Appendix A, "Seismic and Geologic Siting Criteria for Nuclear Power Plants", and Regulatory Guide 1.70 (Rev 3), this section describes and evaluates the geologic and seismologic conditions for the region around LGS. Foundation conditions are evaluated, and the foundation design is described. The seismic history of the region is examined, and the earthquake design criteria are described.

This section presents the results of the evaluation of the regional and site geology and demonstrates that the evaluation is in sufficient detail to ensure the safe design and operation of the nuclear power facility. Results of the literature study, field studies, foundation exploration, and laboratory test programs are presented. Static and dynamic properties of the foundation materials are described, and design criteria are outlined. Groundwater factors affecting plant construction and operation are discussed. (Groundwater conditions are presented in detail in Section 2.4.13).

Investigations were performed to define the site foundation conditions, and regional and site geologic, geohydrologic, and seismological conditions. These investigations determined the characteristics of the foundation materials and their suitability for supporting the structures, the depth and configuration of the groundwater table, and the characteristic effects of the soil and rock materials on the migration of radioactive solutions, should such solutions come in contact with them. Investigations were also performed to evaluate tectonics, faulting, and seismicity of the area, so that appropriate parameters for seismic design could be selected. Results of fault investigations are discussed in Section 2.5.3 and are presented in detail in a report prepared by Dames and Moore in 1974 (Reference 2.5-1). Site exploration is discussed in detail in Section 2.5.4.

LGS is approximately 3 miles southeast of Pottstown, Pennsylvania, adjacent to the Schuylkill River. It is in the Triassic Lowland section of the Piedmont physiographic province. The area is within the Newark-Gettysburg Basin, which is underlain by red sandstones, shales and siltstones of the Triassic Newark Group. These sedimentary basin deposits are gently tilted and warped, and are cut by diabase dikes and sills and by minor faulting.

Faulting has occurred in association with the major episodes of deformation in the region, and some faulting of the Triassic sediments occurred during Triassic and Jurassic time, over 140 million years ago. Some minor Jura-Triassic faults occur near the site; detailed studies carried out by the licensee show that they are not significant to the construction and operation of the plant.

Earthquake activity in historic time within 200 miles of the site has been moderate. Zones of major earthquakes in the eastern United States are far away, and have not had an appreciable effect at the site. Evaluation of tectonic structures and the historical seismic record indicate a design intensity of VII (Modified Mercalli Scale) is adequately conservative for the site. Intensity VII corresponds to a peak ground acceleration of 0.13 g; for additional conservatism, 0.15 g has been adopted for the SSE.

The principal plant structures are founded on competent bedrock, about 100 feet above the river.

Bedrock at the site, which consists of Triassic siltstone, sandstone, and shale, is moderately to closely jointed, and joints are generally vertical to nearly vertical. Two major joint systems are prevalent, striking N 20 to 50 E, and N 50 to 60 W. The bedding strikes approximately east-west and dips 8 to 20 to the north. P-wave velocities in rock in the area of the major plant structures average about 12,000 ft/sec.

CHAPTER 02 2.5-1 REV. 13, SEPTEMBER 2006

LGS UFSAR Minor faults, inactive since mid-Mesozoic time, occur west and south of the construction area.

Fracture zones with a few inches of offset have been encountered in the excavation; however, they can be attributed to Mesozoic events and are not significant to plant structures. Minor clay seams, which occur locally along bedding, and the fracture zones are treated as required where encountered in the foundation excavations.

Soils at the site, predominantly clayey silts, are residual and range in thickness from zero to approximately 40 feet.

The natural groundwater at the site occurs in joints and along bedding. The emergency spray pond, excavated partly in soil and partly in rock, is lined to preclude excess seepage. The pond will not significantly affect groundwater levels beneath the plant structure.

2.5.1 BASIC GEOLOGIC AND SEISMIC DATA 2.5.1.1 Regional Geology 2.5.1.1.1 Regional Physiography and Geomorphology The site is located in the Triassic Lowland section of the Piedmont physiographic province, as shown on Figure 2.5-1. The Piedmont is bounded on the southeast by the Coastal Plain and on the northwest by a projection of the New England Upland (the Reading Prong) and by the Valley and Ridge province (References 2.5-2 and 2.5-3).

The Valley and Ridge province is characterized by folded Paleozoic sedimentary rocks, which form a series of generally northeast-southwest trending parallel ridges. The Reading Prong of the New England Upland is underlain by Precambrian crystalline rocks. The Coastal Plain is underlain by sedimentary formations of Cretaceous and Tertiary ages.

The northeast-southwest trending Piedmont province is an eroded plateau of low relief and rolling topography. The surface of the plateau slopes gently to the southeast. The Piedmont is divided into an upland and a lowland section. The upland section is underlain by metamorphosed sedimentary and crystalline rocks of Paleozoic and Precambrian age (Reference 2.5-4). The rocks are relatively resistant to erosion, and they support an irregular, hilly surface. The higher hills are capped by Cambrian quartzites and Precambrian crystalline rocks, while broad valleys characterize areas underlain by limestone and calcareous shales. The less rugged lowland section, in which LGS is located, is north and west of the Piedmont uplands and is formed largely on shales and sandstones of Triassic-age. Ridges in the Triassic Lowland trend northeast-southwest along the strike of the more resistant bedrock formations. Higher and more rugged terrain exists where these formations have been intruded by diabase dikes and sills (Reference 2.5-4).

2.5.1.1.2 Regional Stratigraphy In the region surrounding the site, the rocks exposed at the surface form lithologic belts that correspond to the physiographic provinces and sections. These rocks include crystalline igneous materials, metamorphic sequences, and sedimentary formations. The distribution of the geologic units of the region surrounding the site is shown on Figure 2.5-2. The stratigraphic relationships and thickness of the various formations are shown on Figures 2.5-3 and 2.5-4.

CHAPTER 02 2.5-2 REV. 13, SEPTEMBER 2006

LGS UFSAR The oldest rocks of the region are of Precambrian age. They occur to the northwest of the site in the Reading Prong, and to the south and southwest in the upland section of the Piedmont and the Blue Ridge. These rocks are highly metamorphosed and faulted and have been intruded by a wide range of igneous materials. The Precambrian rocks, as shown on Figure 2.5-2, have been grouped into a single unit. The principal formations within this unit include the Baltimore and Pickering gneiss, and the Franklin limestone and marble. Other Precambrian rocks include granite gneiss and hornblende gneiss which underlie large areas of the Reading Prong. Gabbroic gneiss and gabbro in the upland section of the Piedmont are also of Precambrian age.

The Glenarm Series is a group of metamorphosed sedimentary rocks. In the region around the site, the Glenarm Series is represented primarily by the Wissahickon Formation which consists of schist and gneiss. In early geologic publications the Glenarm Series was identified as Precambrian because of the high degree of metamorphism of these rocks and their intrusion by igneous materials unknown in recognized Paleozoic formations. However, more recent studies (Reference 2.5-5) which include radiometric age determinations, show the Glenarm Series must be younger than 1100 million years (age of the basement gneiss) and older than 550 million years, the age of intrusives that cut the Wissahickon Formation. The Glenarm Series is therefore either late Precambrian or early Cambrian.

Paleozoic sedimentary rocks are exposed in the Piedmont upland, the Reading Prong, the Valley and Ridge, and the Appalachian Plateau provinces within the region. The occurrence of Paleozoic rocks is shown on Figure 2.5-2. Rocks of early Cambrian age consist primarily of quartzites, while the higher Cambrian and Ordovician formations are characterized by shales, limestones, and dolomites. The quartzites are prominent ridge formers in the Piedmont Upland and occur as outliers on the Precambrian rocks of the Reading Prong. The Upper Cambrian and Ordovician calcareous formations constitute what is known as the valley limestones and occur mainly in the lower-lying areas.

Sandstone, conglomerate, and shale with some limestone begin the Silurian sequence, giving way to carbonate formations in the upper units. The Devonian consists of shale and sandstone, with some limestone in the Middle and Lower Devonian.

The Mississippian rocks are divided into the Pocono Group, which is predominantly conglomerate and sandstone with some shale, and the Mauch Chunk Formation, typically consisting of red shales, but also consisting of sandstone and some limestone.

Sandstone and conglomerate are the predominant rock types of the Pottsville Group, oldest of the Pennsylvanian units in the region. Some coal is present in the Pottsville group, but most of the minable coal in the region occurs in the overlying units, which consist of cyclic deposits of shale, clay, coal, limestone, and sandstone. Similar deposits extend through the Permian; in some areas the deposition was continuous, with no clearly defined boundary between Pennsylvanian and Permian time.

Rocks of Triassic-age are contained in basins that parallel the northeast-southwest structural trend of the region. The Triassic basins of eastern North America occur from Canada to the Carolinas.

The site is located in one of these Triassic basins, the Newark-Gettysburg Basin, which extends from the Hudson River across north-central New Jersey, through a portion of southeastern Pennsylvania, and into Maryland (Figure 2.5-38). The basin reaches a maximum width of 32 miles in Bucks County, Pennsylvania, near the Delaware River.

CHAPTER 02 2.5-3 REV. 13, SEPTEMBER 2006

LGS UFSAR The Triassic rocks characteristically occur as red shales and sandstones, which are locally interbedded with basalt flows and intruded by diabase dikes and sills. The rocks of the Triassic basins have been designated by various authors as the Newark Group.

In the northern (Newark) portion of the Newark-Gettysburg Basin, where the site is located, the Newark Group is divided from oldest to youngest into the Stockton arkosic sandstone, the Lockatong gray shale and argillite, and the Brunswick red shale and sandstone. Of these three units, the Stockton is the only unit that can be classified as a true stratigraphic formation (Reference 2.5-6). The other sedimentary units of this group were deposited more or less contemporaneously, and they interfinger with each other. Since no definite three-dimensional boundaries can be placed on the Lockatong and Brunswick, they have been designated as lithofacies related to the Hammer Creek lithofacies which occurs west of the Schuylkill River.

The outcrop of the Stock Formation is found in a 4 mile wide belt along the southeastern border of the Newark Basin. The outcrop of the formation narrows, and ultimately pinches out to the west, in southern Berks County, Pennsylvania. Another body of the Stock Formation occurs between Center Bridge and Lumberville, along the Delaware River, having been brought to the surface by the Furlong and Flemington faults. The formation extends along the strike about 12 miles southwest of the river, where it is cut off by the Chalfont fault. The rocks of the Stock Formation consist of beds of coarse conglomerate at the bottom, and they range from coarse-to-fine arkosic sandstones to fine-grained sandstones and shales. The source of the sediments of the Stock Formation lay to the south of its present outcrop area. The formation is estimated to be at least 3000 feet thick (Figure 2.5-4). It unconformably overlies Precambrian and Paleozoic rocks to the south, at the margin of the basin. The formation is overlain conformably from east to west successively by the Lockatong, the Brunswick, and, west of the Schuylkill River, the Hammer Creek lithofacies.

The remaining sedimentary rocks of the Newark Group fill the basin from the upper contact of the Stockton Formation to the northern border of the basin. A great alluvial fan, formed by a large Triassic-age river that flowed from the north into the basin, provided the source of these sediments.

The sediments of this alluvial fan were distributed to the northeast into the Newark Basin, and to the southwest into the Gettysburg Basin. The Hammer Creek conglomerates and sandstones in southern Berks County and in the northern portions of Lancaster and Chester Counties are the remnants of the main body of this great alluvial fan. The stratigraphic thickness of the Hammer Creek lithofacies is about 15,000 feet (Reference 2.5-6). This thickness is about the same as, or greater than, the combined thickness of the Lockatong and Brunswick lithofacies that occur in the site area.

The main body of the Lockatong lithofacies overlies the Stockton Formation in a belt with a maximum width of four miles. About 15 miles east of the site, this belt has been offset by the Chalfont fault. The Lockatong thins to the west. It passes about five miles to the south of the site, where the outcrop is only one-half mile wide. North of the main body of the Lockatong, tongues of this unit are interbedded with strata of the Brunswick lithofacies. The uppermost continuous member of the Lockatong pinches out approximately three miles east of the site. The next lower, the Sanatoga member, passes just north of the site and becomes discontinuous.

The Lockatong lithofacies consist of gray shales and massive mudstone; it has been classified as an argillite although it is not a low rank metamorphic facies. The rock may be calcareous, may contain carbonaceous horizons, and often contains mud cracks. The Lockatong was deposited in a CHAPTER 02 2.5-4 REV. 13, SEPTEMBER 2006

LGS UFSAR shallow lake environment. The maximum stratigraphic thickness of this formation is about 3800 feet (Reference 2.5-6).

The Brunswick lithofacies are more closely related to the Hammer Creek, which occurs west of the Schuylkill River, than to the Lockatong. The Brunswick represents the depositional extension of a great alluvial fan. Thus, the western margins of this formation contain more sand members, whereas to the south and east the strata are typically more shaly. The main body of the Brunswick lithofacies extends from the northern border of the basin to its interfingering contact with the Lockatong to the south. To the west of the Schuylkill River, the Brunswick interfingers with the conglomerates and sandstones of the Hammer Creek. The site is located about five miles east of the interfingering contact of the Brunswick and Hammer Creek.

The Brunswick consists of relatively soft red shales, siltstones, and sandstones. The cementing agent is largely iron oxide derived from the hematite contained in the sediments. In areas where the Brunswick approaches the tongues or the main body of the Lockatong, it is sandier and harder.

The stratigraphic thickness of the Brunswick strata is about 7000 feet (Reference 2.5-6).

Along the north border of the basin, the upper Brunswick strata locally consists of coarse fanglomerates derived from minor streams pouring into the basin from the north. As shown on Figure 2.5-2 these broader fanglomerates, as well as the basal conglomerate of the Stockton Formation, appear under the same lithologic symbol as the Hammer Creek.

Two types of igneous rocks occur within the Newark Basin. The majority of these are diabase dikes and sills, which have intruded into the sedimentary sequence. A minor basalt flow has been mapped in the area southwest of Reading, Pennsylvania. Both the basalt flow and the intrusives are of Triassic-age. A large sill to the west of the site was intruded along bedding-plane surfaces in the sedimentary rocks, and locally has "baked" the sediments in a zone up to 500 feet wide.

Numerous dikes of Triassic-age cut the basin rocks, as well as the Precambrian and Paleozoic rocks surrounding the basin. The most prominent of these dikes, the Downington Dike, extends from 11 miles southwest of Downingtown, Pennsylvania, through Sanatoga Station, just north of the site, and continues about three miles to the northeast. It occupies a fracture zone, and in places it enters zones of weakness provided by pre-existing faults. The major mineral constituents of the Triassic diabase are augite and labradorite. The diabase intrusions postdate Triassic sedimentation and tectonics, as discussed in the following sections on structural geology and geologic history. Radiometric ages obtained during geologic investigations for LGS (see section 3.2.3 and 3.7.1 of Reference 2.5-1) indicate some of the diabase dikes near the site are probably of Jurassic-age.

Cretaceous sedimentary deposits occur in the Coastal Plain province. These sediments consist of clays and sands, which are moderately to poorly consolidated.

Deposits of Cenozoic age consist of Tertiary sediments and terrace deposits, Quaternary gravels, sandy gravels, glacial drift, and Recent alluvium. Tertiary and Quaternary deposits occur in the Coastal Plain province.

Glacial deposits of Pleistocene age cover large areas of the bedrock surface north of the site, but do not extend into the area adjacent to the site.

CHAPTER 02 2.5-5 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5.1.1.3 Regional Geologic Structure The dominant structural feature in the region surrounding the site is the Appalachian Orogenic Belt.

This belt is marked by northeast-southwest orientation of the axes and alignment of most of the structural features and stratigraphic contacts. The major structural features of the region are shown on Figure 2.5-5. Figure 2.5-38 is a compilation from the recent literature (References 2.5-10, 2.5-60, 2.5-17, 2.5-112, and 2.5-119) showing faults mapped within and in the vicinity of the Newark-Gettysburg Basin. Major faults and structures within 5 miles of the site are shown in more detail on Figure 2.5-6. A Bouguer gravity anomaly map of the region is shown on Figure 2.5-7, and aeromagnetic intensities for the Limerick-Pottstown area are shown on Figure 2.5-8.

The Paleozoic and Precambrian rocks of the Piedmont Uplands to the south and west of the site, and of the Reading Prong and the Great Valley section to the north, are structurally complex.

These strata are intensely folded and faulted. The axes of foliation, and of the major folds involving these rocks, generally strike to the northeast. Precambrian and Lower Paleozoic rocks south of the site are strongly metamorphosed, and they exhibit tightly folded structures superimposed on broader synclinoria and anticlinoria. The Paleozoics to the west, and in particular those to the north of the site, exhibit recumbent folds, with basal truncation or decollement by thrust faults.

The majority of the faults that displace Precambrian and Paleozoic strata are thrust faults. The thrust sheets are relatively thin, and they generally exhibit movement to the northwest. The north-south trending faults at their flanks are tear faults. These faults do not involve Mesozoic or Cenozoic strata. To the west of the site, in the Piedmont Uplands, some of the major thrust faults are transected by Triassic-age diabase dikes that show no displacement. Similar relationships of Triassic dikes crossing Paleozoic faults are found in the Readin Prong area, north of the site. The last movement along these faults probably occurred over 200 million years ago, and certainly no later than 140 million years ago, based on the absence of displacement of the Triassic diabase dikes.

A major postulated fault system of Paleozoic age in the region is the Martic Line of thrust faulting.

The Martic Line was introduced in earlier geologic publications (Reference 2.5-7), primarily to account for the fact that the Glenarm Series (thought to be of Precambrian age) overlies sedimentary rocks of known lower Paleozoic age in southern Lancaster County, Pennsylvania.

The fault was subsequently drawn for hundreds of miles along the Appalachians as the boundary of the inner and outer Piedmont. Cloos and Hietanen (Reference 2.5-8) prove imbricate thrusting in the type area. Wise (Reference 2.5-9) concluded that the Martic Line is a "complex polygenetic feature made up of distinct segments...." He indicates that the Martic Line is a Paleozoic feature, which may be largely a facies boundary marking the edge of a deeply subsided basin to the southeast. Along this basin margin there has been imbricate thrusting, regional metamorphism, brittle movement of large basement blocks, widespread development of kink bands and joints, and local intense folding. Wise states: "The final event was the injection of basaltic dikes of probable Triassic-age with north to northeast trend across the region."

The geologic map of Pennsylvania (References 2.5-10 and 2.5-60) and the tectonic map of the central and southern Appalachians (References 2.5-11) show Triassic dikes crossing the Martic Line without offset, indicating that no movement has occurred in this zone for at least 140 million years.

The major recognized Paleozoic faults in the region occur to the south and west in the Piedmont Uplands, and to the north in the Reading Prong and Great Valley sections. The Huntington CHAPTER 02 2.5-6 REV. 13, SEPTEMBER 2006

LGS UFSAR Valley-Cream Valley and Rosemont faults occur in the Piedmont Uplands, 15 miles south of the site. The major structural features in the Piedmont Uplands west of the site are, from south to north, the Mine Ridge anticline, Brandywine Manor fault, Welch Mountain anticline, and the Elverson thrust fault (Reference 2.5-12). Of these features, the nearest to the site is the Brandywine Manor fault, which lies nine miles to the southwest. None of the major Paleozoic faults has a mapped extension to the east of the Triassic border or into the Newark-Gettysburg Basin.

To the north in the Reading Prong and Great Valley sections, nappe structures exhibited by the strata greatly complicate their interpretation. An interpretation presented for the Valley and Ridge province to the northwest (Reference 2.5-13) is that the Paleozoic rocks have been transported westward along nonoutcropping, low angle, detachment thrust faults. The group of rocks observed in any one locality represents an allochthon, entirely unrelated stratigraphically to the surrounding rocks. This hypothesis is supported by deep bore hole and geophysical data, which indicate that the basement below the relatively thin skin of folded and faulted Paleozoic sedimentary rocks does not reflect connected or similar structures (References 2.5-60, 2.5-30, 2.5-87, and 2.5-89).

Over the past several years additional deep seismic reflection studies have suggested that these various thrust faults conveying the allochthonous slices westward are splays of a deeper, regionally extensive detachment surface or decollement which may be a fundamental tectonic structure of the central and southern Appalachians, extending eastward beneath the Piedmont province and possibly continuing beneath the Coastal Plain and Atlantic shelf (References 2.5-60, 2.5-75, 2.5-76, 2.5-77, 2.5-78, 2.5-88, 2.5-97, 2.5-64, 2.5-94, 2.5-68, 2.5-69, 2.5-121, 2.5-122, and 2.5-82).

The seismic profiles from Pennsylvania, Virginia, eastern Tennessee, South Carolina, and Georgia show laterally continuous subhorizontal reflectors ranging in depth from less than 6 km in the Valley and Ridge to more than 11 km beneath the Coastal Plain. Cook (Reference 2.5-79) and Harris and Bayer (Reference 2.5-88) interpret these subsurface reflectors to be elements of a regionally extensive decollement underlain by relatively undeformed, unmetamorphosed, layered Paleozoic sedimentary rocks. The subhorizontal east-sloping decollement presumably accommodated crustal foreshortening associated with continental plate convergence in late Paleozoic time when metamorphic Precambrian and early Paleozoic rocks of the Blue Ridge and Piedmont were thrust to the west and northwest over relatively unmetamorphosed sedimentary rocks of the Valley and Ridge.

The eastern extent of the decollement is still poorly known. The seismic reflection profiles confirm the presence of a decollement from the Valley and Ridge to at least as far as the inner (western)

Piedmont where the subsurface reflectors thicken, become discontinuous, and dip steeply to the southeast (References 2.5-75, 2.5-76, 2.5-77, and 2.5-78). In one interpretation, the steeply dipping reflectors mark the root zone of the decollement beneath the outer Piedmont (References 2.5-92, 2.5-93, and 2.5-94). Reflections from the coastal plain and continental shelf, however, suggest that the decollement may extend southeastward to the coast (References 2.5-78, 2.5-88, 2.5-68, 2.5-69, and 2.5-122). In the second interpretation, the steeply dipping reflectors are interpreted as sedimentary deltaic deposits accumulating on the shelf of an autochthonous proto-North American continental margin (Reference 2.5-78). Cook et al (Reference 2.5-78) suggest that the decollement may be rooted offshore where a complex structural configuration is evident on the seismic profiles.

A recent reinterpretation (Reference 2.5-97) of these seismic reflection profiles supports the interpretation by Hatcher and his coworkers (References 2.5-90, 2.5-91, 2.5-92, 2.5-93, and 2.5-94). Based on palinspastic restorations of the crust, they suggest that the decollement is probably rooted beneath the Kings Mountain belt of the Piedmont province. Subhorizontal reflectors farther to the southeast described by Cook et al (Reference 2.5-78) as a decollement CHAPTER 02 2.5-7 REV. 13, SEPTEMBER 2006

LGS UFSAR probably represent subhorizontal ultramylonites or layered intrusions related to the internal structure of an accreted island arc (Reference 2.5-97).

Age constraints for displacement on the decollement are not well established. Many investigators believe that the principal displacement occurred during the Alleghenian orogeny and that it is responsible for the thin-skinned fold and thrust deformation of the Valley and Ridge Province, emplacement of the crystalline Blue Ridge and Piedmont rocks as thin, allochthonous sheets, and the extensive metamorphism and plutonism along the eastern margin of the Appalachians (References 2.5-92, 2.5-94, 2.5-64, and 2.5-93). Although principal displacement probably did occur during the Alleghenian, Harris and Bayer (Reference 2.5-88) suggest that the decollement is also long-lived and may have grown intermittently from the late Precambrian to the late Paleozoic.

Based on radiometric age dates, paleomagnetic data and geologic field relations, however, Ellwood et al (Reference 2.5-81) believe that a 350 million year old intrusive complex was emplaced through the decollement after major thrusting. The 350 million year old age for the intrusion would thus provide a minimum age for thrusting.

A second conspicuous feature indicated by both deep and shallow seismic reflection profiles across the Appalachian orogen is the presence of upwardly concave reflectors beneath the Piedmont and Coastal Plain. Most of the reflectors are listric into the suspected decollement and may represent late Paleozoic thrust faults, Mesozoic listric normal faults, or both (Reference 2.5-78). In this interpretation, the Brevard zone is a splay off the main decollement (Reference 2.5-77). There is some uncertainty as to whether, and to what extent, late Mesozoic crustal rifting was accommodated by reactivation of these listric faults as normal basin faults. It seems reasonable to presume that at least some Mesozoic basin faults are deep-seated, as evidenced by basaltic magmatic activity, in which case they would displace and disrupt the Paleozoic decollement surface.

A postulated major east-west fault zone, named the Cornwall-Kelvin wrench fault, has been inferred by Drake and Woodward (Reference 2.5-14) on the basis of subsea topography and geophysical surveys and has been postulated to extend through the Triassic Lowland of southeastern Pennsylvania. However, there is neither geological nor geophysical evidence of a westward continuation of this fault into the continent, either at the surface or at depth (Reference 2.5-15).

The strata of the Newark Group, unlike those of the older Paleozoic rocks, are only locally and gently folded. In general, the rocks dip between 5 and 20 to the north and northwest. The Brunswick lithofacies and the upper tongues of the Lockatong are involved in broad cross-folded structures north and east of the site. The Triassic strata are characteristically faulted by normal strike and transverse faults, some of which show considerable lateral displacement. Unlike most of the Paleozoic thrusts, the Triassic faults are high angle gravity faults, with the downthrown side usually to the south.

The faults of Triassic-age have generally not displaced the younger Cretaceous and Cenozoic strata in the Coastal Plain to the southeast. According to a recent compilation (Reference 2.5-134, plate 1; Reference 2.5-135, figure 3), the only instance of a fault which offsets coastal plain strata overlying Triassic sediments within 200 miles of the site is the inferred Brandywine fault zone located 9 miles southeast of Washington, D.C., about 140 miles from the site. Evidence for both the faulting and the presence of Triassic sediments at depth is entirely from subsurface data, including seismic reflection, drill holes and geophysical logging, as reported by Jacobeen (Reference 2.5-98). These data indicate that two northeast striking, southeast dipping en echelon reverse faults offset the top of the Lower Cretaceous Arundel Formation a maximum of about 250 feet. The relation of this faulting to Triassic basin structure, however, is uncertain because both CHAPTER 02 2.5-8 REV. 13, SEPTEMBER 2006

LGS UFSAR Triassic sediments and granitic rocks were penetrated by deep drill holes on both sides of the fault zone. (It should be noted that Jacobeen (Reference 2.5-98) states he believes the faults to be unrelated to Mesozoic structures.) Up section, the drill hole data confirm that the Paleocene-Eocene Aquia Formation is monoclinally folded without being offset by faulting. This folding occurred prior to Oligocene erosion and only minor flexuring is displayed in overlying Miocene strata (Reference 2.5-98). Jacobeen concludes that there is no evidence for post-Miocene movement on the Brandywine fault zone.

Mixon and Newell (References 2.5-105, 2.5-106, and 2.5-107) suggest that the Brandywine fault zone may be an extension of faults bordering the Richmond Triassic basin in Virginia, 60 miles farther to the southwest, based on the alignment of both fault zones with a linear gravity anomaly that extends between the two locations. However, evidence to associate the Richmond basin faults with Cretaceous or younger faulting apparently could not be found because neither Mixon and Newell (References 2.5-105, 2.5-106, and 2.5-107) nor Wentworth and Mergner-Keefer (References 2.5-134 and 2.5-135) indicate any offset of coastal plain strata where these strata overlap the Richmond basin faults.

Offshore, reported instances of significant Cenozoic displacements within 200 miles of the site are rare. Hutchinson and Grow (Reference 2.5-96) report offset of the basement-Cretaceous contact of up to 280 feet on a 20 mile long, north-northeast trending fault (New York Bight fault) off the northern New Jersey coast, about 100 miles east of the site. However, evidence for post-Cretaceous offset on the fault is highly ambiguous, due to poor seismic resolution in the shallow section, lack of stratigraphic control, and the upward-decreasing amount of possible offset.

Hutchinson and Grow (Reference 2.5-96) state that their data are not sufficient to determine the dip of the fault or whether it is reverse or normal. They conclude the data show "potential offset" in the Tertiary and possible "warpage" of strata they infer to be Quaternary in age.

Another fault zone about 15 miles west of, and parallel to, the Brandywine has been named the Stafford fault zone (References 2.5-105, 2.5-106, and 2.5-107). Surface exposures of the Stafford fault zone show that coastal plain strata are offset along high angle basement faults. Although the Stafford fault zone is not associated with Triassic sediments and has no demonstrable relation to any Triassic basin structure, the relative proximity, similar trend, and presence of reverse sense of displacement in both the Brandywine and the Stafford fault zones suggest they may be structurally related.

The Stafford fault zone consists of four subparallel, en echelon basement faults which offset coastal plain strata. Each fault has a maximum throw of about 150 feet. Two of the faults display reverse throw on high angle, west dipping planes (down-to-coast movement).

On one of these faults (the Fall Hill fault) an exposure near the Stafford-Spotsylvania county line shows a high angle fault contact between basement gneiss and Cretaceous coastal plain sediments. The gneiss projects about 14 inches upward into overlying Plio-Pleistocene fluvial gravels, consistent with reactivated reverse displacement along the fault-plane (Reference 2.5-107, figure 7). On the other hand, such relationships are also consistent with features produced by normal stream erosion, particularly in view of the great contrast in erodability between the gneiss and the Cretaceous sands.

A detailed investigation of the Stafford fault zone, including an extensive drilling and trenching program, was conducted by Dames and Moore for Potomac Electric Power Company (Reference 2.5-113). Tertiary strata were identified above the faulted Cretaceous strata in trenches across one of the four faults (Hazel Run fault). The trenches show that this fault truncates Paleocene-Eocene CHAPTER 02 2.5-9 REV. 13, SEPTEMBER 2006

LGS UFSAR strata (Aquia Formation) and is overlain by unfaulted Miocene strata (Calvert Formation) which exhibit only minor flexure not necessarily caused by fault movement. Significantly, all three trenches excavated to the base of the alluvial gravels across the trace of the Fall Hill fault disclosed that the gravels are not offset by the fault, strongly suggesting that the feature described by Mixon and Newell is erosional. Therefore, no unequivocal evidence was found for post-Miocene movement on any of the faults; latest movement could be as early as Paleocene or early Eocene.

A discussion of surface faulting is presented in Section 2.5.3 and was presented in detail in a report by Dames and Moore in 1974 (Reference 2.5-1).

2.5.1.1.4 Geologic History The geologic history of the region can be traced from the Precambrian; the sediments that were deposited to form the Baltimore and Pickering Gneisses and the Franklin Limestone were subjected to magmatic intrusion, metamorphism, and erosion before the onset of Cambrian time.

In late Precambrian, Cambrian, and early Ordovician time, a thick sequence of sediments accumulated. They were derived from a highland that lay southeast of the region and were deposited in a broad inland sea. During Cambrian and early Ordovician time, thick sequences of sand and calcareous and dolomitic ooze accumulated. In the latter part of Ordovician time, the region was subjected to folding and uplift as a result of the Taconic orogeny. Farther to the northwest, sediments were deposited throughout the remainder of the Paleozoic Era.

The major tectonic events during the Paleozoic era began with the Taconic orogeny during the Middle Ordovician, continued with the Acadian orogeny in late Devonian, and culminated in the Appalachian orogeny at the end of the era. Major folding of the strata occurred predominantly during the Appalachian orogeny. The less competent units in the Middle Cambrian, Upper Ordovician, and Upper Silurian became thrust zones (Reference 2.5-16). Thrusting is discussed in Sections 2.5.1.1.3 and 2.5.1.1.4.1.

From the close of the Paleozoic to the present, the region has remained above sea level, and most of the area has undergone erosion. During the late Triassic, a series of down-warped or down-faulted basins developed in the central portion of the Appalachian Highlands. Streams from the higher areas adjacent to these basins filled them with the alluvial deposits that make up the Triassic sedimentary rocks of the region. After sedimentation ceased in the Newark Basin, the Triassic strata were tilted to the northwest along a synclinal hinge line near the north border of the basin. The faulting and broad folding of the strata occurred during or after the tilting.

The igneous activity in the Triassic occurred in three stages. The first stage was the extrusion of basaltic lava, which occurred during Triassic sedimentation. The second stage was the intrusion of diabase sills. The sills are cut by later Triassic faulting, and therefore they predate the development of the major fracture system of the surrounding rock. The final stage was the intrusion of diabase dikes along the major fractures. Most of the dikes in the region are not faulted and represent the final activity affecting the Triassic strata. Paleomagnetic age dating of these Triassic dikes indicates that they were emplaced between 180 and 140 million years ago.

Radiometric dates obtained during geologic investigations for the site are in good agreement, ranging from 151 to 198 million years (see Reference 2.5-1, section 3.7).

During Jurassic and early Cretaceous time, the region underwent a prolonged period of erosion.

The ancestral patterns of the present day drainage system of the region probably began during the late Triassic or Jurassic. Early in the Cretaceous, the ocean encroached on the Coastal Plain to CHAPTER 02 2.5-10 REV. 13, SEPTEMBER 2006

LGS UFSAR the southeast, and alluvium and marine sands, marls, and clays were deposited. These deposits constitute the unconsolidated Cretaceous and Tertiary sedimentary formations of the Coastal Plain.

There is no record in the eastern United States of any major tectonic event in the interval between Cretaceous time and the present. The region was subjected to broad, gentle periodic warping of the crust.

During the Pleistocene epoch, the maximum advance of the continental glaciers reached as far as the northern portion of the Reading Prong, about 20 miles to the northeast of the site. During periods of glacial melting, the major rivers of the region were choked with great quantities of glacial outwash. These materials were deposited along the valleys of the major streams and remain today as terrace deposits. The glaciated region was north of the Schuylkill River headwaters, and the river was little affected by continental glaciation.

2.5.1.1.4.1 Appalachian Plate-Tectonic History The geologic history of the region summarized above can be stated in terms of lithospheric plate-tectonic interactions based on the considerable geologic research reported in the current technical literature on Appalachian tectonics (References 2.5-26, 2.5-30, 2.5-75, 2.5-76, 2.5-77, 2.5-78, 2.5-88, 2.5-92, 2.5-93, 2.5-94, 2.5-64, 2.5-97, and 2.5-138).

In the late Precambrian (approximately 820 million years ago) extensional tectonics rifted proto-North America from proto-Africa, proto-Europe and proto-South America (References 2.5-116 and 2.5-117). Various continental fragments were distributed in the expanding proto-Atlantic or "Iapetus" Ocean as the major continents separated (References 2.5-116, 2.5-76, 2.5-77 and 2.5-92). Some major continental fragments include the Avalon and Inner Piedmont-Blue Ridge provinces. During the ensuring plate convergence or contraction of the Iapetus Ocean, several island arcs developed offshore of the continental margin. The Carolina slate belt, for example, is believed to be one of these island arcs subsequently accreted to the margin of North America.

In the southern and central Appalachians, closure of the oceanic basin between North America and the Inner Piedmont-Blue Ridge fragment and subsequent collision of the land masses during the Cambrian and Ordovician is recorded by the Taconic orogeny (480-450 million years ago). The Inner Piedmont-Blue Ridge fragment was accreted to North America by overthrusting (References 2.5-117, 2.5-91, and 2.5-76). Extreme deformation, metamorphism and plutonism in the Inner Piedmont accompanied the orogeny. Six hundred to 500 million year old volcanics on the offshore Carolina slate belt suggest that eastward subduction and development of a volcanic arc on the slate belt accompanied closure of the basin (Reference 2.5-91). In western New England, the Taconic orogeny produced extensive thrust and fold deformation, low grade regional metamorphism, gravity-slide nappe structures, and extensive granodioritic and ultramafic intrusions (References 2.5-26, 2.5-30, and 2.5-133).

Continued closure of the Iapetus Ocean by eastward subduction resulted in the collision of the Carolina slate belt with the Inner Piedmont-Blue Ridge terrain of the southern and central Appalachians during the Acadian orogeny (400-300 million years ago) (Reference 2.5-92). Rocks of the Kings Mountain-Charlotte belt probably represent allochthonous portions of the intervening oceanic crust caught in the collision and obducted onto the margin of North America by overthrusting. The orogeny resulted in additional extensive metamorphism and deformation. The allochthonous Inner Piedmont-Blue Ridge fragment was thrust further westward over the old CHAPTER 02 2.5-11 REV. 13, SEPTEMBER 2006

LGS UFSAR continental margin. In the northern Appalachians, extensive high grade regional metamorphism and granitic intrusions accompanied the orogeny.

The last phase of plate convergence along the margin of North America culminated in the continental collision between proto- North America and proto-Africa during the Alleghenian orogeny (300-250 million years ago). Most of the prominent southern and central Appalachian structures and provinces assumed their present configuration in this orogeny. Extensive igneous activity in the eastern Piedmont accompanied the orogeny and large-scale overthrusting deformed the sedimentary rocks in the Valley and Ridge Province (References 2.5-76 and 2.5-78). Thrusting of the Piedmont over the Blue Ridge also created the Brevard zone at this time. In the northern Appalachians, the Alleghenian orogeny is manifest primarily in southeastern New England by folding, low and medium grade local metamorphism and granitic intrusions.

In the Triassic, following the Alleghenian orogeny, extensional tectonics again dismembered a megacontinent (Pangaea). Rifting separated North America from Africa, Europe and South America, forming the present Atlantic Ocean in its wake. Extensional Triassic-Jurassic basins bordered by normal faults (such as the Newark-Gettysburg Basin) were superimposed on the Paleozoic compressional structures. As rifting continued, Cretaceous and Tertiary sediments of the Coastal Plain province accumulated on the trailing eastern margin of North America. The composition of these sediments and their gentle dip away from the Appalachian Mountains implies that the Appalachians have stood as an eroding structural high for over 200 million years (Reference 2.5-84). Although many of the Triassic-Jurassic basins are exposed at the surface in the Piedmont Province, most are buried beneath the Coastal Plain sediments and have been tentatively located on the basis of drill holes (Reference 2.5-98), aeromagnetic and gravity anomalies (Reference 2.5-100), and seismic reflection and refraction profiles (Reference 2.5-81).

In summary, the Appalachians evolved principally by a sequence of plate rifts and collisions during the late Precambrian and Paleozoic eras. Subsequent rifting in late Triassic and early Jurassic time initiated the latest cycle of Atlantic sea floor spreading and produced the present configuration of the continental margin.

2.5.1.2 Site Geology 2.5.1.2.1 General The site is located approximately 3 miles southeast of Pottstown, Pennsylvania, adjacent to the Schuylkill River, as shown on Figure 2.5-6. The principal plant structures are located on a broad ridge, approximately 100 feet above the river. Bedrock, encountered at shallow depths, consists predominantly of red siltstone, sandstone, and shale of late Triassic-age. The soils are residual, derived from the weathering of the underlying bedrock. Minor Triassic-age faults, inactive since Middle Mesozoic time, occur to the west and south of the construction area (see Sections 2.5.1.1.3 and 2.5.3). Fracture zones with a few inches of offset were encountered in the excavation.

However, they are not significant to the plant structures.

2.5.1.2.2 Physiography The topography of the site area is characterized by broad, gently rolling ridges that are dissected by the entrenched courses of the Schuylkill River and its tributaries. Elevations in the area range from a high of 200-300 feet above sea level on the ridge tops to a low of 100 feet along the Schuylkill River. Higher elevations occur in the more rugged terrain to the north and west of the site.

CHAPTER 02 2.5-12 REV. 13, SEPTEMBER 2006

LGS UFSAR Topography and drainage of the area are controlled to a large degree by the lithologic and structural characteristics of the bedrock. The ridges generally trend east-west along the strike of the bedrock. The prominent ridge tops are capped by the more resistant rocks: sandstone, argillite, and diabase. Higher order drainage courses trend along the major joint directions.

Tributary streams in the area (Sanatoga Creek, Possum Hollow Run, and Brooke Evans Creek) flow along the strike of the bedding or along major joints and fracture zones.

2.5.1.2.3 Stratigraphy Bedrock at the site includes both the Brunswick and Lockatong lithofacies. Bedding dips toward the north at 8 to 20. In the power block area, the dip is approximately 10 to 18 to the north.

Contours on top of rock are shown on Figure 2.5-9.

Strata of the Brunswick lithofacies underlie most of the area. This rock unit is several thousand feet thick and is basically a reddish-brown siltstone, which grades locally to a shale. Calcite commonly fills joints and bedding-planes in the rock. Sandstone beds interfinger with the siltstones and range from reddish-brown, very fine-grained, silty sandstone to light gray, and medium-to-coarse-grained sandstone. Lateral gradation between sandstones and siltstones is common throughout the area.

Geologic sections through the site area are shown on Figures 2.5-10 and 2.5-11. The stratigraphic column at the site is shown on Figure 2.5-12.

The Lockatong lithofacies, represented by the Sanatoga Member, interfinger with the Brunswick in the northern part of the site area. The Sanatoga Member of the Lockatong lithofacies consists of bluish-gray, calcareous argillite, with two distinct beds of black carbonaceous shale. The Sanatoga member occurs in the spray pond area, but does not occur beneath the cooling towers or the main plant structures.

Diabase has been intruded into the sedimentary rocks along major fractures, trending about N 20 to 30 E. The most extensive of these intrusives is the Downingtown Dike, about 1300 feet west of the site. The rocks at the borders of the dikes have been baked to a tough gray hornfels.

For the most part, soils consist of red sandy and clayey silts, with numerous rock fragments derived from the underlying bedrock. Soil thickness ranges from 0-40 feet with an average of 10-15 feet. A typical soil section consists of the following:

0-5 feet - Reddish-brown clayey silt 5-10 feet - Reddish-brown clayey silt with numerous rock fragments 10-20 feet - Highly weathered and fractured rock with some silts and clays Residual soils overlying the diabase and hornfels are usually thicker and contain fewer rock fragments than the section described above.

Typical flood plain deposits occur along the Schuylkill River. They consist of silt, sand, and gravel that have been transported from upstream areas.

2.5.1.2.4 Structure The principal structural features and relationships in the site area are shown on Figures 2.5-6 and 2.5-13. The bedrock strata dip to the north at between 8 to 20. Two major joint directions are CHAPTER 02 2.5-13 REV. 13, SEPTEMBER 2006

LGS UFSAR prevalent in the area. Both are nearly vertical; they strike approximately N 20 to 50 E and N 50 to 60 W. Several fracture zones with minor offsets of Jura-Triassic age were encountered in foundation excavations and are described in Section 2.5.1.2.5.

As previously indicated, the site lies within a down-warped Triassic basin. Subsequent to the deposition and consolidation of the basin sediments, the region was uplifted, and the brittle basin materials were broken by numerous small faults and fractures.

Three faults, the Sanatoga, the Brooke Evans, and the Linfield, occur within 2 miles of the site.

These faults are shown on Figure 2.5-6 and discussed in Section 2.5.3. The nearest approach of the Sanatoga fault to the reactor enclosure area is 1300 feet to the west. Vertical displacement on the fault is approximately 290 feet, with the downthrown side to the east. The fault-plane is intruded by Triassic diabase, which is part of the Downingtown Dike.

The Brooke Evans Fault passes within 2800 feet to the south of the plant area and trends approximately N 50E. The apparent vertical displacement is about 350 feet, with the downthrown side to the south.

The Linfield fault is located east of the town of Linfield, about 2 miles southeast of the site. Field investigations in 1974 (Reference 2.5-1) identified five vertical or nearly vertical faults, trending N 40 to 55 E, with apparent vertical offsets of from 1-20 feet.

All three of these faults, the Sanatoga, the Brooke Evans, and the Linfield, are associated with the Jura-Triassic events that occurred 140-200 million years ago (Section 2.5.3.2).

2.5.1.2.5 Fracture Zones in the Site Area Three fracture zones were encountered during excavation of the main power block. Figure 2.5-13 shows their locations, widths, and attitudes at final foundation grades. None of the zones presents a hazard to the structures. The zones have relatively lower strength than the rock adjacent to them; consequently, they were treated to preclude any unacceptable differential settlement across them. Treatment of these zones is discussed in Section 2.5.4.12.

A fracture zone (Zone A), consisting of discrete fractures about 20 feet apart with hard, relatively unfractured rock between them, was encountered in the Unit 1 foundation excavation, at the south edge of the reactor enclosure. To the northeast, these fractures converge and intersect other fractures, resulting in a zone of closely fractured, partly weathered rock varying in width from 2-7 feet. This zone extends across the northern part of the reactor enclosure foundation and the entire control structure and turbine enclosure foundations (Figure 2.5-13). No offset was observed along this zone. At the south wall of the control structure, most of the fractures in this zone, and essentially all of the weathering, terminated at a bedding-plane (clay seam 1 described later in this section) at about el 158. At this wall, the fracture zone was removed down to the bedding-plane.

To the northeast, it was treated, where necessary, by dental excavation, as described in Section 2.5.4.12. Treatment of the zone was not required at the south wall of the Unit 1 reactor enclosure, since the "zone" consists only of narrow, widely spaced fractures with hard unweathered rock between them.

Another fracture zone (Zone B) occurs at the southeast corner of Unit 2 reactor enclosure (Figure 2.5-13). It strikes about N 30 E, is essentially vertical, and contains highly fractured, partly weathered rock in a zone about one foot wide. This zone has about 8-10 inches of apparent CHAPTER 02 2.5-14 REV. 13, SEPTEMBER 2006

LGS UFSAR vertical offset, down on the southeast side. Since this fracture zone is vertical and narrow, dental treatment was probably not needed, but was done as a conservative measure, as described in Section 2.5.4.12.

A third zone (Zone C) in the foundation of the Unit 2 turbine enclosure was mapped as a minor fracture zone. Dames and Moore indicate that the zone is similar in character to fracture Zone B, with apparent vertical offsets of up to two feet (Reference 2.5-1, section 2.3.4.3). This zone contains from 1/4-4 inches of clay and decomposed rock. Treatment of this zone was not necessary.

Two clay seams were encountered in the plant excavation. They occur along bedding, striking generally northeast-southwest and dipping 10 to 18 northwest. The first seam is near the top of rock south of Unit 1, where it is about 10 inches thick in the south slope of the excavation. The thickness of this zone decreases rapidly down-dip, toward Unit 1, and is generally less than 11/2 inches thick under Unit 1; in places it is barely discernible. Clay, where present in the seam, contains numerous hard rock particles of small gravel size. Some shearing is evident along this seam.

The second clay seam is stratigraphically lower than seam 1 and intersects foundation grade in part of Unit 2. This seam occurs along shaly beds, which are relatively softer than the adjacent siltstone. The soft material in seam 2 is usually only about 1/4-1/2 inch thick where exposed in foundations or excavated slopes, and it consists of plastic clay with shaly, sheared rock fragments.

Only minor local treatment of these seams was required. The method of treatment is described in Section 2.5.4.12.

Geologic mapping in the spray pond excavation (Figures 2.5-42 and 2.5-43) disclosed only few minor instances of offset along near-vertical joints and fractures. The maximum apparent displacement observed does not exceed 3 inches. These minor offsets observed in the spray pond excavation are consistent with the late Triassic or Jura-Triassic deformational features found elsewhere at the site and in the region around the site. Extensive geologic investigations conducted by Dames and Moore (Reference 2.5-1) have confirmed the noncapability of these deformational features (Section 2.5.3).

Near the western end of the spray pond excavation, a vertical joint trending N 20 E displayed about 1 inch of offset (east side down); the joint was traced a distance of about 100 feet A narrow (2-6 inches wide) east-west zone of closely spaced jointing and fracturing, locally accompanied by more widely spaced jointing, was exposed in the southern part of the pond. It was traced from the western edge of the excavation eastward for a distance of about 600 feet. The maximum observed offset on this feature is no more than 2 inches (south side down); slickensides were not generally apparent but were noted in a few places. These features were examined by an NRC geologist and a geotechnical engineer during their site visit on September 10, 1981. Locally altered and weathered rock within the zone was excavated to sound rock following procedures employed in the main power block excavation (Section 2.5.4.12).

In the rock slope at the eastern end of the spray pond excavation, an east-west fracture zone was exposed that displayed between 2 and 3 inches of total apparent lateral displacement (left-lateral sense) as measured by offsets between matching joints on either side of the fracture. No slickensides were visible along this fracture. The fracture terminated downward at a bed of hard, calcareous siltstone. No trace of the fracture was present on the exposed bedding-plane surface of this underlying siltstone, thus demonstrating the superficial nature of the fracture. Several other CHAPTER 02 2.5-15 REV. 13, SEPTEMBER 2006

LGS UFSAR narrow east-west zones of closely spaced fractures were also mapped in the spray pond excavation, none of which showed evidence of offset.

There is little evidence of shearing in the excavation, despite the proximity of the Sanatoga fault to the western end of the spray pond. As investigated and mapped by Dames and Moore (Reference 2.5-1), the Sanatoga fault - Downington dike passes within about 250 feet of the western end of the spray pond. The N 20 E joint with about 1 inch offset noted above strikes parallel to the trend of the Sanatoga fault. The structural investigations conducted by Dames and Moore recognize east-west fracturing and shearing as a secondary but persistent occurrence accompanying the dominant northeast and northwest structural trends. East-west shearing was mapped at the site in a trench excavated across the trace of the Sanatoga fault. Similar shears were also recorded in exposures of the Downington dike south of the site and in road-cuts about 3 miles northeast of the site.

Dames and Moore concluded that lateral shears occur as a numerically minor east-west regional fracture trend which includes both extension fractures and shears. In several cases these joints appear to have resulted from simple shear along northeast-trending fractures (Reference 2.5-1).

2.5.1.2.6 Site Geologic History The geologic history of the region is discussed in Section 2.5.1.1.4. The Newark-Gettysburg basin, in which the site is located, was formed in two phases (Reference 2.5-17). The first phase consisted of formation of the sedimentary basin, probably by down-warping rather than faulting.

The bulk of the intrusives were emplaced late in basin development, but before the end of deposition. The second phase consisted of a deformational period. Faill (Reference 2.5-17) states:

"All the deformation appears to have been postdepositional, with the monocline, the folds, and the faults developing contemporaneously.... With few exceptions, faults are not intruded with diabase -

in general, the faults offset diabase plutons. This indicates that most of the faulting occurred after the diabase intrusions."

Studies by Dames and Moore (Reference 2.5-1), which include radiometric age dating of the diabase dikes in the vicinity of the site, indicate that the age of the diabase in the site area is 191

+/-10 million years. The age of the faulting is uncertain but is considered by the Geology Review Board, a group of authorities on Appalachian geology, to be associated with the Jura-Triassic development of the Newark Basin 140-200 million years ago (Reference 2.5-1).

Reported instances of Triassic basin faults that also offset Cretaceous and Tertiary deposits are quite rare along the Atlantic coastal plain. These rare occurrences are characterized by relatively small displacements compared to those present in known Triassic basins. All such displacements investigated have been found to be at least 500,000 years old, and no proof has been found for post-Miocene movement. These occurrences are considered to be unrelated to the capability of Triassic faults in the site region.

According to a recent compilation (References 2.5-135, plate 1, and 2.5-134, figure 3), the only instance within 200 miles of the site of a fault that offsets coastal plain strata overlying Triassic sediments is the inferred Brandywine fault zone located 9 miles southeast of Washington, D.C.,

about 140 miles from the site. Evidence for both the faulting and the presence of Triassic sediments at depth is entirely from subsurface data, including seismic reflection, drill holes and geophysical logging, as reported by Jacobeen (Reference 2.5-61). These data indicate that two northeast striking, southeast dipping en echelon reverse faults offset the top of the Lower Cretaceous Arundel Formation a maximum of about 250 feet. The relation of this faulting to CHAPTER 02 2.5-16 REV. 13, SEPTEMBER 2006

LGS UFSAR Triassic basin structure, however, is uncertain because both Triassic sediments and granitic rocks were penetrated by deep drill holes on both sides of the fault zone. Jacobeen (Reference 2.5-61) states he believes the faults to be unrelated to mesozoic structure. Up section, the drill hole data confirm that the Paleocene-Eocene Aquia Formation is monoclinally folded without being offset by faulting. This folding occurred prior to Oligocene erosion, and only minor flexuring is displayed in overlying Miocene strata (Reference 2.5-61). Jacobeen concludes that there is no evidence for post-Miocene movement on the Brandywine fault zone.

Mixon and Newell (References 2.5-105, 2.5-106 and 2.5-107) suggest that the Brandywine fault zone may be an extension of faults bordering the Richmond Triassic basin in Virginia, 60 miles to the southwest, based on the alignment of both fault zones with a linear gravity anomaly that extends between the two locations. However, evidence to associate the Richmond basin faults with Cretaceous or younger faulting apparently could not be found because neither Mixon and Newel nor Wentworth and Mergner-Keefer (References 2.5-135 and 2.5-134) indicate any offset of coastal plain strata where these strata overlap the Richmond basin faults.

Another fault zone about 15 miles west of, and parallel to, the Brandywine has been named the Stafford fault zone (References 2.5-105, 2.5-106, 2.5-107 and 2.5-151). Surface exposures of the Stafford fault zone show that coastal plain strata are offset along high angle basement faults.

However, the Stafford fault zone is not associated with Triassic sediments and has no demonstrable relation to any Triassic basin structure at this location. Nevertheless, the relative proximity and similar trend of the Brandywine and the Stafford fault zones suggest they may be structurally interrelated.

The Stafford fault zone consists of four subparallel, en echelon basement faults that offset coastal plain strata. Each fault has a maximum throw of about 150 feet. Two of the faults display reverse throw on high angle, west dipping planes (down-to-coast movement).

On one of these faults (the Fall Hill fault), an exposure near the Stafford-Spotsylvania county line shows a high angle fault contact between basement gneiss and Cretaceous coastal plain sediments. The gneiss projects about 14 inches upward into overlying Plio-Pleistocene fluvial gravels, consistent with reactivated reverse displacement along the fault-plane (Reference 2.5-107). Such relationships are also consistent with features produced by normal stream erosion, particularly in view of the great contrast in erodability between the gneiss and the Cretaceous sands.

A detailed investigation of the Stafford fault zone, including an extensive drilling and trenching program, was conducted by Dames and Moore for Potomac Electric Power Company (Reference 2.5-113). Tertiary strata were identified above the faulted Cretaceous strata in trenches across one of the four faults (Hazel Run fault). The trenches show that this fault truncates Paleocene-Eocene strata (Aquia Formation) and is overlain by unfaulted Miocene strata (Calvert Formation) which exhibit only minor flexure not necessarily caused by fault movement. The Aquia and Calvert Formations were not present in trenches across the remaining faults; instead, these faults are directly overlain by undisturbed Quaternary materials (generally Plio-Pleistocene upland gravels).

Along the Fall Hill fault, three trenches were excavated to expose the basal contact of the upland gravels crossing the fault. In all three cases, the upland gravels were not disturbed by faulting. The investigation concluded (Reference 2.5-113) that the minimum age of fault displacement on the Stafford fault zone is at least 500,000 years. Moreover, no unequivocal evidence was found for post-Miocene movement on any of the faults. Absence of post-Miocene movement is also CHAPTER 02 2.5-17 REV. 13, SEPTEMBER 2006

LGS UFSAR supported by an investigation by Mixon (Reference 2.5-150), who maps a late Miocene erosional scarp trending across the Stafford fault zone without observable offset.

It is concluded that the indication of Cretaceous and early to middle Tertiary offset above possible Triassic basin faults is a highly exceptional occurrence within the site region and does not warrant the presumption of similar movement on Triassic basin faults in general. Furthermore, detailed investigations of these known exceptional occurrences within the site region have established that their age of latest movement is at least 500,000 years, and is probably Miocene or older.

Therefore, these occurrences are considered to be unrelated to the capability of Triassic faults in the site region.

Since late Mesozoic time, the area around the site has been a land mass subject to erosion.

Continental glaciation that occurred during the Pleistocene did not extend to the site area. Bedrock at the site is overlain by up to 40 feet of residual soil derived from the bedrock by weathering.

These residual soils, which overlie the Sanatoga fault and the Downington Dike without offset, have been dated in studies by Dames and Moore as being of Yarmouthian age, or 500,000-850,000 years before present (Reference 2.5-1, p. 4-17).

Additional discussion related to post-Mesozoic faulting in the site region is provided in Section 2.5.2.3.1.2.

2.5.1.2.7 Engineering Geology Evaluation Site subsurface exploration is described and discussed in Section 2.5.4.3. Laboratory tests of foundation materials and in situ geophysical tests of the foundation materials are discussed in Sections 2.5.4.2 and 2.5.4.4, respectively. Geologic mapping of the foundation excavations is described in Sections 2.5.1.2.5 and 2.5.4.3. It is concluded from these studies and evaluations that the site geologic and foundation conditions are entirely suitable for plant construction and operation.

2.5.1.2.7.1 Geologic Conditions Under Category I Structures All seismic Category I plant facilities are founded on bedrock, except part of the spray pond, portions of the underground piping, and electrical ducts, diesel oil tanks, and valve pits, which are founded on weathered rock, natural soil or fills. For more detail, refer to Section 2.5.4.5. The locations of the major Category I facilities are shown on Figure 3.8-58.

The foundation rock at the site consists of reddish-brown siltstone, interbedded and lensing with shale and sandstone. These rocks are part of the Brunswick and Lockatong lithofacies of Triassic-age (Section 2.5.1.2.3). The bedrock strata dip to the north at angles of from 8 to 20. Several fracture zones with minor offsets were encountered during site excavation; these zones and their treatment are described in Sections 2.5.1.2.5 and 2.5.4.12. All the Category I rock foundations were excavated to unweathered bedrock. Geologic maps and sections of the Category I excavations at the main power block are shown in Figures 2.5-10, 2.5-11, and 2.5-13. Engineering properties of the foundation rock are described in Section 2.5.4.

The natural soils at the site consist of materials derived from the in situ weathering of siltstone, sandstone, and shale. Soil from 0 feet to about 40 feet thick was encountered in the borings at the site. In some parts of the area, the bedrock has not completely broken down into soil, and the soil materials are mixed with weathered, decomposed rock fragments. Weathering decreases with CHAPTER 02 2.5-18 REV. 13, SEPTEMBER 2006

LGS UFSAR depth, and grades gradually to fresh rock. The evaluation of the stability of the natural soils at the site is presented in Section 2.5.4.

2.5.1.2.7.2 Landslide Potential There are no steep or unstable natural rock slopes in the construction area. No old landslides, rock slips, or landslide scars have been noted near plant structures. The natural rock slopes present no hazards to plant structures. Stability of soil slopes is discussed in Section 2.5.5.

2.5.1.2.7.3 Areas of Potential Subsidence, Collapse, or Uplift Rocks in the area around the site are primarily well-indurated siltstones, sandstones and shales to a depth of several thousand feet. No cavernous or karstic terrain exists in the area. There is no mining or significant fluid withdrawal in the area.

Analyses of precise leveling surveys indicate that some broad crustal warping may be occurring in the area (Reference 2.5-18). If so, it is of a broad, regional nature and does not significantly affect plant structures. Measurements of residual stresses (Reference 2.5-1) using overcoring methods indicate very low stress in the area near the site. It is concluded that available data indicate that uplift or subsidence, either from man's activities or natural geologic conditions, does not have any significant effect on the safe operation of the LGS.

2.5.1.2.7.4 Behavior of Site During Prior Earthquakes There is no evidence at the site of any effects, such as landslides, fissuring, or subsidence, that could be attributed to prior earthquakes.

Within historical time, the maximum intensity of earthquakes that have occurred near the site was probably V or less. Ground motion at this intensity has no significant effect on the dense soil and well-indurated rock at the site. For additional information on historical seismicity, see Section 2.5.2.1.

2.5.1.2.7.5 Zones of Deformation or Structural Weakness As reported in the PSAR, the preconstruction investigation defined major joint and fracture systems, minor faults in the area around the site, and indicated the possibility of fracture zones with small offsets within the site area.

During the foundation excavation some fracture zones with small displacements were encountered and were treated as required. Descriptions of the fracture zones are presented in Section 2.5.1.2.5, and their treatment is discussed in Section 2.5.4.12. Figure 2.5-13 shows their locations at final foundation grades, their widths, and their attitudes. None of the zones presents a hazard to plant structures.

2.5.1.2.7.6 Zones of Alteration or Irregular Weathering Bedrock at the site is overlain by from 0-40 feet of residual soil, developed in situ by the gradual decomposition of the parent rock.

CHAPTER 02 2.5-19 REV. 13, SEPTEMBER 2006

LGS UFSAR The soil horizon has developed over a long period of time, probably in excess of 500,000 years (Reference 2.5-1, section 4.0). The soil grades gradually into fresh, unweathered rock; no clearly defined boundary between soil and rock exists. Weathering appears to be a function of lithology and fracture spacing. Where the rock is closely jointed or shaly, weathering has progressed more rapidly, and a thicker soil horizon is present. In the fracture zones, weathering extends below the general level of the fresh rock surface. These zones are generally narrow, such as the fracture zones described in Section 2.5.1.2.5, and were treated locally when encountered in foundation excavations.

Within the site area, the narrow zones of irregular, relatively deeper weathering that were encountered in the foundation excavations were treated by standard methods, as described in Section 2.5.4.12, and pose no hazard to the construction and operation of the LGS plant.

2.5.1.2.7.7 Potential for Unstable or Hazardous Rock or Soil Conditions The bedrock at the site contains no unstable minerals, and the soils derived from this rock are also composed of stable minerals. There are no potentially unstable or hazardous conditions present in the site foundation materials.

2.5.1.2.7.8 Unrelieved Residual Stress In Bedrock Geologic investigations by Dames and Moore in 1974 (Reference 2.5-1) include several measurements of residual stress in rock, made using overcoring methods. These tests were made in closely jointed rock near the Sanatoga fault and in a boring in Possum Hollow Run southeast of the major plant structures. Of the 15 tests made, only two indicate valid results. Dames and Moore state: "Only the results of test FTa-1 and test ET-1 are considered valid (Reference 2.5-1, page 5-4). They show relatively low stresses: major principal stresses of 450 psi and 950 psi, respectively; and minor principal stresses of 175 psi and 650 psi, respectively. The orientation of the major stresses in the two tests is almost identical; N 75 W and N 85 W, respectively."

The report of the geologic evaluation committee, consisting of Dr. Donald U. Wise, Dr. Carlyle Gray, and Dr. Paul B. Myers (Reference 2.5-1), evaluates the residual stress-field as follows: "The present WNW orientation of compressive stresses in the residual stress measurement differs markedly from the NNE compression which produced the Jura-Triassic faults at the site and nearby folds. The effect is to increase normal stresses across the faults, increasing fractional resistance to movement, and decreasing available space for additional graben displacement."

Calcite is present in vein form along joints and fractures in the site area. Despite the fact that calcite twins easily at low stress levels, calcite twins are rare in the latter stages of the mineralization in the site area. The low stress levels indicated by the calcite correspond well with values measured in situ.

It is concluded that the stress regime at the site is low and stable, and residual stress is not a factor in the safe operation of LGS.

2.5.1.2.7.9 Conclusions and Summary Consideration of all the engineering geologic factors at the LGS site leads to the conclusion that the site is suitable for constructing and operating the plant. The bedrock in the construction area is competent and provides satisfactory foundation support for all major plant structures.

CHAPTER 02 2.5-20 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5.1.2.8 Site Groundwater Conditions This section provides a summary of groundwater conditions at the site. Reference for more detailed information on this subject may be made to Sections 2.4.13 and 2.5.4.6.

LGS will have no effect on the future use of groundwater in the region. The site is hydrologically isolated from public groundwater supplies and areas of extensive groundwater development.

Groundwater at the site occurs in sedimentary rocks of Triassic-age. The Brunswick lithofacies, the formation underlying the plant site, is consolidated rock consisting of bedded sandstone, siltstone and shale. It yields small-to-moderate quantities of water to wells. Most of the groundwater movement within the Brunswick follows secondary openings, the most important of which are nearly vertical joint planes that cross each other at various angles throughout the beds, giving the rock a low to moderate permeability. About 85% of the permeabilities measured in bedrock are less than 390 ft/yr.

Recharge to the Brunswick occurs through the relatively impervious soil cover as precipitation percolates down to the water table. The water table approximately parallels the topographic surface, groundwater flows from high to low topographic areas. Groundwater beneath the plant flows toward the Schuylkill River in a southwesterly direction, eventually discharging to the river.

North of the plant, a groundwater divide is present beneath the topographic high that occurs there.

Groundwater north of the divide flows northward, discharging to tributaries of the Schuylkill River.

A map of the potentiometric surface at the site determined from levels measured in May, 1979, is shown in Figure 2.4-15. This indicates the groundwater level is at el 250' east of the spray pond, decreasing through the area of the principal plant facilities to about el 120' southwest of the power block enclosure. Fluctuation of water levels in observation wells are illustrated by the hydrographs in Figure 2.4-18, and discussed in Section 2.4.13.

The emergency spray pond, excavated partly in soil and partly in rock, was lined to preclude excess seepage. The pond will not significantly affect groundwater levels beneath the plant (refer to Section 2.5.4.6 for a discussion of groundwater conditions beneath the spray pond).

2.5.2 VIBRATORY GROUND MOTION This section presents the evaluation of the geologic and seismic conditions in the site region as they affect design for vibratory ground motion. The investigation undertaken for the PSAR includes the following:

a. A study of geologic structure and tectonic history of the site region

b. A review of seismicity of the region, primarily based on a literature search and supplemented by a review of contemporary newspaper accounts

c. An evaluation of seismicity of the region that considers the relationship of historic earthquakes to geologic and tectonic features

d. Field geophysical measurements to determine physical properties of the bedrock at the site CHAPTER 02 2.5-21 REV. 13, SEPTEMBER 2006

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e. Selection of appropriate OBE and SSE

f. Estimation of maximum level of ground motion to be expected at the site due to the occurrence of the OBE and SSE

g. Characterization of seismic design criteria in the form of response spectra and recommended time histories Detailed descriptions and the results of the geologic and geophysical investigations that provide background information for some of the results discussed in this section are presented in Sections 2.5.1, 2.5.3, and 2.5.4. Information gained after the submittal of the PSAR has been considered and added, and discussion of this information is provided where appropriate in this report.

2.5.2.1 Seismicity The site is located in a region that has experienced a moderate amount of historic earthquake activity. The record of earthquake occurrence in southeastern Pennsylvania and the surrounding area dates back to the early 1700s. Since this region has had a relatively large and well distributed population since this time, it is probable that any major earthquake activity (defined in this context as intensity greater than or equal to VIII on the Modified Mercalli Scale, Table 2.5-1), would have been reported in local newspapers, private journals, or diaries. No evidence in documentation of this type has been found.

As indicated in the PSAR, zones of major earthquakes in the eastern United States, such as the St.

Lawrence River region and the New Madrid, Missouri region, are too far removed from the site, both tectonically and geographically, to have an appreciable effect on the seismic evaluation.

Maximum historic intensities at the site from earthquakes in these two active regions are estimated to be about IV or less on the Modified Mercalli Scale. Earthquakes near Charleston, South Carolina in 1886 are the only major earthquakes recorded in the Coastal Plain physiographic province of the eastern United States. These shocks, which had a maximum intensity of about IX, were centered about 550 miles SSW of the site. It is estimated that these shocks were felt in the site area, with an intensity of about III. The more modest local and regional seismicity of the site area is more important to the seismic evaluation of the site area than effects from events in the distant zones of major earthquakes; thus, only seismicity within 200 miles of the site is considered in the discussion below. For the purposes of this discussion the site region is defined as that area within a 200 mile radius.

2.5.2.1.1 Regional Seismicity The epicenters of earthquakes of maximum intensity IV-V or greater within approximately 200 miles of the site are shown in Figure 2.5-14 and listed in Table 2.5-2. This data set covers the interval from the first historical account in the beginning of the 18th century through January 1982.

A total of 131 events are shown and listed. Sixteen of these earthquakes (3 of intensity VI, 10 of intensity V, and 3 of intensity VI-V) occurred after 1967 and thus are an update of the PSAR compilation. A number of additional small pre-1968 earthquakes were found (2 of intensity V-VI, 9 of intensity V, and 18 of intensity IV-V), predominantly from two recent catalogs by Winkler (References 2.5-53 and 2.5-54). Data on post-1967 events have been taken from various publications of the USGS (References 2.5-19, 2.5-20, 2.5-21, and 2.5-55), Bulletins of the Northeastern U.S. Seismic Network (Reference 2.5-56), Bulletins of the Southeastern U.S. Seismic Network (Reference 2.5-57), and supplementary data from a National Geophysical and Solar CHAPTER 02 2.5-22 REV. 13, SEPTEMBER 2006

LGS UFSAR Terrestrial Data Center earthquake list (Reference 2.5-58). The new earthquakes do not alter the seismic characterization of the site region implied by pre-1968 seismic history.

The parameters of the Moodus, Connecticut, event of May 16, 1791 have been altered from the data in the PSAR. Although two events are listed in the PSAR, one on May 16th and another on May 18th, no single account describes two large events, two days apart. Rather, two very similar descriptions of a single earthquake, quoted as occurring on these two different dates, are the source of the two events in the PSAR. It is probably confusion over the date of occurrence of this shock that has led to duplication in the original listing. More importantly, the maximum intensity at East Haddam, Connecticut, is anomalous with respect to reports from nearby localities. This observation led Linehan (Reference 2.5-22) to reevaluate the intensity of this earthquake, which he assessed as V-VI. Seismologists from the NOAA, U.S. Department of Commerce, have recommended that the 1791 earthquake be reclassified as intensity VII. This value is conservatively adopted in this report.

Although no modification of the maximum intensity associated with the 1929 Attica, New York earthquake is exercised here, it may be noted in passing that, in a recent study of this event (Reference 2.5-23), a revised intensity of VII has been suggested.

As Table 2.5-2 shows, 59 earthquakes with maximum intensities of IV-V or greater have been reported within 100 miles of the site since the first historical account, at the beginning of the 18th century. Of these shocks, 5 were of intensity VII, 10 were of intensity VI, 40 were of intensity IV-V or V, and 4 were of uncertain intensity. Of the intensity VII earthquakes, the closest to the site was a shock near Wilmington, Delaware in 1871, approximately 30-40 miles from the site. Smaller earthquakes have occurred closer to the site. A series of shocks occurred in the Reading-Sinking Spring area in 1954 and 1955, about 20-25 miles to the west of the site. The largest of these had a maximum intensity of VI. Such smaller shocks are discussed in detail in the section on local seismicity below.

A discussion of the most significant earthquakes in the region follows:

a. Wilmington earthquake of October 9, 1871:

This shock is the largest earthquake originating in or near the Piedmont, and close enough to the site to be of significance in the current report. Little information about this earthquake is available, and it is therefore difficult to accurately locate its epicenter or estimate its maximum intensity. It is believed that the epicenter was located somewhat to the south of Wilmington, Delaware, where the shock was felt with intensity VII. The shock was felt from Chester, Pennsylvania, to the north, to Middletown, Delaware, to the south; and from Salem, New Jersey, to the east, to Oxford, Pennsylvania, to the west. The initial shock was followed by a much smaller aftershock just after midnight on October 10th. A contemporary newspaper account indicates that the shock was felt at Wilmington, "with great distinctness."

Buildings were shaken severely, and a number of chimneys in various parts of the city were shaken down. Windows and glass were broken. The shock also was felt strongly and resulted in broken chimneys in the surrounding towns of Oxford, Pennsylvania, and New Castle and Newport, Delaware.

An interesting aspect of this earthquake is the fact that it was accompanied by a very loud sound, as of an explosion. This loud noise, in fact, led to the belief that CHAPTER 02 2.5-23 REV. 13, SEPTEMBER 2006

LGS UFSAR the shock was caused by an explosion, probably at the powder mill of the E.I.

Dupont deNemours Company, near Wilmington. This possibility was carefully investigated at the time, however, and it was concluded that the shock was due to an earthquake.

b. Wilkes-Barre earthquake of February 21, 1954:

This earthquake and a strong aftershock two days later were probably not of tectonic origin. The shock occurred in an area in Wilkes-Barre, Pennsylvania, which is underlain by coal mines extending to a depth of 400 feet below the ground surface. Wilkes-Barre is about 65 miles north of the site. The affected area was confined to the east side of the Susquehanna River, in a five-block residential area.

The effects of this disturbance were locally severe. Occupants fled to the streets.

Sidewalks pushed upward with a heaving motion and then collapsed. Hundreds of homes were damaged, ceilings and cellar walls split, and backyard fences were pushed over. Gas and water mains snapped. Methane gas rising from cracks in the earth presented a temporary emergency. The estimated property damage was

$1,000,000; however, this figure may be somewhat high.

The second shock, two days later, occurred in the same area. Again, hundreds fled to the streets. Cracks appeared in ceiling, and walls of apartment buildings. Curbs pulled away from sidewalks; street pavements buckled from curb to curb.

Additional water and gas mains were broken.

The small affected area of these shocks suggests a very shallow focus and a source mechanism of limited lateral extent. Detailed investigations of the event shortly after it occurred lead to these conclusions: it was closely related to slip along a zone of weakness that exists in the strata (in which coal mining was taking place) along the south side of the affected area, and, whatever the cause-and-effect relationship was between this slip and roughly concurrent failure of coal pillars and partial mine collapse, this event would not have occurred without the previous existence of the coal mine excavations.

c. Earthquake of March 8, 1889:

This shock had its epicenter in southeastern Pennsylvania, probably near York, about 60 miles southwest of the site, where the shock was felt most strongly.

The maximum intensity of the shock was VI. The shock was felt throughout northeastern Maryland, northern Delaware, and southeastern Pennsylvania. The area in which it was felt was roughly elliptical in a NE-SW direction from Trenton, New Jersey, to Hagerstown, Maryland. The shock lasted about ten seconds.

No serious damage was reported. The greatest damage occurred at York, where some chimneys were knocked over and people were badly frightened. A man was thrown from a sofa, and articles were thrown from shelves. The shock was accompanied by a loud rumbling noise. It is interesting to note that there were reports of a "ball of fire" passing over the area at the same time as the shock.

CHAPTER 02 2.5-24 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5.2.1.2 Local Seismicity Some minor earthquake activity has occurred in the site vicinity. Seventy-two events within approximately 50 miles of the site are listed in Table 2.5-8 and plotted in Figure 2.5-15. Forty-five of these events have been of less than intensity IV-V (or unassigned) and therefore do not appear in the regional catalog above. Data on the additional events have been taken from various publications of the USGS (References 2.5-19, 2.5-20, 2.5-21, and 2.5-55), Bulletins of the Northeastern U.S. Seismic Network (Reference 2.5-56), Bulletins of the Southeastern U.S. Seismic Network (Reference 2.5-57), Winkler (References 2.5-53 and 2.5-54), and supplementary data from a National Geophysical and Solar Terrestrial Data Center earthquake list (Reference 2.5-58).

The completeness and accuracy of earthquake reporting in the site area is not uniform over the entire period of record. Before instrumental epicenters were reported in the northeastern and southeastern U.S. Seismic Network Bulletins, most earthquake locations were based on sensory and damage reports from area surveys following the earthquake, from newspaper articles, historical journals and publications, and private diaries. No formal location accuracy estimate is attempted for earthquakes occurring before 1975. However, it is believed reasonable to assume that these events are generally located to an accuracy of better than ten to several tens of kilometers. The instrumental locations of earthquakes occurring after 1975 are generally better constrained. The degree of uncertainty in the locations of the small site area events that have occurred after 1975 is about one-half the degree of uncertainty associated with pre-1975 event locations. In addition, the disproportionate number of post-1975 earthquakes indicates an increased earthquake detection capability in the site area. Location uncertainties, and the nonuniformity of location accuracy and event detection capability over the total time period of record, should be recognized in considering the seismicity of the vents listed below and of those shown in Table 2.5-8 and Figure 2.5-15.

Descriptions of the historical seismicity for the Reading-Sinking Spring, Allentown and Philadelphia areas follow. A "detected" event as used in this discussion indicates that the event was felt, heard, or noticed in some way.

a. Reading-Sinking Spring Area Earthquakes: (about 20-25 miles west of the site)

1. May 28 and 29, 1906 - A small shock of reported intensity III was felt on May 28 over an area within about 2-3 miles of Geigerstown. On May 29 a water tank collapsed in New York City, tentatively associated with a IV-V shock in Reading. The occurrence of this latter event is suspect.

2. June 8, 1937 - This shock was very slight, with a possible maximum intensity of III. The epicenter was in the vicinity of Reading. Little information is available.

3. January 7, 1954 - This shock and a series of aftershocks originated in the Sinking Spring area west of Reading. The shock had a maximum intensity of VI. Damage in Sinking Spring consisted of plaster ripped from walls, dishes and bottles tumbled from tables, upset furniture, and slight damage to brick and frame buildings. People were thrown out of bed. In the surrounding communities, many people felt the tremor, homes rocked, furniture toppled, and dishes and lamps rattled. The main shock was followed by many minor aftershocks over a period of a few weeks.

CHAPTER 02 2.5-25 REV. 13, SEPTEMBER 2006

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4. January 23, 1954 - An aftershock of maximum intensity III shook a 10 mile square area surrounding Sinking Spring.

5. August 10, 1954 - About 12 tremors, four of "good size," maximum intensity IV, were felt in the Sinking Spring area within a 10 minute period.

6. September 24, 1954 - A series of minor tremors, maximum intensity III, were felt in the Sinking Spring area.

7. January 19, 1955 - A shock of maximum intensity IV occurred in the Sinking Spring Area. The shock was felt from West Reading west to Wernersville; houses vibrated, windows and dishes rattled, and a lamp tumbled to the floor. A noise accompanied the shock. Many felt the shock and were alarmed.

b. Allentown Area Earthquakes:

1. May 31, 1884 - This shock had a maximum intensity of about V. The shock was felt in a small area at Allentown. No damage was reported other than dishes thrown from tables.

2. May 31, 1908 - This shock had a maximum intensity of VI. This shock was felt in all parts of Allentown, in a total area of about 50 square miles, but not in the surrounding communities. The shock lasted about one or two seconds and was accompanied by a low rumbling sound followed by a loud report, as of an explosion. Many people were frightened by the shock.

Some were thrown to the ground. Damage consisted of a few fallen chimneys, broken windows, dishes, etc, and other minor building damage.

3. November 23, 1951 - An extremely local shock was reported in the northwest section of Allentown. The maximum intensity of the shock was IV.

Many were awakened. It was reported that a bed was moved 6-7 inches. A vase was knocked off a shelf and a pump was put out of order. Reports indicated that the water level in a well used for 25-30 years dropped 60 feet.

The shock was accompanied by a low rumbling noise.

4. September 14, 1961 - This shock was centered in the Lehigh Valley, about 5 miles east of Allentown. The maximum intensity of the shock was V. It was felt in a larger area than the foregoing earthquakes; however, no damage was reported. The earthquake was strongest in Allentown, Bethlehem, and surrounding communities where the shock was felt by nearly all and where many were alarmed and awakened. Many were shaken from their beds. Loose bricks fell from a chimney. Buildings shook, loose objects rattled, and thunderous sounds were heard.

c. Philadelphia Area Earthquakes:

1. November 26, 1755 - Detected at Philadelphia.

2. November 22, 1963 - Detected at Delaware County, Pennsylvania.

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3. October 13, 1763 - Detected at Philadelphia.

4. October 30, 1763 - Intensity IV-V at New Jersey; IV at Philadelphia; detected at Bensalem, Pennsylvania.

5. November 22, 1777 - Detected at Philadelphia.

6. November 23, 1777 - Detected at Delaware County, Pennsylvania.

7. November 29 and 30, 1783 - For November 29 event at 22:15: intensity IV-V at Philadelphia; IV at New York City and New Haven, Conn.; detected at Hartford, Conn., Boston, Mass., N.H., N.J., and R.I.; shocks also felt at 21:00 and 01:00-02:00 the next morning (November 30) at Philadelphia and New York City.

8. March 17, 1799 - Detected at Philadelphia.

9. March 17, 1800 - Severely felt at Philadelphia.

10. November 29, 1800 - Severely felt at Philadelphia.

11. November 12, 1801 - Detected at Philadelphia.

12. December 8, 1811 - Intensity III at parts of Pennsylvania and Delaware (including Wilmington, DE).

13. November 11 and 14, 1840 - Both had intensity IV at Philadelphia; great and unusual swell on the Delaware River was noted.

14. October 12, 1870 - Intensity III near Wilmington, Delaware.

15. October 9, 1871 - Intensity VII at Wilmington, Delaware. A press report stated chimneys toppled and windows broke. Damage also was reported at New Castle and Oxford. Event was felt in New Jersey (IV-V at Haddonfield; IV at Salem) and Pennsylvania (IV at Philadelphia).

16. August 10, 1877 - Intensity III near Trenton, N.J.

17. September 10, 1877 - Intensity IV-V in the Delaware Valley. Felt from Trenton, N.J. to Philadelphia, over an area 20 miles wide with center near Burlington, N.J.

18. March 25, 1879 - Intensity IV-V in the Delaware Valley, near Dover, Delaware and below Philadelphia. Felt from Chester, Pennsylvania to Salem, N.J., a distance of 30 miles. Felt most strongly on the east side of the Delaware River.

19. October 10, 1892 - Intensity III near Wilmington, Delaware.

20. November 20, 1895 - Intensity IV at Claymont, Delaware and Chester, Pennsylvania; felt at Wilmington, Delaware and Linwood, Pennsylvania.

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21. March 17, 1900 - Intensity III near Philadelphia.

22. April 28, 1900 - Intensity III near Philadelphia. Felt at Camden, N.J.,

southern N.J., and Philadelphia.

23. November 29, 1900 - Intensity III near Philadelphia.

24. February 6, 1909 - Intensity III near Trenton, N.J.

25. January 26, 1921 - Intensity V at Moorestown, N.J. Rumbling noise reported. Felt in Philadelphia. Fissure reported in Burlington.

26. January 24, 1933 - Intensity V near Trenton, N.J. A Sharply felt shock.

Pictures thrown from wall at Lakehurst.

27. August 22, 1938 - Intensity V in central New Jersey. Felt throughout central New Jersey, southeastern Pennsylvania, and northern Delaware. Slight damage at Gloucester City and Highstown, N.J., and Ardmore, PA. Four smaller shocks occurred on the 23rd and one on the 27th.

28. November 14, 1939 - Intensity V in Salem County, N.J. Reported felt from Trenton, N.J. to Baltimore, MD., and from Cape May, N.J., to Philadelphia, PA and its adjoining counties. About 6000 square miles were affected.

Little or no damage resulted. Intensity V at Deep Water, N.J. where small objects overturned.

29. January 8, 1944 - Felt from Wilmington, Delaware to the southwestern outskirts of Philadelphia. Windows rattled and houses shook.

30. December 27, 1961 - Intensity V near Pennsylvania-New Jersey border. At Bristol and in the northeast portion and suburbs of Philadelphia, buildings shook, dishes rattled, and disturbed objects were observed by many. Police and newspaper offices were swamped with calls from alarmed citizens inquiring about the loud rumbling sounds. Felt by many at Levittown and Langhorne, Pennsylvania and Bordertown and Trenton, New Jersey, where houses shook and creaked, windows and dishes rattled. Felt by several at Burlington.

31. February 28, 1973 - Felt in all or parts of Connecticut, Delaware, Maryland, New Jersey, Pennsylvania, and Virginia. Observers reported cracked plaster at North East and Perryville, MD, and Laurel Springs and Penns Grove, N.J. At Harrisonville, N.J., cinder block basements reportedly cracked. An observer at Palmyra, N.J. reported damage (not described) and that a driveway "dropped 3 inches." At New London, PA, patio, walls, and plaster cracked. "Silent cracks where concrete meets house," loosened rain spouts and cracked plaster were observed by Norristown, PA, residents. At Thornton, PA, pipe connections to a well broke. Two mirrors fell from a wall and shattered at Wallingford, PA. Small objects shifted and fell in several towns in Delaware, New Jersey, and Pennsylvania.

Magnitude 3.8.

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32. March 2 to May 2, 1980 - Sequence of small earthquakes in the vicinity of Abington, PA. The largest events were on March 5 (magnitude 3.5) and March 11 (magnitude 3.7); both were felt with maximum intensities of IV to V. A small magnitude 3.1 aftershock 13 minutes after the March 5 event was also felt in the Abington area. Bischke (1980) reports that both of the March 5 and 11 events were felt and heard by many people as explosive or low distant rumbling sound. "In one case a plant overturned; in several cases small objects vibrated and shifted slightly on their foundations, plaster cracked and pictures tilted on the walls. In general, windows and dishes rattled, and over half of the people reporting the March 11 event at 1 in the morning were awakened by the earthquake."

33. August 30, 1980 - Event near Medford, N.J.; magnitude 3.0.

The descriptions above discuss all known seismic activity in the Reading-Sinking Spring, Allentown, and Philadelphia areas. As discussed below, none of these seismic activities are associated with any known structures within the context of 10CFR100, Appendix A.

Six events with maximum intensities ranging from II to VI have been tabulated near Reading-Sinking Spring. These occurred between 1906 and 1955 and were located only by sensory reports.

Four shocks of intensities from IV to VI are reported to have occurred in the Allentown area between 1884 and 1961. Only sensory report information constrains the locations of these events.

Estimates of location accuracy for these earthquakes indicate an uncertainty of several miles or more. In addition, small events of this type could easily occur on small dislocations in the earth's crust with no surface representation. Because of their small size and uncertain locations, none of these events are associated with known structures in the context of 10CFR100, Appendix A.

As discussed in Section 2.5.2.3, a number of earthquakes in the site region appear to parallel the NE-SW geologic structural trend in a relatively narrow belt along the axis of the Fall Zone. This belt passes through the Philadelphia area about 30 miles southeast of the site. A total of 33 events and event sequences are noted in this area. These range in size from indeterminate intensity to intensity VII and occurred from 1755 to 1980. For all events before 1973, only sensory report locations are available. For the sequences beginning February 28, 1973 and March 2, 1980, instrumental and intensity report data exist. For a single earthquake on August 30, 1980, only instrumental data was found.

The Philadelphia area earthquakes occurring before the availability of instrumental data are, as with the Allentown and Reading-Sinking Spring events, too poorly located and too small to suggest association with specific known structures in the context of 10CFR100, Appendix A. A similar conclusion is reached for the August 30, 1980 event.

Sbar et al. (Reference 2.5-148) investigated the February 28, 1973 earthquake sequence and speculated that this sequence may have occurred on a northeast continuation of a graben described by Spoljaric (Reference 2.5-149). This continuation coincides with the Fall Zone in the Wilmington, Delaware area. Sbar et al. noted that more accurate locations of earthquakes and extensive mapping of faults are necessary to verify their hypothesis.

CHAPTER 02 2.5-29 REV. 13, SEPTEMBER 2006

LGS UFSAR Bischke (Reference 2.5-146) investigated the association of the March 2, 1980 sequence with the Huntingdon Valley fault and concluded that the fault should be considered active on the basis of meizoseismal epicenters for several earthquakes of this sequence. Instrumental epicenters lie from several to over ten miles from the Huntingdon Valley fault. Bischke attributes these locations to inadequate instrumental coverage and possible errors in local crustal model velocities. Bischke concludes further that additional study of this fault is warranted.

Although both the 1973 and 1980 sequences have been speculated to occur on specific structures near the Fall Zone in the Philadelphia area, all events of these sequences have been small (less than magnitude 4.0), therefore they are not associated with large source structures. Both sequences have had disparate sensory and instrumental epicenter locations indicating some real uncertainty in the epicenters of these events. Each sequence has been an isolated instance of a cluster of events near a known or postulate structure. For these reasons; it is believed that these earthquakes should not be associated with known structures in the context of 10CFR100, Appendix A. Rather they are seen as part of the historic series of earthquakes near the Fall Zone in the Philadelphia area as noted above.

As discussed below, the April 1982 and March 1980 earthquakes are not correlated with the same structure.

A magnitude of 2.9 earthquake occurred on April 12, 1982 near Burlington, New Jersey. The epicenter was at 403.5' N, 7448.9' W, according to an instrumental location by Lamont-Doherty Geological Observatory. Bischke (Reference 2.5-147) located this event about 6 miles to the west of the instrumental location based on sensory reports. The earthquake was both felt and heard, and many people detected the trembling motion for 5-10 seconds. An intensity of III to IV has been proposed for this event, although intensity V effects were felt locally (Reference 2.5-147).

Bischke (Reference 2.5-147) associates the April 12, 1982 earthquake with the Fall Zone beneath the Delaware River and attributes the failure of the instrumental epicenters to agree with the intensity data to instrumental location errors. This event should not be associated with a particular structure in the context of 10CFR100, Appendix A. Because the March 2, 1980 sequence and the April 12, 1982 event are associated with separate features, it is concluded that they are not correlated.

2.5.2.2 Geologic Structures and Tectonic Activity 2.5.2.2.1 Regional Geologic Setting The site is located in the Triassic Lowland section of the Piedmont physiographic province of the Appalachian Highlands. The Piedmont is bounded on the southeast by the Atlantic Coastal Plain physiographic province. The Fall Zone, which trends NE-SW approaching within about 30 miles southeast of the site, marks the topographic boundary between the Piedmont and Coastal Plain provinces.

Bedrock in the Piedmont is generally at shallow depth, covered only by a thin veneer of residual soil. In some places, bedrock is exposed. Crystalline Precambrian rocks outcrop to the north and northeast of the site (the Reading Prong) and in a narrow belt parallel to the Fall Zone southeast of the site. The site is located in a down-warped or down-faulted Triassic basin, the Newark Basin, which contains sedimentary rocks, predominantly sandstones and shales. The geologic history of the Piedmont is complex. Major tectonic activity has occurred in the ancient part and many zones of faulting have been identified. Many intrusions are known; for example, the rocks of the Newark CHAPTER 02 2.5-30 REV. 13, SEPTEMBER 2006

LGS UFSAR Basin are crossed by numerous dikes and sills, which in some instances have intruded pre-existing faults.

Coastward, the bedrock surface is down-warped to the south and east, with the line of flexure generally within the Fall Zone. In the Coastal Plain, the bedrock is covered by unconsolidated sediments consisting of interbedded silt, sand, and gravel. These sediments form a wedge-shaped mass that thickens toward the southeast. These sediments range in age from Cretaceous to Recent. The down-warping of the bedrock surface of the Coastal Plain has occurred under the weight of accumulating sediments.

There are indications that some down-warp of the coastal plain relative to the Appalachian chain may be occurring today, although no significant crustal warping during Pliocene and Quaternary time is evident in the record of offshore sedimentation (References 2.5-24 and 2.5-25).

North of the site, the Triassic Lowland rocks abut Precambrian crystalline rocks of the Reading Prong, a narrow highland belt separating the Piedmont rocks from the intensely folded Lower Paleozoic rocks of the Great Valley sequence farther north. Westward, Triassic rocks of the Gettysburg Basin lie directly upon recumbent folds and nappes of the Valley and Ridge province.

2.5.2.2.2 Regional Tectonic Provinces The site lies in a Triassic Lowland section (Newark Basin) of the Appalachian Piedmont tectonic province (Figure 2.5-16). This part of the Appalachian Piedmont in Pennsylvania, New Jersey, and Maryland is typified by the presence of several triassic basins (Culpeper, Gettysburg, and Newark) that have been postorogenically down-warped in the contorted metamorphic and igneous rocks exposed elsewhere in the Piedmont (Figure 2.5-16). Southwest of Reading, Triassic basin sediments of the Piedmont province rest directly on tightly folded and overturned beds of the Northern Valley and Ridge province.

A few miles north of the site, a belt of metamorphosed Precambrian (Grenville) crystalline rock, capped by patches of resistant basal Cambrian quartzite and dolomite, intervenes between the Triassic basin sediments and the slightly metamorphosed, but tightly folded, strata of the Great Valley section of the Northern Valley and Ridge province, here occupied by the Lehigh Valley. This intervening strip of Precambrian rock is the Reading Prong, the southwestern most extension of a distinctive belt of Grenville rocks that forms a series of highlands that can be traced northeastward into western Vermont. Traditionally, the Reading Prong has been viewed as a physiographic and lithologic extension of the crystalline rocks of the New England Appalachians.

Precambrian rocks also occur along South Mountain, southwest of Harrisburg. Here they form the northeastern arm of the Blue Ridge tectonic province, an anticlinorium of late Precambrian metavolcanics (greenstones) and associated Grenville age rocks, enclosing the southwestern flank of the Gettysburg Basin. Grenville rocks may possibly persist in a continuous zone from the Blue Ridge province to the Reading Prong-Hudson Highlands-Green Mountain belt, but if so, the relationships are obscured beneath the Triassic sediments of the Newark and Gettysburg Basins.

Proceeding northwestward from the site, across the folds of the Northern Valley and Ridge province, the folds are at first appressed and complexly faulted in the Great Valley belt but become more open farther northwest, where they assume their classic cylindrical forms. Beyond the Allegheny front (Figure 2.5-16), the folding at once becomes open and subdued, with the strata descending into a broad, gently folded basin or synclinorium. Classically this region of broad, gentle folding is part of the Appalachian Plateaus, though truly undeformed Paleozoic strata lie as much as 100 miles further beyond the Allegheny front.

CHAPTER 02 2.5-31 REV. 13, SEPTEMBER 2006

LGS UFSAR Southeast of the site, the wedge overlap of Coastal Plain strata on the Piedmont crystalline rocks is about 30 miles distant; beyond, undeformed Lower Cretaceous through Recent sediments of the Atlantic Coastal Plain province thicken uniformly seaward.

The New England part of the Appalachian orogenic belt is an extensive terrain composed predominantly of metamorphic rocks, with lesser amounts of igneous and sedimentary rocks included. Like the rest of the Appalachian orogen, the New England Appalachians (Figure 2.5-16) are characterized by elongated structural systems and tectonic belts; however, the dominant orientation of these features, at least in the southern and western parts of the New England Appalachians, is more nearly north-south than NE-SW, as it is in most of the Appalachians.

Recent application of plate-tectonic concepts to New England geology has led to a recognition of several fundamentally distinct crustal units within the New England Appalachians (References 2.5-26, 2.5-27, and 2.5-28). Several investigators have used findings to subdivide the New England Appalachians into tectonic provinces (for example, see Reference 2.5-29. The provinces are not discussed in this report since, for the reasons explained in Section 2.5.2.3, the detailed subdivision of the New England Appalachians is not significant to the LGS site.

2.5.2.2.3 Regional Tectonic Structures Faulting of Precambrian age undoubtedly exists in the Piedmont, but the rocks of the Piedmont have been so metamorphosed since that time that no significant Precambrian faults are recognizable. Paleozoic faults generally are characterized by thrusting toward the northwest, a reflection of compressional forces imposed during the late Paleozoic (Alleghenian) orogeny, as well as during earlier tectonic events.

These thrusts are especially well displayed in the Northern Valley and Ridge province, just northwest of the Blue Ridge-Reading Prong anticlinoria, where the compressive forces in the sedimentary pile were at a maximum. In the Valley and Ridge provinces, thrusting is believed not to have involved basement rocks; rather, the foreshortening necessitated by the great Valley and Ridge folds is thought to have been taken up by buried ramp thrusts riding over incompetent shale lying on top of the basement rocks (References 2.5-13 and 2.5-30). Highly complex fault patterns are also displayed in the Precambrian rocks of the Blue Ridge and Reading Prong anticlinoria; their appearance attests to the fact that basement rocks were very much involved with thrusting southeast of the Great Valley. Southeastward, in the Appalachian Piedmont, the crystalline rocks have been subjected to multiple deformational and metamorphic episodes the Paleozoic, and many zones of pre-Triassic faulting have been identified.

Recent studies using COCORP (Consortium for Continental Reflection Profiling) seismic reflection data suggest that the entire Blue Ridge and at least part of the Piedmont of the southern and central Appalachians are allochthonous, underlain at several kilometers depth by a regional decollement which was active during the Alleghenian orogeny. This same decollement would extend westward beneath the Valley and Ridge province, where upward-splaying ramp thrusts, listric into the decollement, absorbed the westward translation and crustal foreshortening achieved along the decollement, thereby producing the well-known thrust fault and fold pattern of the Valley and Ridge province.

The closest major mapped fault of Paleozoic age is the Brandywine Manor fault, which approaches within about 9 miles to the southwest of the site. The Huntington Valley-Cream Valley fault (and the related Rosemont fault) approach within about 15 miles to the south of the site. None of the major Paleozoic faults has a mapped extension eastward into the Triassic Newark Basin. These CHAPTER 02 2.5-32 REV. 13, SEPTEMBER 2006

LGS UFSAR faults have been completely healed and are inactive. Some major thrust faults are transected by subsequently intruded Triassic-age dikes that show no displacement.

Faults of Triassic-age occur within the Newark Basin (Figures 2.5-2 and 2.5-38). The most significant of these to the east of the site. The Chalfont fault is the closest of these, approaching within about 9 miles to the northeast of the site. Some minor Triassic faults exist closer to the site (Figure 2.5-6). Some of these faults have developed along the major joint systems in the Triassic rocks. In some instances, Triassic-age dikes have intruded along some of these fracture zones, thus indicating their lack of activity since the Triassic (more than 140 million years ago). These minor faults are discussed in Section 2.5.1.2.3.

Faulting younger than Jura-Triassic is reported in a few places in the Coastal Plain. Reported displacements are steeply dipping; reverse motion on northeast-trending planes has occurred in some instances. All such displacements are minor compared to older faulting. Within 200 miles of the site, these faults include the Brandywine fault (140 miles southwest of the site), the Stafford fault system (150 miles southwest of the site), and the New York Bight fault (off the New Jersey coast, about 100 miles east of the site). On none of these is there unequivocal evidence for latest movement younger than Miocene, or possibly no younger than Paleocene or early Eocene (Section 2.5.1.1.3).

A postulated major east-west fault zone, named the Cornwall-Kelvin wrench fault, has been suggested to extend through the Triassic Lowland of southeastern Pennsylvania. The postulation of this fault zone is based on subsea topographic and geophysical surveys. However, there are neither geological nor geophysical bases for projecting this postulated fault zone into the continent.

Two small faults occur in the Newark Basin near the site: the Sanatoga fault, in the northwest portion of the site; and the Brooke-Evans fault, in the southeast portion of the site. The maximum displacement along these faults is about 350 feet. The Sanatoga fault has been intruded by the Triassic-age Downingtown Dike. These faults were investigated by Dames and Moore (Reference 2.5-1) and evaluated by an independent review committee consisting of experts in Appalachian geology. The review committee concluded that these faults are closely related to the well-dated Jura-Triassic dikes injected along them.

Geologic structure is discussed further in Sections 2.5.1.1.3 and 2.5.3.

2.5.2.3 Correlation of Earthquake Activity with Geologic Structures or Tectonic Provinces As shown in Figure 2.5-14, the trend of epicenters in the site region is generally NE-SW, parallel to the trend of geologic structure in the Appalachian Mountains. As a particular example, within 100 miles of the site there appears to be a pattern of seismicity paralleling geologic structure in a relatively narrow belt, roughly along the axis of the Fall Zone. At distances greater than 100 miles northeast of the site, the seismicity becomes more diffuse and scattered throughout that part of the New England Appalachians within 200 miles of the site. To the southwest of the site, at distances greater than 100 miles, the seismicity is again more diffuse and, moreover, there are fewer events until the Virginia Piedmont is reached just beyond the 200 mile radius. A modest clustering of historic earthquakes occurs here, in the general area of Richmond, Virginia. Finally, both to the southeast and northwest of the Fall Zone, there is scattered earthquake activity within the Coastal Plain and Appalachian physiographic provinces, respectively.

The earthquake activity in northeastern Massachusetts, although well in excess of 200 miles from the site, has been considered. In November 1755, an intensity VIII earthquake occurred offshore CHAPTER 02 2.5-33 REV. 13, SEPTEMBER 2006

LGS UFSAR near Cape Ann, at least 300 miles from the site. This is the largest historic earthquake known to have occurred in the New England region. It has been associated variously with the Southeastern or Avalon Platform tectonic province, a Cape Ann-New Hampshire tectonic province (each more than 200 miles from the site) (Reference 2.5-29), an east-west zone of intense thrust and strike-slip faulting (Reference 2.5-31), a northwest trending seismic and plutonic zone (Reference 2.5-32),

and a specific plutonic structure (Reference 2.5-33). None of the above alternatives would relate the event to provinces or structures that are closer than 200 miles from the site; therefore, the 1755 Cape Ann event is not significant to the site.

2.5.2.3.1 Tectonic Models Relating Seismicity to Geologic Structure A number of widely divergent hypotheses have recently been advanced relating seismicity in the Appalachian region to specific geologic structures. Direct field evidence, such as surface faulting or fault-generated topographic features, has not been found. This wide divergence of opinion results because of the limited data base available and because the hypotheses are constructed largely from permissive rather than compelling evidence. These hypotheses have been summarized and evaluated previously by others (Reference 2.5-79).

The hypotheses that have been advanced can be grouped into one of four general mechanisms:

(1) reactivation of steep basement faults, (2) decollement reactivation, (3) stress amplification at the margins of mafic and ultramafic intrusions, and (4) subsidence of the continental margin. The following discussion reviews and assesses these hypotheses.

2.5.2.3.1.1 Reactivation of Steep Basement Faults Steeply dipping, northeast-trending faults exhibiting Cretaceous and Cenozic displacement are present in the Atlantic Coastal Plain. Several authors suggest that the faults are a reactivation of former basement structures and relate historic seismicity to recurrent movement on these faults.

Cretaceous and Cenozoic displacement is indicated from geologic and seismic reflection data for a number of faults, including the Stafford fault zone (References 2.5-110, 2.5-106, and 2.5-107), the Brandywine fault zone (Reference 2.5-98), and the New York Bight fault (Reference 2.5-96) within 200 miles of the site; and south of the 200 mile radius, the Belair fault zone (in Georgia)

(References 2.5-114, 2.5-111, and 2.5-115) and the Cooke and Helena Banks faults (onshore and offshore South Carolina, respectively) (References 2.5-68 and 2.5-69). The faults displace strata as young as Eocene or Miocene in age and, in possibly two cases (New York Bight and Stafford faults), as young as Pliocene or early Pleistocene in age. (The interpretation of Plio-Pleistocene movement on the Stafford fault zone discussed in Reference 2.5-107 seems questionable in view of several trenches across the same fault which show no such offset (Section 2.5.1.1.3).)

Wentworth and Mergner-Keefer assume these faults have experienced recurrent movement, based on the indication of smaller displacements on them in younger strata.

Zoback and Zoback (Reference 2.5-140) suggest that if these faults are zones of weakness, they may localize stress and cause brittle failure (earthquakes) in the upper crust. Wentworth and Mergner-Keefer (References 2.5-134 and 2.5-135) further suggest that a domain of northeast-trending, specifically reverse faults exists along the east coast, in keeping with the inference by Zoback and Zoback of persistent NW-SE regional compression. They point out that northeast-trending reverse faults in the basement must have been reactivated by compressive stresses.

They propose that the reactivated basement structures are most likely to be normal faults associated with Triassic and possibly Precambrian rifting. (It should be noted, however, that evidence for reverse displacement on the Helena Banks fault is quite tenuous, and there is no CHAPTER 02 2.5-34 REV. 13, SEPTEMBER 2006

LGS UFSAR indication that the New York Bight fault is reverse. Furthermore, there is no convincing proof of post-Miocene offset on any of the investigated reverse faults.) Mergner-Keefer (References 2.5-134 and 2.5-135) propose that historic seismicity in the eastern U.S., including the 1886 Charleston event, is related to continued reactivation of these reverse faults but that long periods of quiescence and small cumulative offsets make it difficult to recognize surface rupture and to document historic fault activity.

In support of their reactivation hypothesis, Wentworth and Mergner-Keefer infer the existence of a persistent, northwest directed regional compressive stress-field by reference to focal mechanism solutions, and to orientation and sense of displacement of Cretaceous and Cenozoic faults. In particular, they refer to focal mechanism solutions in the Giles County, Virginia seismogenic zone (Reference 2.5-70) and the Newark Basin area (References 2.5-63 and 2.5-119), and selected solutions from the Charleston, South Carolina area (Reference 2.5-131), and conclude that reverse faulting on northeast-trending high angle or subhorizontal faults is the predominant mode of contemporary fault movement.

However, the evidence is not fully supportive of such a regional stress regime. As Stephenson and Pratt (Reference 2.5-127) and Talwani (Reference 2.5-129) note, fault-plane solutions in the southeast imply considerable local variations in stress; for example, the data from Monticello Reservoir in central South Carolina and from Lake Jocassee in northwest South Carolina imply northeast directed compression, as does hydrofacturing and overcoring data at Bad Creek (Reference 2.5-130). At Charleston itself, focal mechanism solutions strongly favor northwest-oriented fault-planes (northeast compression). Tarr et al (Reference 2.5-131) consider that all except three of 16 solutions define relatively well constrained, northwest-oriented faulting, whereas but two solutions, defining a nearly vertical plane loosely constrained between a northwest and an E-NE strike, merely permit inference of northeast-trending faulting. Tarr et al (Reference 2.5-131, p. 1895) in this latter case select the northeast, rather than northwest trend on the basis of evidence (geophysical) unrelated to focal mechanism solutions. The favored indication of northwest-trending structure is corroborated by recent detailed focal plane studies by Talwani (Reference 2.5-129) which imply shallow northwest-oriented, steeply dipping reverse faulting and deep-seated, northeast-oriented strike-slip (but not reverse) faulting.

In summary, the hypothesis of Wentworth and Mergner-Keefer that contemporaneous northwest directed crustal compression causes large eastern U.S. earthquakes by reverse reactivation along steep basement faults has a number of inconsistencies. Among these are:

a. Even with a long recurrence interval of 500,000 to 1 million years and a displacement rate of 0.35-0.5 m per million years on the reactivated faults (Reference 2.5-134), many late Cenozoic displacements should be evident.

However, there is little convincing evidence for post-Miocene offset on any of the recognized Cenozoic reverse faults. Consequently, there is little basis for extrapolating continued movement from Miocene to the present time on reverse faults presumed to be present throughout the eastern U.S. Therefore, there is little justification for associating present day seismicity with these faults.

b. Implicit in the model is regional NW-SE directed compression. However, several authors describe either random and weak regional compression (Reference 2.5-83) or NE-SW directed compression (References 2.5-127, 2.5-129, and 2.5-132).

c. To the extent that NW-SE compression in the central and southern Appalachians is postulated to exist based on the premise of indicated northeast-trending CHAPTER 02 2.5-35 REV. 13, SEPTEMBER 2006

LGS UFSAR Cretaceous and Cenozoic reverse faults (Reference 2.5-140), this inferred compressive stress direction cannot then serve as evidence to justify postulating the existence of additional, as yet undiscovered reverse faulting (References 2.5-134 and 2.5-135).

d. NW-SE regional compression, required by the model, is not consistent with subsidence known to be occurring in regions along the Atlantic Coast. Subsidence implies a tensional horizontal stress regime with associated normal faults.

e. Composite focal mechanism solutions by Talwani (Reference 2.5-129) suggest that northeast-trending reverse faults are not active in the Charleston area. Instead two clusters of seismicity are present: a shallow cluster (4-8 km) which defines a northwest-trending reverse fault, and a deep cluster (9-13 km) which defines a northeast-trending strike-slip fault. Either the Charleston area is unique or the reactivation model is not regionally applicable. Furthermore, composite fault-plane solutions by Tarr et al (Reference 2.5-131) show two relatively well constrained vertical nodal plane striking northwest. Even if Wentworth and Mergner-Keefer's hypothesis were correct, it does not appear to be applicable to the Charleston area.

f. The present day Charleston seismogenic zone trends northwest (Reference 2.5-131), rather than northeast, and composite focal mechanism solutions suggest that the seismic activity is associated with a northwest-trending structure (References 2.5-129 and 2.5-131). In addition, the extended meizoseismal area of the 1886 earthquake trends northwest (References 2.5-123, 2.5-65, and 2.5-124).

Thus, although seismic reflection data show structures trending both northwest and northeast, there is little evidence associating either the 1886 earthquake or recent seismicity to a northeast-trending structure.

One of the few, and certainly one of the best documented, instances of possible relation of seismic activity to tectonic structure involves the Ramapo fault in northern New Jersey (Figure 2.5-38). This activity, detected by seismic network monitoring for microearthquakes (Nuttli magnitude less than

3) appears to be concentrated near the trace of the Ramapo fault and its northern splays east of the Hudson River (Reference 2.5-119). Hypocentral locations of microearthquakes near the fault appear to fall approximately on the downward projection of mapped dips on the Ramapo and associated Paleozoic shear zones (Reference 2.5-119, figure 4). Nodal plane solutions for three microearthquakes positioned close to the fault indicate reverse-slip (Reference 2.5-137, figure 14).

In contrast, macroseismicity shows no obvious tendency to cluster near the Ramapo zone.

Historical earthquakes with epicentral locations within at least 25 kilometers of the Ramapo do not exceed intensity VI (Reference 2.5-82, figures 1 and 2). Considering the complete record, many of the microearthquakes recorded by the seismic network (including those with reverse-slip fault-plane solutions) are located at considerable distance from the Ramapo fault system. Further, there have been some earthquakes located close to the Ramapo that have fault-plane solutions indicating reverse-slip on northwest trending, southeast dipping planes (Reference 2.5-82). Thus, the record shows a diffuse scattering of historical epicenters in the region, with a clustering of some, but by no means all, of the detected small earthquakes along the Ramapo.

It is interesting to examine the geologic evidence for offset along the Ramapo. Detailed mapping of the area in the vicinity of the Ramapo, including trenching and examination of drill cores across the Ramapo fault, have revealed no evidence of reactivation along the Ramapo fault surface. In drill hole cores which penetrated the actual fault contact (cataclastic Triassic fanglomerate over CHAPTER 02 2.5-36 REV. 13, SEPTEMBER 2006

LGS UFSAR gouge over phyllonitized basement rock), the last recorded displacement is normal; no evidence of reverse reactivation is seen (Reference 2.5-119, figure 7). No instance has been discovered of offset glacially polished surfaces or of offset surficial deposits in natural exposures or in trenches across the trace of the Ramapo fault (Reference 2.5-119, p. 282). Ratcliffe (Reference 2.5-119, p.

281) concludes, "my present feeling is that we have not yet demonstrated either from experiment, theory or empirical observation that the premise restricting seismic activity to surface brittle fracture zones (e.g., Ramapo fault) is at all valid."

2.5.2.3.1.2 Decollement Recent seismic reflection profiles reveal the presence of a low angle detachment surface or decollement beneath much of the southern and central Appalachians (References 2.5-88, 2.5-76, 2.5-77, 2.5-78, and 2.5-75). The decollement was active primarily during the Alleghenian orogeny when rocks of the Valley and Ridge, Blue Ridge and Piedmont were thrust tens of mile to the west and northwest (Reference 2.5-92). Harris and Bayer (Reference 2.5-88) and Cook and others (Reference 2.5-77) suggest that the decollement may extend from beneath the Valley and Ridge to the continental margin. Recent seismic profiles from the Coastal Plain and continental shelf, although not conclusive (Reference 2.5-97), appear to substantiate this eastward extent (References 2.5-78, 2.5-67, 2.5-68, 2.5-69, 2.5-121, and 2.5-122). The postulated depth of the decollement based on those profiles beneath Charleston, S.C., also agrees well with a seismic velocity discontinuity at about 10 km depth observed by Talwani (Reference 2.5-129).

Several authors suggest that seismicity in the eastern U.S. is caused by reactivation of the decollement (References 2.5-68, 2.5-69, and 2.5-123). The decollement is a fundamental mechanical discontinuity in the crust and if it has not been disrupted by subsequent Mesozoic extensional faults or intrusive complexes, it may be a zone of crustal weakness susceptible to reactivation by properly oriented intraplate stresses. Reactivation would be in either a forward (thrust) sense in response to NW-SE regional compression (Reference 2.5-68) or in a gravity-driven backward sense (References 2.5-123, 2.5-65, and 2.5-124). Large earthquakes such as the Charleston 1886 event would represent displacement or slippage on the decollement.

The decollement reactivation hypothesis as thus proposed may not be incompatible with the reactivation of steep basement faults. Seismic reflection profiles suggest that many high angle faults are listric into the decollement (Reference 2.5-78). Movement on the high angle faults would thus be viewed as second order displacements (but capable of producing earthquakes) accommodating horizontal movement on the decollement (Reference 2.5-68, pp. 121-122). Noting that at least one listric fault appears to project to a Triassic basin border fault, Cook et al (Reference 2.5-78, p. 742), speculate that perhaps the listric thrust (reverse) faults were reactivated during Mesozoic rifting as listric normal faults to produce the Triassic basins.

The following geologic and seismic data are cited as evidence supporting decollement reactivation:

(1) the decollement is the most likely fault that could achieve significant differential offset without producing surface displacement. (2) Common to all the focal mechanism solutions supporting displacement on high angle northeast-trending reverse faults (References 2.5-132, 2.5-70, and 2.5-

63) is one subhorizontal nodal plane which supports displacement on the decollement. (3) Seismic reflection profiles indicate the structural disruptions and other geometric impediments to reactivation are not present across the decollement (References 2.5-78, 2.5-123, and 2.5-140). (4)

Releveling data by Citron and Brown (1979) indicate that vertical crustal movements as far west as the Blue Ridge front may be related to gravity-induced back-slip along the decollement (Reference 2.5-123). The maximum primary stress orientation may be different above and below the decollement supporting a decoupled upper sheet (Reference 2.5-123).

CHAPTER 02 2.5-37 REV. 13, SEPTEMBER 2006

LGS UFSAR On the other hand, active movement along a decollement appears unlikely for the following reasons:

a. Continuation of the decollement beneath the coastal plain is not certain. The subhorizontal seismic reflectors representing the decollement (Reference 2.5-78) widen, become discontinuous and dip steeply southeastward beneath the outer Piedmont where magnetic and gravity data have been interpreted to suggest a root zone (References 2.5-102, 2.5-92, and 2.5-93). Recent palinspastic restorations of the crust by Iverson and Smithson (Reference 2.5-97) indicate that the decollement is rooted to the east beneath the outer Piedmont and does not extend to the east beneath the Coastal Plain.

b. Although Cook et al (Reference 2.5-78) and Seeber and Armbruster (Reference 2.5-123) suggest that no offsets of the decollement or other geometric impediments to reactivation exist, figure 7 of Reference 2.5-78 seems to show a series of normal faults within the inner Piedmont which displace basement Precambrian Grenville rocks as well as overlying Precambrian and Paleozoic sedimentary rocks beneath of the decollement. If these extensional faults are related to Mesozoic rifting, which they appear to be because they displace Paleozoic strata, then they must also displace the overlying decollement. Ellwood et al (Reference 2.5-81) also cite paleomagnetic data, radiometric age data and geologic relationships indicating that a major, relatively undisplaced intrusive complex has been emplaced through the decollement. Mafic dikes and aulocogens related to Mesozoic rifting (Reference 2.5-117) would also have to be truncated after rifting. It is not probable that gravitational back-slip could overcome these impediments to reactivation, and it is difficult to account for a source of relatively shallow crustal compression of a passive, presumably subsiding continental plate margin.

c. Seismic data from the Giles County, Virginia seismogenic zone are not consistent with decollement reactivation. Hypocentral depths range from 5 km to 25 km (Reference 2.5-70) which places it beneath the decollement in the crystalline basement under the deformed Paleozoic sediments of the Valley and Ridge province.

d. Seeber and Armbruster (Reference 2.5-123) speculate that the stress-field above and below the decollement may differ. Composite focal plane solutions from the Charleston area, however, indicate a consistent orientation of maximum stress across the postulated depth of the decollement suggesting that there is no decoupling of the stress-field (Reference 2.5-129).

e. If gravitational backsliding is responsible for seismicity, then extensional stresses resulting in normal faulting should be evident in the overthrust plate. However, many focal mechanism solutions indicate reverse displacements in the overthrust plates, and geologic and geophysical evidence suggest that some Cenozoic faults in the overthrust plate have reverse displacements (References 2.5-134 and 2.5-135).

f. It does not seem likely that the gravitational force resolved parallel to the postulated, extremely low angle decollement is sufficient to overcome normal frictional shear resistance of rock materials within the decollement.

CHAPTER 02 2.5-38 REV. 13, SEPTEMBER 2006

LGS UFSAR Deep seismic reflection data (COCORP and other lines) suggest that the central and southern Appalachian system, including at least the Blue Ridge and inner Piedmont provinces, are underlain at depth by a low angle regional thrust, or decollement, resting on relatively unmetamorphosed sedimentary rocks, extending westward beneath the Valley and Ridge province. This decollement was presumably active during the Alleghenian orogeny at the end of the Paleozoic Era; ramp-like displacements branching upward from the decollement produced the well-known thrust sheets of the southern Appalachian Valley and Ridge and were taken up as faulted anticlinal folds in the central Appalachians. The eastern extent, or "root zone", of the decollement is a matter currently in dispute (Section 2.5.1.1.3). The inclination of the decollement plane is flat; profiles (Reference 2.5-78, figure 7, and Reference 2.5-88, figure 5) show it to be a horizontal to subhorizontal surface, broken perhaps by local offsets or minor ramps, but not exceeding about 21/2 overall, sloping gradually downward to the east.

This new evidence of a decollement east of the Blue Ridge helps to explain the mechanics of the Alleghenian orogeny. Heretofore it has been difficult to explain the large-scale westward thrusting that produced the Valley and Ridge structures in late Permian or early Triassic time, in the face of so little surface evidence for such thrust faulting, or for a source of compression, in the metamorphosed rocks to the east, which show abundant evidence of only earlier Paleozoic orogenies. On the other hand, its relevance, if any, to current seismicity is unclear. Speculations that gravity-induced backsliding may be occurring on the decollement and thereby causing earthquakes, in particular the Charleston event of 1886 (References 2.5-123 and 2.5-124) seem implausible in view of the extremely low coefficient of friction that would be necessary for gravitational slippage to occur on a subhorizontal plane, even discounting structural and lithologic impediments that would be expected to exist along the decollement surface. Similarly, the inverse hypothesis of a compressional reactivation of the decollement proposed in Reference 2.5-68 to account for events such as Charleston seems improbable. The source of such compression at the edge of a passive, presumably subsiding continental margin acting on a relatively shallow crustal feature is difficult to envision.

The possibility of a continuation of the Alleghenian decollement into the New England Appalachians is not in accordance with recent studies (Reference 2.5-133). Such decollements as may occur there relate to the Taconic orogeny (Reference 2.5-26); these are found along the western and northern margin of New England (References 2.5-133 and 2.5-136), east of Logan's line. Subsequent Acadian orogeny and metamorphism (Reference 2.5-26) have obliterated the regional mechanical significance of such decollements, and there is no evidence of extensive regional Alleghenian thrusting in New England. As pointed out in Reference 2.5-28 and by others since 1975, strike-slip rather than thrust faulting appears to have characterized at least the latter part of late Paleozoic deformation in New England. This deformation is best developed as displayed in eastern New England and as inferred in the Gulf of Maine.

The idea of reverse reactivation of pre-existing NE-SW striking faults generated by inferred, present day NW-SE oriented regional crustal compression has gained popularity with some investigators (References 2.5-140, 2.5-134, and 2.5-135). Early Mesozoic (i.e., Jura-Triassic) faults having this orientation are attractive candidates because they presumably had a deep-seated origin related to continental rifting, but other smaller faults seaward of the edge of the Coastal Plain showing evidence of reverse offset of Cretaceous and younger strata are also considered, provided they have the appropriate NE-SW strike.

It seems somewhat presumptuous to define by inference a regional stress-field on the basis of fault orientation, or even of focal plane mechanism solutions, in the Appalachians. The reason is that CHAPTER 02 2.5-39 REV. 13, SEPTEMBER 2006

LGS UFSAR the occurrence of strain release of modern stress would be expected to be constrained by the existing structural planes of rock weakness, regardless of the "real" stress orientation, provided the stress levels are moderate, as seems appropriate for a passive continental margin. Because the structural grain of the Appalachians, which was imposed during Paleozoic orogenies, has a very pronounced and persistent NE-SW orientation, often almost to the exclusion of other trends, earthquakes in the Appalachians occur mostly on NE-SW trending planes. The same problem exists when the argument is turned around to infer the orientation of modern stresses from the trend and sense of displacement of existing faults (for example, faulting of coastal plain strata) because presumably these faults are simply new expressions of old basement faults that have been reactivated. If the faults do not extend to basement, they are of no significance.

Here there is an additional critical problem; namely, the justification for assuming that displacements for which there is no proof of movement over the past several million to several tens of millions of years have any meaningful relationship to modern stresses or seismicity.

In References 2.5-134 and 2.5-135, it states that a northeast- trending reverse fault may have been responsible for the Charleston event of 1886 and cites the inference of existing northwest directed compressive stress in the eastern U.S. from focal plane and fault orientation studies (Reference 2.5-140). They review the evidence indicating Cretaceous and Cenozoic reverse offsets on NE-SW trending faults in the eastern U.S. and propose that such faults should be viewed as potential sources of large earthquakes, such as the Charleston event of 1886. Their conclusion is based on dating different measured displacements on the same or different faults and deriving rate curves, extrapolating continued episodic movement to the present from the most recent documented movement on any of the faults (Reference 2.5-134, figure 5). These rate curves are assumed to represent a hypothetical family of both known and as yet undiscovered NE-SW reverse faults, an assumption Wentworth and Mergner-Keefer consider reasonable in view of the limited area for which such documentation can be obtained.

Critical to the presumption (References 2.5-134 and 2.5-135) of significant potential for present day movement on these hypothetical faults is the age and amount of the most recent offset plotted in their data set because the older and smaller an offset is, the less justification there is in presuming continued activity to the present. They (Reference 2.5-134, figure 5) present only one data point younger than Eocene: namely, an offset measured at an exposure of the Stafford fault system in Virginia. This particular exposure, which occurs along the Fall Hill fault of the Stafford fault system, is diagrammed in a field trip guide by Mixon and Newell (Reference 2.5-107, figure 7). It shows a sharp, 14 inch projection of metamorphic rock (gneiss) upward into overlying upland gravels (late Pliocene to early Pleistocene in age), consistent with the gneiss having moved upward in a reverse sense on a steep, west dipping plane. Below this, the metamorphic rock has been reverse-faulted against Cretaceous alluvial deposits (Potomac Group) with obvious deformation and significant offset. However, the above described offset in the overlying fluvial gravels apparently is accompanied by little or no deformation, in that no shear planes are shown in Reference 2.5-107 extending upward into the gravel along the plane of the fault, and the gravels in the sketch appear structureless or vaguely "draped" across the offset. Thus, it would seem that an equally plausible explanation for the origin of the "offset" in the gravels is that it is an erosional feature, particularly when one considers the contrast in erodability between the Cretaceous sands and the metamorphic rock, the relatively high energy depositional environment of the gravels, and the presence of angular metamorphic boulders in the gravel (Reference 2.5-107, figure 7).

CHAPTER 02 2.5-40 REV. 13, SEPTEMBER 2006

LGS UFSAR This alternative explanation is reinforced by the trenching across the Fall Hill fault that was performed by Dames and Moore (Reference 2.5-113). In the three trenches that exposed the base of the Plio-Pleistocene alluvial gravels resting on the trace of the Fall Hill fault, the gravels were plainly not offset across the trace of the fault.

The above evidence calls into question the validity of the single data point younger than Eocene (40 million years) of Wentworth and Mergner-Keefer's rate curve (Reference 2.5-134, figure 4) for reverse fault movement. Therefore, there is no solid evidence for movement younger than Eocene on any of the NW-SE trending faults used by Wentworth and Mergner-Keefer in their analysis.

Such faults should not be considered to be capable within the content of the siting criteria.

The basis for presuming that the Charleston seismicity is associated with a northeast-trending structure is also questionable. Although Wentworth and Mergner-Keefer cite the data of Tarr et al (Reference 2.5-131) in support of a northeast structure, Tarr et al include only three of a total of 16 focal mechanism solutions from the Middleton Place-Summerville (i.e., Charleston) area in such a trend; moreover, Tarr et al choose this trend not so much from focal mechanism solutions but from gravity anomalies and inferred fault trends (Reference 2.5-131, p. 1892, 1898). Specifically, Tarr et al state that the three focal mechanism solutions are "consistent with early vertical faulting on a plane striking northwest to E-NE, with one inconsistency... A nearly vertical E-NE striking plane was chosen because of independent geophysical evidence..." (Reference 2.5-131, p. 1892). The other 13 focal mechanism solutions which Tarr et al studied define two alternate, relatively well constrained nodal planes that strike northwest, thus implying a stronger likelihood of northwest-faulting at Charleston, and suggesting the presence of unique structure. This is further corroborated by recent detailed focal plane studies by Talwani (Reference 2.5-129), with no evidence found for northeast-trending reverse faulting. It does define reverse motion along a relatively shallow, steeply dipping northwest-oriented plane and strike-slip motion along a deeper, northeast-trending plane. Therefore the Charleston area, rather than being structurally and tectonically typical for the Coastal Plain and surrounding region, seems more likely to be just the opposite, based on current data and analysis.

2.5.2.3.1.3 Stress Amplification at the Margins of Mafic Plutons The presence of mafic plutons in certain areas of historical or current seismic activity, such as in the New Madrid area and at certain locations in New England, has been pointed out by a number of observers; whether these plutons have the capacity to localize earthquakes is a matter of considerable conjecture. Investigators have suggested that stress is mechanically concentrated in or near the plutons because of contrasts in elastic moduli with the surrounding country rock (References 2.5-103, 2.5-104, and 2.5-99). They theorize that the stress concentrations may be sufficiently high to result in earthquakes.

According to these stress modeling studies, stress can be concentrated in or near mafic plutons by one of three methods: (1) The instructions are "softer" or less rigid than surrounding rocks because of serpentinization, and deform by creep rather than fracture. Stress is thus concentrated in the more rigid country rock surrounding the intrusion (Reference 2.5-99). (2) The intrusions are "stiffer" or more rigid than the surrounding rocks and stress is concentrated within the intrusion (References 2.5-104). (3) The intrusions are weakened by pervasive fracturing and high pore pressure caused by residual volatiles associated with their origin and emplacement. Stress is therefore concentrated in the more rigid country rock.

CHAPTER 02 2.5-41 REV. 13, SEPTEMBER 2006

LGS UFSAR McKeown (References 2.5-103 and 2.5-104) studied the association between mafic plutons and seismicity primarily in the New Madrid and Charleston area on the basis of geologic, gravity, aeromagnetic and seismic data. He noted that the trend of nodal planes of many fault-plane solutions coincides with the trend of many mafic dikes. He suggests that earthquakes are controlled by the presence of mafic intrusions and that ancient rift zones may have been the primary control for the location of the intrusions.

References 2.5-132, 2.5-118, 2.5-62, 2.5-128, 2.5-101, 2.5-102, and 2.5-140 also report an empirical spatial association of seismicity and mafic plutons interpreted from surface exposures and shallow source gravity and magnetic anomalies. Tarr and Rhea (Reference 2.5-132) further note that in the broad areas of flat or negative anomalies, seismicity is lacking. They indicate that, although earthquakes in the Piedmont tend to be scattered, those in the Coastal Plain tend to Cluster. They interpret the clustering to result from localized stresses on or near intersecting faults and mafic plutons.

Kane (Reference 2.5-99) speculated that the mafic bodies at Charleston are serpentinized.

Analytical modeling by Campbell (Reference 2.5-73) supports Kane's hypothesis as an explanation for the Charleston earthquake. Kane suggests that the lack of seismicity at other geophysically inferred mafic and ultramafic plutons is due to the lack of serpentinization, an insufficiently large or randomly changing stress-field, or inappropriate geometric relations between the intrusive body and stress-field direction.

The stress amplification hypotheses are beset with a number of uncertainties and unknowns due chiefly to insufficient information. Some of the major problems and uncertainties are:

a. In most cases where earthquakes are clearly related to geologic structures, the earthquakes occur along planes or zones of weakness (i.e., faults) rather than in areas of concentrated stress in the neighborhood of relatively rigid bodies.

b. Although some intrusions are spatially associated with seismicity, there are many large and small intrusions (such as the Cortlandt Complex near the Ramapo fault, Reference 2.5-120) which are either not associated with seismicity or have no effect on localizing seismicity.

c. Presently it does not appear feasible to obtain field data in sufficient to test the proposed models.

d. Mafic/ultramafic intrusions are common in zones of crustal rifting. If seismicity is caused by the reactivation of basement structures such as triassic association of seismicity with mafic/ultramafic plutons may be simply a secondary coincidence.

e. The inferred mafic/ultramafic bodies are not necessarily localized intrusions. The mafic body beneath Charleston penetrated by core holes in a Jurassic basalt flow.

The flow is laterally continuous beyond the limits of concentrated historic seismicity (References 2.5-78, 2.5-122, and 2.5-68).

2.5.2.3.1.4 Subsidence Several workers have noted that east coast seismicity may be spatially associated with subsiding embayments on the Cretaceous continental margin and Triassic basin border faults inland from the CHAPTER 02 2.5-42 REV. 13, SEPTEMBER 2006

LGS UFSAR continental margin. They suggest that the seismic activity may represent minor adjustments in the crust associated with continued opening of the Atlantic Ocean.

Barosh (Reference 2.5-66) empirically observed the relationship between seismicity and structural depressions. He hypothesizes that current seismicity along the Atlantic seaboard is caused by small-scale vertical movements within pre-existing fracture zones and subsiding embayments. The seismicity, for example, is concentrated along the northeastern margin of the subsiding Southeastern Georgia Embayment.

Following the Triassic-Jurassic extension that resulted in the development of Triassic basins and the Atlantic Ocean, the trailing continental margin has sagged and accumulated a mantling wedge of Cretaceous and Tertiary sediment. Subsidence of the margin may be due to sediment loading (Reference 2.5-126), thermal contraction (Reference 2.5-125), thermally induced phase changes (Reference 2.5-109), or oceanward creep of lower continental crust due to gravity (Reference 2.5-71). Barosh (Reference 2.5-66) suggests that seismicity is localized in those areas along the coast where the lowlands are subsiding over the ancestral margin of the continent.

Subsidence in areas along the Atlantic coast is known to be occurring (Reference 2.5-95).

Subsidence as a cause for seismicity, however, is not consistent with several facts and observations:

a. Many focal mechanism solutions indicate reverse displacement on either high angle or low angle faults. Reverse displacement is not consistent with a tensional, subsiding stress regime.

b. Many observed Cenozoic faults in the Coastal Plain have reverse and not normal displacement.

c. The nearly aseismic nature of the offshore region (continental shelf) where subsidence has been greatest argues against subsidence as a primary cause of seismicity.

2.5.2.3.1.5 Conclusion Although the recent geologic and geophysical data summarized above demonstrate the significant role of compressional and extensional deformation in the evolution of eastern North America, they do not provide consistent evidence that these older deformational structures have been regionally reactivated and pose a widespread, significant potential for damaging earthquakes.

Correlation of Charleston seismicity and the Charleston 1886 earthquake with a steep reverse northeast-trending basement fault is extremely tenuous. Composite fault-plane solutions, hypocentral location patterns and the inferred and measured orientation of the principal maximum stress direction implies a coherent pattern of deformation unique to the Charleston area. Fault planes trend northwest with reverse displacement in the upper crust and trend northeast with strike-slip displacement in the lower crust. The vertical maximum principal stress axis near the surface (Reference 2.5-127) and the deeper northeast-trending maximum principal stress axis (Reference 2.5-129) are not consistent with the northwest- oriented stress axis required by the hypotheses for reactivation of steep basement faults (References 2.5-134 and 2.5-135) or thrust reactivation of the decollement (Reference 2.5-68).

CHAPTER 02 2.5-43 REV. 13, SEPTEMBER 2006

LGS UFSAR Apparent association of small earthquakes with the Ramapo fault is not reflected by any obvious concentration or association of historical macroseismicity in its vicinity; the widely scattered historical epicenters within at least 25 kilometers of the Ramapo fault do not exceed an intensity of VI (Reference 2.5-82). The Ramapo fault has been intensively studied by drilling, trenching and detailed mapping. No evidence for reactivation of the fault surface has been found, and no instance of surficial offset along the fault, either in natural exposures or in numerous trenches, has been discovered (Reference 2.5-119).

It is concluded that the lack of recent faulting and the modest size of even the largest historic earthquakes within 200 miles of the site and the general scatter and infrequency of reported earthquakes argue against the meaningful association of regional macroseismicity with specific faults.

2.5.2.4 Maximum Earthquake Potential No earthquake within the 200 mile radius site region has exceeded intensity VII during the historic record for this area, which began in the early 18th century. There have been 6 intensity VII shocks during this period in the site region. Of these, 2 occurred near New York City, in 1737 and 1884, at the edge of the Newark Basin, near the junction of the Piedmont, New England, and Coastal Plain provinces; one occurred near Wilmington, Delaware in 1871, about 35 miles south of the site, near the Piedmont-Coastal Plain boundary; one occurred along the New Jersey coast in 1927, about 80 miles northeast of the site in the Coastal Plain; one occurred near Moodus, Connecticut in 1791, about 180 miles northeast of the site in the New England Upland section of the New England Province; and one occurred in Wilkes-Barre, Pennsylvania in 1954 in the Folded Appalachians to the N-NW of the site and was almost certainly not related to either tectonic strain accumulation or release, as these terms are normally understood. One additional intensity VII event, which was slightly more than 200 miles from the site, occurred near Richmond, Virginia in 1875, close to the Piedmont-Coastal Plain province boundary. Thus, most of the intensity VII events recorded, and particularly those south of the New England Appalachian-Piedmont Coastal Plain physiographic province junction, have occurred near the Fall Zone. An apparent NE-SW trend of smaller earthquakes occurs along the same zone, although intensity VI and smaller events are scattered throughout the site region.

Considering the historic seismicity of the site region alone, a reasonable interpretation of maximum potential earthquake might be either an intensity VII event along the Fall Zone at its closest approach to the site or an intensity VI event very near the site. Because of the uncertainties involved in associating regional activity with specific structures, however, the maximum potential earthquake is specified as being equivalent to the Io = VII 1871 Wilmington, Delaware earthquake occurring near the site. This is equivalent to translating the largest historical earthquake that has occurred anywhere within 200 miles of the site.

2.5.2.5 Seismic Wave Transmission Characteristic of the Site Detailed descriptive data on the foundation material properties appear in Section 2.5.4. All Category I structures are founded on competent bedrock, except some buried structures including portions of underground piping and electrical ducts, diesel oil tanks, and buried valve pits. An analysis was performed to determine the response of the soil column and the effects of the properties used in the analysis are given in Section 2.5.4.2.2, and the results of the analyses are given in Section 2.5.4.7.1.

CHAPTER 02 2.5-44 REV. 13, SEPTEMBER 2006

LGS UFSAR The seismic wave velocity of the in situ soils and the fill materials, on which the buried structures are founded, are not available. However, because of the shallow foundation soil depth to the competent bedrock and the smooth transition from rock to soil foundation (Figure 2.5-37), LGS Category I underground piping is designed in accordance with the procedure identified in Section 3.7.3.12 for seismic load, assuming the seismic shear-wave velocity of the foundation soil is the same as the competent bedrock. The application of the bedrock seismic wave velocity to the foundation soil material in buried piping analysis is shown in Reference 2.5-53.

The LGS design response spectra, discussed in Section 3.7.1 and Figures 3.7-1 and 3.7-2, are based on data developed from records of previous earthquake activities representing an envelope of motion expected at a sound rock site.

2.5.2.6 Safe Shutdown Earthquake The SSE is defined in terms of a peak ground acceleration and a design response spectrum. As indicated in Section 2.5.2.4, the site design intensity is VII on the Modified Mercalli Scale. This intensity may be associated with a peak ground acceleration of approximately 0.13 g (References 2.5-34 and 2.5-35). For additional conservatism, a peak acceleration of 0.15 g the design response spectrum shown in Figure 3.7-2 for critical damping values of 0.0%, 0.5%, 2.0%, 5.0%, and 10.0%.

2.5.2.7 Operating Basis Earthquake The OBE is defined as one-half the peak ground acceleration associated with the SSE, 0.075 g.

The design response spectrum is identical in shape to that specified for the SSE. This value (0.075 g) is the high frequency asymptote of the OBE design response spectrum shown in Figure 3.7-1. Curves for the same percent critical damping values appearing in the SSE design response spectrum are shown.

2.5.3 SURFACE FAULTING 2.5.3.1 Geologic Conditions of the Site LGS is located approximately 3 miles southeast of Pottstown, Pennsylvania, adjacent to the Schuylkill River. It is situated within the western portion of the Newark Basin, which is part of the Triassic Lowland subdivision of the Piedmont physiographic province (Section 2.5.1.1). The Triassic Lowlands represent sedimentary basins that form outcrop patterns broadly parallel to the sinuous, folded Appalachians and extend from Maine to South Carolina. The basins contain rocks of relatively similar gross lithologic types, usually occurring in the same stratigraphic sequence (Reference 2.5-6).

The Newark Basin, together with the contiguous Gettysburg Basin further west, forms the largest of the Triassic basins. This basin extends in a W-SW direction from the Hudson Palisades, near the New Jersey-New York boundary to the Catoctin Mountains, near Frederick, Maryland (Figure 2.5-38). The basin varies in width from 3 miles in its narrowest section to almost 35 miles at its greatest width. The Newark-Gettysburg Basin represents a series of some 18,000-20,000 feet of nonmarine sedimentary rocks with associated basaltic flows, and diabase dikes, sills, and stocks (Figure 2.5-6). For detailed descriptions of the regional geology, see Section 2.5.1.1. Site geologic conditions are described in Section 2.5.1.2.

CHAPTER 02 2.5-45 REV. 13, SEPTEMBER 2006

LGS UFSAR Several faults of apparently large displacement occur within the Newark Basin. These are the Chalfont and Furlong faults in Pennsylvania, and the Flemington and Hopewell faults in New Jersey. These faults, which are of Triassic-age, result in Paleozoic age rocks being exposed at the surface, indicating at least 10,000 feet of displacement. The orientation and direction of movement of these faults are not known. Although generally considered to be steeply south dipping normal faults (Reference 2.5-36), Sanders (Reference 2.5-37) suggests predominant strike-slip movement, and Faill (Reference 2.5-17) indicates these faults may be high angle reverse faults resulting from intersection of two different axes of monoclinal folding within the basin.

Smaller Triassic faults cross the basin margins and extend well into the surrounding rocks. They usually show less than 3000 feet of displacement. Associated with these faults are local concentrations of smaller faults.

The major fault closest to the site is the Chalfont fault, which passes 9 miles to the northeast (Figure 2.5-2). The maximum throw on this fault is 6500 feet, in Bucks County, Pennsylvania, near Bennetts Corner (Reference 2.5-38). The throw decreases to about 3000 feet at the Montgomery County line and eventually dies out to the northeast of the site. The northeastward branch of the Chalfont fault (at Bennetts Corner) is called the Furlong fault. The fault crosses the Delaware River near Center Bridge, Pennsylvania and continues through New Jersey (where it is called the Flemington fault) to just beyond the north border of the Triassic basin. The maximum throw on the Furlong-Flemington fault is 10,000 feet, with an undetermined amount of right-lateral, strike-slip displacement (Reference 2.5-37). The displacement along the Furlong fault in eastern Pennsylvania is sufficient to offset the entire thickness of the Newark Group and expose the Cambrian rocks on the floor of the basin. Five and one-half miles southeast of the Furlong fault, the east-west trending Buckmanville fault crosses the river and becomes the Hopewell fault of New Jersey. A throw of 15,000 feet has been measured on the Hopewell, with a right-lateral, strike-slip displacement of 12 miles (Reference 2.5-37). Other faults associated with the Furlong-Buckmanville complex and their eastward extensions into New Jersey are, the Holmquist School, Bridge Valley, Pidcock Creek, and Danborough faults. These faults are all normal, with the downthrown side to the south. Some show lateral displacement, and all are strike faults.

Many of the minor faults in the Newark Basin of southeastern Pennsylvania are transverse normal faults. These have developed along the major joint system exhibited by the Triassic strata, and they trend approximately N 30 E. Some of the Triassic dikes have intruded along these fracture zones subsequent to movement along them. The Downingtown Dike is an example of this type of intrusion.

In this report the minor faults that occur in the site area have been named the Sanatoga fault, the Brooke-Evans fault, and the Linfield fault, as shown on Figure 2.5-6. The Sanatoga fault passes within 1300 feet to the west of the reactor location, the Brooke-Evans fault is 2800 feet to the south, and the Linfield fault is 2 miles to the southeast.

Detailed field investigations of faults in the site area were performed by Dames and Moore in 1974 (Reference 2.5-1).

2.5.3.2 Evidence of Fault Offset As previously indicated, the site lies within a down-warped basin of Triassic-age. Subsequent to the deposition and consolidation of the basin sediments, the region was uplifted, and numerous small faults and fractures were formed.

CHAPTER 02 2.5-46 REV. 13, SEPTEMBER 2006

LGS UFSAR The Sanatoga fault, shown on Figure 2.5-6, trends approximately N 25 E, becoming discontinuous approximately 2 miles to the north and 1 mile south of the site area. The nearest approach of the fault trace to the reactor area is 1300 feet to the west. Detailed studies by Dames and Moore in 1974 (Reference 2.5-1) indicate vertical displacement on the fault is approximately 290 feet, with the downthrown side to the east. The fault-plane is occupied by a Triassic diabase intrusive, which is part of the Downingtown Dike.

The Brooke-Evans fault passes within 2800 feet to the south of the plant area and trends approximately N 50 E. Slickensides on the fault-plane indicate some lateral movement. Field evidence based on the offset of marker beds indicates that the movement was probably both right-lateral and vertical. The apparent vertical displacement is about 350 feet, with the downthrown side to the south. West of the Schuylkill River, magmatic material from the Downingtown Dike occupies the fault for a short distance. A high concentration of calcium carbonate filling occurs in the fracture zone associated with fault.

A third fault, the Linfield fault, is located east of the town of Linfield about 2 miles southeast of the site. This fault trends N 20 E, with the downthrown side to the west. Displacement on this fault is approximately 130 feet. Field evidence based on slickensides in the fault-plane indicates an undetermined amount of lateral movement. This fault is associated with a diabase dike that occupies a parallel fault approximately 700 feet to the east.

Detailed inspection of rock cores and nearby outcrops disclosed the existence of other very small displacements (maximum less than 2 feet), especially in those areas close to the faults described above. On the basis of detailed stratigraphic correlation in the site area done for the PSAR, it was concluded that the strata between the Brooke-Evans fault to the south and the Sanatoga fault to the north have not been offset by faulting. The rock at the site is moderately-to-well jointed. A statistical analysis of joints by Dames and Moore in the rock cores revealed a higher joint frequency along the northeast trend through the southeastern portion of the construction area, an the PSAR noted that the presence of these abundant joints may be related to an undetected fracture zone. Fracture zones with minor offsets were encountered in the excavation. These are described in Section 2.5.3.2.1.

As noted in the PSAR, the presence of minor faults within the Newark Basin in general, and near the site area in particular, is not surprising, nor is it hazardous to the operation of the plant. As previously discussed, such minor faulting is common throughout the basin. Regional crustal stresses during Mesozoic time resulted in numerous small-scale readjustments and fracturing of the brittle, flat-lying Triassic strata. Small-scale insignificant displacements are likely to be found in any major rock exposure within the basin.

Evidence that confirms the regional inactivity of the Triassic faults includes the following:

a. The regional uplifting forces that induced the basin faulting no longer exist.

b. There is no evidence that faulting has displaced overlying Coastal Plain sediments.

c. Many faults are known to be healed by secondary mineralization.

d. Within the same Triassic rock-type, no fault exhibits surface expression in the form of fault scarps.

CHAPTER 02 2.5-47 REV. 13, SEPTEMBER 2006

LGS UFSAR These faults were discussed in the PSAR (section 2.5.2.3.4) and were described as inactive for at least the last 140 million years. The PSAR stated: "Perhaps the most conclusive proof of tectonic inactivity near the site is the presence of unfaulted, intact diabase, at least 140 million years old, along the nearby fault planes. The faults near the site have not moved in at least 140 million years and will not move during the life of the proposed facility."

Subsequently, some evidence for possible offset of the diabase was observed, and an extensive geologic study was made by Dames and Moore (Reference 2.5-1) in an attempt to date as closely as possible the age of the faulting in the area.

A committee consisting of Dr. D.U. Wise, University of Massachusetts (Amherst); Dr. C. Gray, Franklin and Marshall College, Lancaster, Pennsylvania; and Dr. P.B. Myers, Lehigh University, Bethlehem, Pennsylvania, was selected to evaluate the study. In evaluating the faults in the area around the site, they state: "The fault and fold patterns are indicative of a Jura-Triassic (150-200 million years) stress system with maximum compression oriented NNE... The present WNW orientation of compressive stresses in the residual stress measurements differs markedly from the NNE compression which produced the Jura-Triassic faults at the site and nearby folds. The effect is to increase normal stresses across the faults, increasing fractional resistance to movement, and decreasing available space for additional graben displacement."

2.5.3.2.1 Geologic Structures in the LGS Site Excavation The site stratigraphy consists of a sedimentary sequence of red shale, siltstone, and sandstone (Figure 2.5-12). Structural mapping of the excavation at the LGS site shown on Figure 2.5-13, confirms the existence of three fracture zones with small displacements (LGS PSAR, supplement 4, question 2.29).

All three zones are characterized by the following:

a. Two vertical, or subvertical, northeast-trending fractures, in addition to numerous closely spaced smaller fractures

b. Observable vertical stratigraphic offsets (1 inch to 4 feet), which vary along the length of a given shear zone

c. Mineralization of the shear zone or adjacent fractures by calcite, quartz, or fibrous clay (Reference 2.5-1, section 3.4)

d. Subvertical and subhorizontal slickensides, indicating both strike-slip and subsequent dip-slip offsets A detailed discussion of these shear zones is presented in a 1974 Geologic Report by Dames and Moore (Reference 2.5-1).

The minor faulting and shearing encountered within the site excavation are not unique to the area, and they have the same geometry as and similar mineralization to adjacent larger faults. These larger faults in turn are part of the pervasive fault-fold pattern that reflects an initial northeast-trending regional compressive stress. Extension features (small grabens and dikes) for the most part strike parallel to the compression axis, and dikes sometimes display an en echelon geometry.

CHAPTER 02 2.5-48 REV. 13, SEPTEMBER 2006

LGS UFSAR As one would expect, the fold axes, together with thrust shear joints in nearby areas, strike normal to the compression.

Upon completion of their 1974 study (Reference 2.5-1), Dames and Moore concluded:

"The results of this comprehensive investigation indicate that shears exposed in the excavation are not capable by AEC (NRC) definition. It is concluded that movement along the shears could not have occurred later than 500,000 years ago and in all probability, have been inactive since Jura-Triassic time some 150-200 million years ago."

The review committee (Dr. Wise, Dr. Gray, and Dr. Myers) concluded:

"We believe the small faults discovered in the LGS containment excavation have been thoroughly tied to regional 150-200 million year structural events of this area. The ancient stress-field orientations which caused the faults have been superseded by the present stress-field orientations which tend to lock the faults against further motion. The faults have undeformed or little deformed delicate minerals in them indicating the absence of strong shearing stresses in the containment area since cessation of mineralization, most likely in the same 150-200 million year range. There is no evidence of displacement of any feature as young as the undisturbed upland terrace surface which has a minimum age on the order of a half million years.

We conclude that a very strong circumstantial case has been made on structural ground for the containment faults being on the order of 150-200 million years old. An equally strong geomorphic case has been made precluding any movements in the last half million years.

The committee sees abundant evidence suggesting geologic antiquity of the faults. We have yet to see the first piece of evidence suggesting recent movements. Based on our collective local experience, none of us has seen anything about the faults onsite to indicate they are other than typical small movement features associated with the 150-200 million year old events in the area."

It is concluded that these small faults are not significant to the operation of the power plant.

2.5.3.3 Earthquakes Associated with Capable Faults There are no capable faults within 5 miles of the site. There have been no earthquakes in historic time with epicentral locations within 10 miles of the site. For correlation of epicenters and geologic structure, see Sections 2.5.2.2 and 2.5.2.3.

2.5.3.4 Investigation of Capable Faults There are no capable faults within 5 miles of the site. Shear zones encountered in foundation excavations have been mapped in detail (Section 2.5.1.2.5 and Figure 2.5-13). The geologic investigations demonstrate that the small fracture zones encountered during foundation excavation and the faults in the vicinity of the site are associated with the 150-200 million year old events in the area.

2.5.3.5 Correlation of Epicenters with Capable Faults CHAPTER 02 2.5-49 REV. 13, SEPTEMBER 2006

LGS UFSAR There are no capable faults within 5 miles of the site. No earthquakes with epicentral locations within 10 miles of the site have occurred in historic time. For regional correlation of epicenters with geologic structures, see Sections 2.5.2.2 and 2.5.2.3.

2.5.3.6 Descriptions of Capable Faults There are no capable faults in the region around the site extending to within 5 miles of the site. For a description of faults within 5 miles of the site, see Section 2.5.3.2.

2.5.3.7 Zone Requiring Detailed Fault Investigation As discussed in Section 2.5.1.2, some shear zones with small offsets were encountered during foundation excavation at the site. Structures of this type are not unusual in the region; however, these zones were mapped in detail and photographed as part of the site geologic record (Figure 2.5-13). The detailed geologic investigation included mapping of an area approximately 5 miles in diameter. A review committee was established (Section 2.5.3.2.1), and the results of all investigations were included in a report issued in July 1974 (Reference 2.5-1). The review committee concluded that the shear zones in the foundation excavation at the site and the faults within the area mapped are typical of features associated with 150-200 million year old events in the area (Section 2.5.3.2.1).

2.5.3.8 Results of Faulting Investigation See Sections 2.5.1.2 and 2.5.3.7; there are no capable faults within 5 miles of LGS.

2.5.4 STABILITY OF SUBSURFACE MATERIALS AND FOUNDATIONS 2.5.4.1 Geologic Features Bedrock at the site consists of well-indurated Triassic sandstones, siltstones, and shales that extend to a depth of several thousand feet. Bedding dips toward the north at 8 to 20. Site stratigraphy is presented in Section 2.5.1.2.3. Geologic structure in the site area is presented in Section 2.5.1.2.4. Bedrock is overlain by from 0-40 feet of residual soil, developed in situ by weathering and decomposition of the parent rock. The soil grades into weathered rock, then into fresh, unweathered rock; no clearly defined boundary exists between soil and rock.

During foundation excavation, some fracture zones with small displacements were encountered and were treated locally as required. Descriptions of the fracture zones are presented in section 2.5.1.2.5; their locations at final foundation grades are shown on Figure 2.5-13. Treatment of these fracture zones is discussed in Section 2.5.4.12.

Engineering evaluation of the site geology is discussed in Section 2.5.1.2.7. The bedrock at the site contains no unstable minerals or hazardous conditions. The stress regime within the bedrock materials is low and stable. There are no mines in the site area and no significant fluid withdrawal.

The bedrock in the construction area is competent and provides satisfactory foundation support for plant structures.

CHAPTER 02 2.5-50 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5.4.2 Properties of Subsurface Materials The principal plant structures are founded on bedrock. The spray pond is excavated partly in soil and partly in rock. All or portions of other facilities not founded on bedrock are founded on natural soil or manmade fills. The locations of the major plant structures are shown on Figure 2.1-3.

Results of laboratory tests for foundation and construction materials are presented in References 2.5-39 and 2.5-51, and in Sections 2.5.4.1.2, 2.5.4.2, and 2.5.4.10.

2.5.4.2.1 Properties of Foundation Rocks The seismic Category 1 reactor and diesel generator enclosures, as well as the turbine and radwaste enclosures, are founded on hard, competent bedrock. The bedrock consists of siltstone, sandstone, and shale of Brunswick lithofacies of Triassic-age. The Brunswick is described in Section 2.5.1.2. The Lockatong lithofacies, represented by the Sanatoga Members, interfinger with the Brunswick in the northern part of the site area. The Sanatoga Member consists of blue-gray calcareous argillite with two distinct beds of black carbonaceous shale. The spray pond is underlain by both the Lockatong and Brunswick lithofacies.

Bedding and jointing patterns are well developed in the foundation rocks. Bedding-plane spacing varies from a few inches to several feet. Bedding-planes strike generally east to west and dip to the north at 8 to 20. Two major joint systems are prevalent in the area. Both are vertical or nearly vertical; they strike approximately N 20 to 50 E and N 50 to 60 W. Three fracture zones and two minor clay seams along bedding were encountered in the main power block foundation excavations; they are described in Section 2.5.1.2.5. Treatment of these zones is described in Section 2.5.4.12.

Rock quality designation values were measured in a total of 81 boreholes, which include the 200-,

300- and M-series borings completed in 1970, and borings 400 and 401 completed in February 1971 (Figure 2.5-22). In general, rock quality designation values measured on the first core run (5-10 ft) in rock were very low, usually zero; with some exceptions, minimum rock quality designation values increased to approximately 50% or greater after the first 10-30 feet of rock cored. The lower rock quality designation values in the upper 10-30 feet of rock reflect poorer rock quality caused by rock weathering; this weathered material was removed during foundation excavation. The data and analyses discussed below demonstrate that the higher rock quality designation values are associated with the sound, unweathered bedrock that supports the foundations for the principal plant structures.

Because of the gradational nature of the upward transition from sound rock to soil, engineering properties in the zone of rock weathering can be expected to vary from soil-like properties near the top of the zone to properties approaching those of sound rock near the base of the zone.

Properties of sound, unweathered foundation rock were determined as follows.

Laboratory tests on 16 core samples, using substantially the same procedure as specified in the subsequently adopted ASTM D2938-71, indicate unconfined compressive strengths ranging from 6370-24, 540 psi with an average of 15,820 psi. Results of these tests are shown in Table 2.5-3.

Laboratory sonic tests on intact cores (ASTM D2845-69) (Table 2.5-13) from four borings yield an average compressional velocity of 12,060 fps.

Seismic refraction surveys were performed to determine P-wave and S-wave velocities for site foundation materials. P-wave velocities in rock range from about 7700 fps to 20,000 fps, with an average of about 12,500 fps; locations of refraction lines are shown in Figure 2.5-21. Shear-wave CHAPTER 02 2.5-51 REV. 13, SEPTEMBER 2006

LGS UFSAR velocities were determined to be about 6100 fps in a line perpendicular to the strike of the bedding (north-south). A line run approximately parallel to the east-west strike of the bedding measured a shear-wave velocity of 5800 fps. Poisson's ratio is calculated to be about 0.3. The dynamic modulus of elasticity calculated from the seismic data is 3x106 psi. Representative engineering properties of foundation rock are summarized in Table 2.5-11.

Because geophysical test methods were used to establish dynamic design rock properties, it is appropriate to consider a range of dynamic moduli. Accordingly, all LGS available geotechnical data have been evaluated to establish the most appropriate range of dynamic moduli to be considered. This approach is presented below. It was concluded from this evaluation that an appropriate LGS dynamic rock modulus range would be between 2.7x106 and 4.0x106 psi, a range of +/-20%. The seismic analysis of the containment structure and reactor control enclosures included the SSE effect as discussed in Section 3.7. These analyses use a Young's dynamic modulus for rock equal to 3.0x106 psi. Additional soil-structure interaction studies were performed for the containment structure and reactor and control enclosures to assess the sensitivity of the structural response to variations in the design basis rock modulus. Modal analyses have demonstrated that for a +/-50% range (of the design elastic modulus), variations in structural frequency do not exceed 10% for predominant modes. These results indicate that a reduction in rock modulus to 1.5x106 psi will not produce significant effects on structural response. It is thus concluded that the average dynamic elastic modulus value of 3.0x106 psi is adequate for design.

Examination of the LGS Design Basis Dynamic Rock Modulus The following actions have been taken to establish an appropriate range of dynamic elastic rock moduli: (1) re-examination of the data on seismic velocities and dynamic moduli of the rock at the site, as given in the Dames and Moore reports and in the PSAR and the FSAR, (2) study of relevant technical literature including the WASH-1301 report (Reference 2.5-155), and (3) discussions with Dr. A.J. Hendron, a recognized authority on rock foundation engineering.

(1) Seismic velocity data and in situ dynamic moduli at the LGS site.

Young's modulus (dynamic) is calculated from either P-wave or shear-wave velocities.

Ambraseys and Hendron (Reference 2.5-153) state, "for fairly competent rock masses (VP greater than 10,000 fps) the compressional wave velocity and an assumed value of Poisson's ratio ranging from 0.27 to 0.35 may be used to evaluate a dynamic Young's modulus for engineering purposes". The average compressional wave velocity in rock at the site is approximately 12,000 fps; therefore, it is appropriate to consider P-wave velocities in computing the dynamic modulus. Accordingly, the dynamic modulus value of 3.0x106 psi for the site has been re-examined considering both P-wave and S-wave velocity data from refraction surveys, and considering up-hole velocities measured in one hole at the site. Also, results of laboratory tests have been re-evaluated to determine sonic velocities of rock cores, at the recommendation of Dr. A.J. Hendron.

Shear-wave velocities were measured along two refraction lines in the power block area (Figure 2.5-21); Dames and Moore (Reference 2.5-154) reported Vs of 6100 fps along strike and 5800 fps parallel to dip direction, for an average of 5950 fps, as described in Table 2.5-11. Poisson's ratio calculated from the average P-wave and S-wave velocities is 0.33. Young's modulus calculated from the average shear-wave velocity and this Poisson's ratio is 3.1x106 psi.

Upper layer P-wave velocities measured across the entire site have approximately the same average value (12,020 fps) as those measured only in the power block area (11,880 CHAPTER 02 2.5-52 REV. 13, SEPTEMBER 2006

LGS UFSAR fps) but, as expected, the former data set has significantly higher scatter (standard deviation 2,660 fps vs. 930 fps). A test (F-test) for equivalence of variance shows conclusively that the scatter of the P-wave values for the whole site is unrepresentative of the data for the power block area only. Therefore, P-wave data primarily from the power block area have been considered for evaluation and analysis.

Within the power block area, Dames and Moore (Reference 2.5-154) measured P-wave velocities on lines 3E, 4 and 6 (Figure 2.5-21). In addition, seismic P-wave velocities east of the Unit 1 turbine building were recorded along line 3F. Neglecting the higher velocities detected from the more deeply buried rock strata, eight refraction velocities (PSAR figure 2.5.5) yield an average P-wave velocity of 11,880 fps with an unbiased 1 sigma range of 10,950 fps to 12,810 fps. Using the lower end of the range (10,950 fps), and a Poisson's ratio of 0.30 as recommended by Dame and Moore, (E) is calculated to be 2.9x106 psi; the upper end of the range corresponds to (E) of 4.0x106 psi. With a Poisson's ratio of 0.33 (as indicated by the ratio between average P-wave and S-wave velocities), the range for (E) is 2.7x106 to 3.7x106 psi.

From the up-hole survey (Dames and Moore, Reference 2.5-154) and PSAR figure 2.5.6, the weighted mean P-wave velocity (weighted by relating velocity of each layer to the relative thickness of that layer) is 11,650 fps, which is in agreement with the profile data.

The corresponding (E), with a Poisson's ratio of 0.3, is calculated to be 3.3x106 psi.

Sonic tests on cores taken at the site (nine determinations) (Table 2.5-13) have an average P-wave velocity of 12,060 fps. The ratio of average field measured P-wave velocity to laboratory measured P-wave velocity (11,880-12,060 fps) is 0.98. This ratio provides support for the statement provided below in this section that: "The agreement between laboratory and field velocity measurements suggests that the seismic refraction results are representative of sound foundation rock, for otherwise the field measurements would be significantly less than measurements on intact core. Therefore the field seismic velocity data should provide a reasonable quantitative index of the general character of the in situ foundation rock".

(2) Reduction of dynamic modulus for higher strain levels during SSE event The rigidity of soil decreases significantly with strain; however, this should not be translated directly to rock, particularly to hard rock. For example, the relative reduction of shear modulus for sands ranges from 15% to 35%, depending primarily on void ratio or relative density, as the strain level increases from 10-4% to 10-2% (Figure 2.5-47). The corresponding reduction in rock moduli for a similar increase in strain level would be expected to be significantly less.

A brief search of the literature was made for data on the effect of strain on the rigidity of rock. Also, the maximum strain that would occur in the foundation rock at LGS during the SSE event was computed.

A simple estimate of strain in rock during a seismic event can be obtained by assuming sinusoidal, plane, vertically propagating SH-waves in an elastic, homogeneous whole space. Under these assumptions, maximum shear strain is equal to the maximum SH-wave particle velocity divided by the propagation velocity of the medium. For rock, a peak particle velocity to peak acceleration ratio of 36 inches/second/g has been recommended CHAPTER 02 2.5-53 REV. 13, SEPTEMBER 2006

LGS UFSAR (Newmark and Hall, Reference 2.5-155). Using this ratio and the 0.15g SSE for the site, and estimate of about 5.4 in/sec is derived for the SSE ground motion at the site. The site near surface SH-wave propagation velocity is about 6000 ft/sec, so that:

STRAIN = 5.4 in/sec = 7.5x10-5 = 7.5x10-3%

72,000 in/sec Actual strain at or near ground surface would not be this high because this calculation ignores the zero stress-free and strain-free surface boundary condition.

Strain levels associated with seismic profiling pulses are presumed to be on the order of 10-4

%, depending on the shot charge and line spread (e.g., Silver and Seed, Reference 2.5-159). There is little published information quantifying to what extent typical rock materials deviate from linear-elastic behavior at these small strain levels. Raphael and Goodman (Reference 2.5-157), drawing upon rock mechanics studies conducted in France, state that the modulus computed from seismic profiling should be greater than the rock modulus which would pertain during earthquakes, but they provide little information on the specific ratio of E (earthquake) to E (seismic test) to be adopted for varying circumstances.

Schnable et al (Reference 2.5-158) plot shear modulus as a function of strain for rock (Figure 2.5-46). In this context, "rock" is generally considered to be a material with an S-wave propagation velocity of about 2500-3000 ft/sec or higher (Schnable et al, 1971, Reference 2.5-158 and Algermissen and Perkins, 1976, Reference 2.5-152). These authors indicate that the shear modulus should decrease by less than 10% as the shear strain increases from 10-4% to 7.5x10-3%.

A.J. Hendron (personal communication, 1983) stated that a reduction in dynamic modulus -

from E (seismic test) to E (earthquake) - is appropriate for rock because strain levels experienced during earthquakes are higher than the strains which are associated with the measurement of seismic velocities. For the site, Hendron favors tying such a reduction to the square of the ratio of field measured seismic velocities to laboratory measured velocities on core samples. Dr. Hendron has developed such a curve from his studies at the Nevada Test Site and other localities; that curve would plot well above the curve shown on Figure 2.5-45. He stated that the amount of reduction for higher strain during a seismic event is significantly less for dynamic conditions than it is for reducing dynamic E to static E, which is the purpose of Figure 2.5-45. On Hendron's revised graph, for example, he stated that the square of the ratio of field-to-laboratory P-wave velocities (VF/VL)2 of 0.87 correlates with a reduction in dynamic modulus Ed of about 22% at a strain level of 10-3 in/in (Hendron, personal communication, January 1983). Subsequently, Hendron prepared a brief report describing his additional data on variation of propagation velocity with strain and commenting on dynamic Young's Modulus for the site under SSE conditions. Based on the data on rock quality provided to him, it is his judgment that the reduction factor at LGS should be about 0.75 to 0.85.

The following is a summary of the measured seismic velocities at the site and the corresponding values of the dynamic modulus of elasticity computed from them.

P-Wave (from In Situ Field Surveys)

Average velocity - 11,880 fps Range of velocities (mean +/-1 S.D.) = 10,950-12,810 fps CHAPTER 02 2.5-54 REV. 13, SEPTEMBER 2006

LGS UFSAR E (Assuming Poisson's ratio = 0.30)

Average - 3.5x106 psi Range - 3.0x106 psi to 4.0x106 psi E (Assuming Poisson's ratio = 0.33)

Average - 3.3x106 psi Range - 2.6x106 psi to 3.7x106 psi S-Wave (From In Situ Field Surveys)

Average velocity - 5950 fps Range - 5800-6100 fps E (Average) - 3.1x106 psi (Poisson's ratio of 0.33)

E (Range) - 3.0x106 psi to 3.3x106 psi Sonic Velocity Tests (On Intact Cores in Laboratory)

Average - 12,060 fps Ratio of field velocity to lab velocity VF/VL

= 11,880/12,060 = 0.98 Up-Hole Survey Measurements Weighted average velocity - 11,680 fps E (with Poisson's ratio = 0.33) - 3.0x106 psi E (with Poisson's ratio = 0.30) - 3.3x106 psi Maximum Range of E - 2.6x106 psi to 4.0x106 psi (Range with +/-1 S.D. of P-wave and S-wave velocities, and up-hole velocities, using Poisson's ratio of 0.30 and 0.33)

E used for design - 3.0x106 psi (lower 28 percentile of the maximum range).

The seismic velocity data of the foundation rock at the site indicates that the dynamic (E) value used for design (3x106 psi) represents the lower portion of the data range. The available data do not justify a range of dynamic moduli of +/-50% of the design basis value. Data from Schnable et al (Reference 2.5-158), for example, indicate that a reduction of less than 10%, for rock of undefined quality, is appropriate. Considering the rock quality at the LGS site is indicated by a velocity ratio squared of about 0.87, Hendron recommends a reduction of about 20%. (Actually, the velocity ratio squared of the rock at is about 0.96, based on the few lab velocities measured.)

It is therefore concluded that 3x106 psi is an appropriate value for the average dynamic modulus of elasticity of the foundation rock at the site. The range of moduli to be taken into consideration for design should be 2.7x106 psi to 4.0x106 psi. Additional discussion of seismic refraction survey techniques is given in Section 2.5.4.4; plate bearing tests and static moduli are described in Section 2.5.4.10.

The agreement between laboratory and field velocity measurements suggests that the seismic refraction results are representative of sound foundation rock, for otherwise the field measurements would be significantly less than measurements on intact core. Therefore, the field CHAPTER 02 2.5-55 REV. 13, SEPTEMBER 2006

LGS UFSAR seismic velocities should provide a reasonable quantitative index of the general character of the in situ foundation rock. Weathering reduces rock quality typically in the upper 10-30 feet of rock as indicated by reduced rock quality designation values in this zone. The weathered rock grades upward with no well defined contact into residual soil. Therefore, the properties of weathered rock are similar to those of sound rock at the base of the weathered zone and approach those of soil at the top of the zone. Because foundations for the main plant structures including the power block, radwaste, and pumphouse enclosures are carried to unweathered rock, rock weathering is not significant to foundation design for these facilities. Seismic Category I structures founded on weathered rock are buried structures including diesel oil tanks and portions of buried piping and electrical duct banks. In the foundation design of these structures, the bearing pressure allowable for dense natural soil (6,000 psf) is conservatively used, as discussed in Section 2.5.4.10.2.

Nevertheless it was recognized that occasional localized features such as fracture zones or clay seams encountered in the foundation rock would require evaluation as they became exposed.

Accordingly, foundation rock was mapped and evaluated for such features by experienced engineering geologists during the course of construction. Measures to improve foundation conditions were carried out at certain areas where potential rock weakness was encountered.

Detailed discussion of these measures is provided in Section 2.5.4.12.

2.5.4.2.2 Properties of Foundation Soils The in situ soils are residual in nature, derived from weathering of siltstone, sandstone, and shale.

The properties of these soils were determined by laboratory testing. Testing of the in situ soils at the spray pond was conducted by Geotechnical Engineers, Inc. The complete laboratory test report is given in Reference 2.5-39. The properties of the soils are given in Sections 2.5.4.2.2.1 through 2.5.4.2.2.3 and are summarized in Table 2.5-4.

The properties of the in situ soils other than spray pond were determined by Dames and Moore.

The complete laboratory test results are included in Reference 2.5-51. The laboratory testing procedures are given in Reference 2.5-141. The properties of these soils are described in Section 2.5.4.2.2.4 and are summarized in Table 2.5-4.

A compaction test was performed on a sample of the overburden soil to obtain soil compaction characteristics. Sieve tests were performed on representative soil samples to evaluate grain-size distribution. In addition, Atterberg limits were performed on the bulk soil sample used for the compaction test and on other soil samples to evaluate the plasticity characteristics. The locations of borings at the site are shown in Figure 2.5-20.

Seismic Category I structures not founded on competent bedrock include the spray pond and buried structures described below. The manner and conservatism in which soil data were used in the design of the spray pond is discussed in Section 2.5.5. Seismic Category I buried structures include portions of underground piping and electrical duct banks, valve pits, and diesel oil tanks.

These buried structures are founded above the high groundwater table and have bearing weight less than the replaced in situ soil. The static stability of the buried structures is discussed in Section 2.5.4.10.1. The seismic inertia effect of buried structures are negligible and the propagating seismic wave effects on buried piping are considered as described in Section 3.7.3.12.

2.5.4.2.2.1 Index Properties of Soils at Spray Pond The index properties include the following:

CHAPTER 02 2.5-56 REV. 13, SEPTEMBER 2006

LGS UFSAR

a. Visual and laboratory classification of samples - ASTM D2488

b. Mechanical analysis - ASTM D422

c. In situ moisture content and unit weight - ASTM D2216

d. Atterberg limits - ASTM D423 and ASTM D424

e. Specific gravity tests - ASTM D854 The in situ soil of the spray pond includes clayey silt, sandy silt, and silty fine sand, with varying amounts of gravel-sized rock fragments. The predominant soil is clayey silt, classified as ML and CL. The in situ moisture ranges from 11.9 to 38.7%, and averages 21.7%. The specific gravity ranges from 2.70 to 2.80, with an average value of 2.76.

Sieve and hydrometer analyses were performed to determine grain-size distribution according to ASTM D422. The mean grain-size (D50) was found to be in the range of 0.006 mm to 4.4 mm, with an average value of 0.32 mm.

The Atterberg Limits were obtained according to ASTM D423 and ASTM D424. The liquid limit ranges from 27 to 51, with an average value of 37. The plasticity index ranges from 2 to 27, with an average value of 15 (Figure 2.5-17).

2.5.4.2.2.2 Static Shear Strength of In Situ Soils at Spray Pond Consolidated-undrained triaxial tests were made in which undisturbed specimens were placed in the triaxial chamber and saturated by the back pressure method. After saturation the samples were consolidated isotropically to a consolidation pressure of 1 ksf. Plots are made of deviator stress vs axial strain, induced pore pressure vs axial strain, and stress paths for each test. These plots are included in appendix J of the PSAR. The failure stresses of samples are shown on Figure 2.5-18. From this drawing, the effective friction angle is determined to be 33.5.

Points plotted on Figure 2.5-18 represent conditions at failure. In general, the failure condition was chosen when the sample reached 20% strain if the shear stress continues to increase. However, if the peak was reached at a lower strain level, the (p-q) point shown on Figure 2.5-18 would represent the stresses at the peak. The point, p = 10.2 and q = 5.2, represents the Test R14 which reached its peak shear stress at 15% strain. The points are plotted in conformance with this criteria.

The undrained shear strength of the material is calculated from the effective stress friction angle and pore pressure parameter (A) at failure using the following relationship (from Reference 2.5-40):

gf Po sin _ (EQ. 2.5-1) 1 ( 2Af 1 ) sin where:

gf = undrained shear strength CHAPTER 02 2.5-57 REV. 13, SEPTEMBER 2006

LGS UFSAR

= effective stress friction angle Af = pore pressure parameter (A) at failure Po = initial mean effective principal stress For the

_ samples tested, Af varies from -0.04 to -0.38. Using the conservative

_ assumption of Af = 0 and = 33.5, the undrained shear strength is calculated to be 1.2 Po .

2.5.4.2.2.3 Dynamic Shear Strength of In Situ Soils at Spray Pond Stress controlled cyclic triaxial tests were made in which the undisturbed samples were placed in a chamber and saturated using the back pressure method. When saturated, the samples were consolidated isotropically to an effective confining pressure of 2 ksf. After consolidation, the drainage valves were closed, and a symmetrical cyclic deviator stress was applied. The axial deformation, axial load, and pore pressure were measured continuously during the cyclic loading.

The cyclic stress ratio plotted against the number of cycles required to cause 5% double amplitude strain is shown on Figure 2.5-19.

The dynamic shear strength of the soil is determined by multiplying the cyclic stress ratio by the effective overburden pressure. The cyclic stress ratio required to cause 5% double amplitude strain in five cycles is determined to be 0.61 (Figure 2.5-19). The selection of 5 cycles simulates the SSE of 0.15 g at the site, based on correlations of equivalent uniform stress cycles and time histories by Seed, et al (Reference 2.5-41).

2.5.4.2.2.4 Properties of In Situ Soils Other Than Spray Pond Area The properties of soils at the site, other than those discussed in Sections 2.5.4.2.2.1 through 2.5.4.2.2.3, are discussed in the Dames & Moore report (Reference 2.5-51). The results of the tests are described below and are summarized in Table 2.5-4.

The predominant soils at the site consist of red sandy and clayey silts with numerous rock fragments. They are classified as ML. The in situ moisture content measured in accordance with ASTM D2216-61 ranges from 8.3% to 21.3%, and averages 13.4%. The average grain-size distribution, based on the results of sieve analyses, in accordance with ASTM D422-63, were found to be 14% in gravel-size, 27% in sand-size, and 59% silt and clay. The average liquid limit and plasticity index in accordance with ASTM D422-63, based on three tests, were found to be 25 and 8, respectively. The plasticity index was calculated using the results of the liquid limit test in accordance with ASTM D423-66 and plastic limit test in accordance with ASTM D424-59.

A limited number of compression tests were performed to determine the shear strength of the in situ soil. The total shear strength parameters, based on two unconfined compression in accordance with ASTM D2166-66 and two unconsolidated undrained triaxial compression tests in accordance with ASTM D2850-70, were found to be c = 3.0 ksf, and Ø = 18. The effective shear strength, based on four consolidated drained triaxial compression tests, were found to be c 0, and Ø 26.5 . Procedures given in Reference 2.5-141 were followed in conducting the consolidated drained triaxial compression tests.

CHAPTER 02 2.5-58 REV. 13, SEPTEMBER 2006

LGS UFSAR No laboratory tests were made to determine the dynamic shear strength of the soil.

2.5.4.2.2.5 Properties of Type I Fill A description of Type I fill and methods of placement are given in Section 2.5.4.5.4. The static or dynamic properties of this material have not been measured.

A total unit weight of 140 pcf and an at rest coefficient of lateral earth pressure of 0.7 have been assumed for Type I fill. The Type I fill is expected to have properties similar to dense sand and gravel and to have a maximum shear-wave modulus determined by the following expression (Reference 2.5-144):

G = 1000 K2 m 1/2 where:

G = maximum shear modulus, psf, at a strain of 10-6 in/in K2 = 140 m = average stress, psf, and is equal to 1

2 3 3

1 = H 3 = 0.7 H

= Unit weight, 140 pcf H = depth, ft 2.5.4.3 Exploration The locations of all field explorations are shown in Figures 2.5-20 and 2.5-21. Summary logs of borings are shown in Figure 2.5-22. Soils are classified in accordance with the Unified Soil Classification System. Rock coring was performed with double-tube, NX equipment.

Site drilling began in September and October, 1969 and continued in the spring of 1970 (Borings 1 through 301). Geophysical surveys (Section 2.5.4.4) were performed in the plant area at this time.

In the spring of 1971, additional drilling was performed in the area of the Perkiomen Creek pump station. In late 1971 and early 1972, additional drilling was performed in the Schuylkill River pump station area. In 1973 auger holes were drilled in the area of the emergency spray pond, and in 1974 detailed spray pond investigations were completed. The cooling tower foundation exploration was also completed at this time. Most of the main plant foundation was exposed at this time, and some fracture zones were noted (Sections 2.5.1.2.5 and 2.5.4.12). Between April and July, 1974, an extensive geologic study of these zones and the area around the site was completed by Dames and Moore. This study consists of geologic mapping, with some drilling and test trenching near the Sanatoga, Brooke-Evans, and Linfield faults. Site exploration was completed in the fall of 1974. A total of 366 borings and four test pits was completed in the site area for geologic, groundwater, and foundation investigations. In March, 1977, 13 additional test pits were excavated to depths of from 5-11.5 feet, to investigate soils for use in construction of the soil-bentonite liner for the emergency spray pond.

CHAPTER 02 2.5-59 REV. 13, SEPTEMBER 2006

LGS UFSAR Static water levels were measured in some of the borings drilled on the site. Perforated plastic pipes were installed in some borings to allow future collection of water level data. These borings are denoted on the plot plan (Figure 2.5-20).

A geologic map of the main power block foundations is presented in Figure 2.5-13. Contours on top of rock in the site area are shown on Figure 2.5-9. Geologic profiles are shown on Figures 2.5-10 and 2.5-11.

2.5.4.4 Geophysical Surveys 2.5.4.4.1 Seismic Refraction Survey Seismic refraction surveys were conducted at the site. A continuous series of seismic lines closely followed the continuous line of borings drilled across the site (Figure 2.5-21). These lines ranged in length from about 400-700 feet, with some overlapping coverage. Two 550 foot refraction lines were performed perpendicular to the major line that traversed the site.

The results of the seismic refraction surveys were used to develop dynamic properties of the foundation materials. Permanent records of the compressional waves generated from the survey were obtained using an Electro-Technical Labs ER75012 seismic timer, which is a 12-trace refraction seismograph. Geophone spacings of 25 feet and 50 feet were used in performing the surveys. Compressional wave velocities (Vp) in rock measured during these studies range from about 7000 fps to 20,000 fps, with an average of approximately 12,500 fps.

2.5.4.4.2 Shear-Wave Velocity Survey A shear-wave velocity survey was performed to further evaluate dynamic bedrock characteristics.

Shear-wave velocities were computed from the records obtained using two Sprengnether 3-component engineering seismographs. Observations were made at distances of between 500 feet and 1000 feet from the shot point, at the locations indicated on Figure 2.5-21. An apparent shear-wave velocity (Vs) of 6100 ft/sec was derived from the line run perpendicular to the rock strike, roughly in a north-south direction. In the second line, which parallels the approximate east-west strike of the bedding-plane, an apparent shear-wave velocity of 5800 ft/sec was measured. Using elastic theory, a Poisson's ratio of about 0.3 can be calculated using these shear-wave velocities and average compressional wave data across the site. Shear-wave data are included in Table 2.5-11.

2.5.4.4.3 Up-Hole Velocity Survey An up-hole velocity survey was performed in the pump test well at the location shown on Figure 2.5-21. Measurements were obtained using a Porta-Seis refraction seismograph, with small explosive charges as energy sources, and an in-hole cable with 25 foot geophone spacings.

Repeated shots were made and the cable was withdrawn in 5 foot increments.

The up-hole velocity survey was made to determine variations in the vertical compressional wave velocity of the underlying rock. Compressional wave velocities (Vp) measured were 7700 fps from depth 110-140 feet, and 12,600 fps from depths37-110 feet and from 140-187 feet.

2.5.4.4.4 Micromotion Measurements Measurements of the ambient background motion of the site and its response to natural motion generators, such as wind and tides, give an index of the dynamic properties of the materials CHAPTER 02 2.5-60 REV. 13, SEPTEMBER 2006

LGS UFSAR underlying the site. An attempt was made to measure these motions in the proposed plant area using the Dames and Moore microtremor equipment. This equipment is a highly sensitive, electronic, vibration recording device capable of magnification up to 150,000 times, and it is accurate over a frequency range of about 0.4-30 cycles per second. The 3 component micromotion observation was made on the existing ground surface (prior to construction), near the test well used for the up-hole velocity survey, as shown on Figure 2.5-21.

The microtremor records indicate that ambient motions caused by natural background excitations at the site are negligible. There is a suggestion of motion at a period of 0.7-0.8 second, but no apparent predominant ground period; this is a condition common at sites where competent materials are exposed at the surface.

2.5.4.5 Excavations and Backfill The location, limits of excavation, and type of fills associated with seismic Category I facilities are presented in Figure 2.5-37. Detailed descriptive data concerning these backfills and excavations are discussed below. Procedural controls, such as inspections and tests, exercised during the preparation of foundations for seismic Category I structures are described below and in Section 2.5.5.

The properties of foundation rock at the site (Section 2.5.4.1 and Tables 2.5-3 and 2.5-11) indicate that the rock has adequate strength to resist significant heave at the depths excavated. Further, the moderately well to well developed conjugate sets of vertical joints present at the site, together with subhorizontal bedding-planes, would presumably have provided for prior release of any excessively high in situ stresses, minimizing the potential for rock rebound.

Seismic Category I facilities on soil, such as pipelines and electrical ducts, are largely buried and backfilled facilities, such that the accompanying loads approximate the original in situ conditions.

The diesel oil storage tank facility, which is founded on weathered rock, is also a buried and backfilled facility. At the spray pond, the pump house, overflow structure and all pipe supports are to be founded on rock; as discussed in Section 2.5.5.2, factors of safety on the seismic Category I soil slopes are considerably in excess of the minimum acceptable requirements.

In view of these circumstances, specific measures for monitoring foundation rebound and heave were considered but are not considered to be necessary. No instances of foundation rebound or heave were noted during excavation or construction of the facilities.

2.5.4.5.1 Main Power Block and Cooling Tower Excavation All seismic Category I rock foundations at the main power block are carried to, or well below, unweathered bedrock. These include the reactor enclosures and control room structure. Rock foundations for the turbine and radwaste enclosures, although they are not seismic Category I structures, are prepared according to the same general procedures and criteria used in preparing the seismic Category I rock foundations.

The excavation of rock proceeded by initial ripping of any weathered surficial rock material, followed where necessary by line blasting and presplitting in holes drilled to provide stable slopes.

Essentially vertical slopes in unweathered rock proved stable throughout the duration of construction, and no special protective measures were required.

CHAPTER 02 2.5-61 REV. 13, SEPTEMBER 2006

LGS UFSAR The surface of the excavated foundation rock is scaled to remove loose debris and jetted with water or air to remove loose fragments and to prepare the surface for concrete. Before placement of structural concrete or concrete backfill to design elevation, all seismic Category I foundations for the main power block are inspected by an engineering geologist to verify the suitability of the rock and its proper surface preparation to receive concrete. Geologic mapping of foundation rock in the main power block area is presented in Figure 2.5-13.

Foundations for each of the cooling towers (nonseismic Category I structures) consist of 40 individual pedestals supporting the columns and extending to bedrock. Excavation proceeded by cutting a ring trench and preparing a suitable surface for each pedestal in unweathered or partly weathered bedrock by ripping or blasting as necessary, followed by scaling and jetting.

2.5.4.5.2 Diesel Generator Enclosure Excavation The plans and sections of the diesel generator enclosure foundations are shown in Figure 3.8-61.

The diesel generator enclosures are founded on bedrock; the geological inspection and preparation of the foundation rock are the same as for the main power block.

The spray pond pump structure and spray network pipe supports are seismic Category I and are founded on bedrock. Geological inspection and preparation of the foundation rock for these structures are the same as that for the main power block.

2.5.4.5.3 Spray Pond Excavation, Slope Protection and Liner Construction The plans and sections of the spray pond showing excavation, slopes, and normal pond level are given in Figures 3.8-56 and 3.8-57. As shown, the spray pond is constructed primarily by excavation. A soil-bentonite liner and a protective soil cover are placed over the entire bottom of the pond and on the soil slopes. The soil cover on the slopes in turn is protected by riprap and riprap bedding. The rock slopes are treated by shotcrete for protection against weathering. Due to a sloping bedrock surface, the bottom of the pond below the soil-bentonite liner is underlain partially by soil and partially by rock. Slopes in soil are excavated at 4 horizontal to 1 vertical; rock slopes are excavated at 1 horizontal to 1 vertical (Section 2.5.5), except for a short transition zone in weathered rock, which is excavated at 2:1.

The soil-bentonite liner, soil cover, riprap bedding and foundation preparation of the spray pond are discussed in detail in Section 2.5.5.

2.5.4.5.4 Fills - General All fills placed at the site, which are associated with seismic Category I structures (except spray pond), include random fill, select granular backfill, cementitious backfill, and concrete.

Two types of random fill, identified as Type I and Type II, were used for general site grading.

Type I fill was placed in the areas adjacent to the main power block and as backfill over part of the RHR and ESW piping. Type II fill was used for finished grading in the outlying areas, including backfill over part of the safety-related duct banks.

The division line between cut and fill and the limits of Type I and Type II fill within the main power block complex are shown in Figure 2.5-37.

CHAPTER 02 2.5-62 REV. 13, SEPTEMBER 2006

LGS UFSAR Type I fill consisted of broken rocks and fines obtained from the site excavations and was graded from fine to coarse material with no rock fragments larger than 8 inches in diameter. It was placed in uniform layers of loose lifts with a maximum thickness of 12 inches and was compacted and tested as described below.

Type I fill placed before July 1971 was compacted to 90% of the maximum dry density in accordance with AASHO T180-61, Method D, and was tested in accordance with AASHO T147-54. Type I fill placed thereafter was compacted to 90% of the maximum dry density in accordance with AASHO T180-70, Method D, and was tested in accordance with AASHO T191-61.

Type I fill was compacted by the following approved types of equipment:

a. Sheep-foot roller with minimum weight of 4000 pound per linear foot of drum, operated at a speed of approximately 3 mph.

b. Rubber tire rollers with minimum of 4 pneumatic tire wheels maintaining tire pressure on the ground of between 80-100 psi. The load per wheel may vary from 18,000-25,000 pounds.

c. Vibratory roller with minimum weight of 21,000 pounds, drum diameter approximately 70 inches and length of 78 inches, and minimum centrifugal force of 40,000 pounds, and frequency 1000-1400 vibrations per minute.

d. In confined areas, hand operated equipment was used with a maximum lift of 8 inches.

e. A total of 379 compaction tests were performed before the end of 1982. The minimum frequency for testing Type I fill for compaction was not less than once for each 500 cubic yards placed. Fills failing to meet the compaction requirements were removed, recompacted, and retested.

Type II fill consisted of broken rocks and fines obtained from the site excavations. It was placed uniformly in unconsolidated lifts not exceeding 24 inches in thickness, producing a reasonably well graded mass with a minimum of stratification of fine or coarse materials. The material was uniformly spread over the entire area by bulldozer prior to compaction. The material was moisture-conditioned to attain satisfactory compaction. Type II fill was compacted by the equipment described for Type I fill under general supervision without testing requirements. Granular Type II fill was also compacted by use of a track-type tractor weighing not less than 60,000 pounds and making a minimum of four passes overlapping one-fourth the width of the track on each pass.

Select granular backfill consisted of imported aggregate or screenings. The maximum particle size was 3/4 inch, with no more than 10% by weight passing the No. 200 sieve. Select granular backfill was placed in loose lifts with a maximum thickness of 6 inches and was compacted to 95% of the maximum dry density in accordance with AASHO T180-70, Method D. As an alternative to the above material, when cohesionless select backfill was used, it was compacted to 90% of the maximum dry density in accordance with ASTM D2049-69 by the use of both dry and wet methods.

In-place select granular backfill was tested in accordance with AASHO T191-61.

CHAPTER 02 2.5-63 REV. 13, SEPTEMBER 2006

LGS UFSAR The cementitious backfill consisted of a mixture of portland- cement, aggregate, and water. The minimum compressive strength at 28 days was 80 psi. Testing for compressive strength of cementitious backfill using sand as aggregate was in accordance with ASTM C109, ASTM C31, and ASTM C39. Testing for compressive strength of the backfill using coarse aggregate was in accordance with ASTM C31 and ASTM C39.

Slump tests were performed in accordance with ASTM C143. The cementitious backfill was consolidated by use of mechanical vibrating equipment for proper placement.

The concrete backfill consisted of a mixture of portland-cement, aggregates, admixtures, and water. The minimum 28 day compressive strength was 2,000 psi. The standards and specifications that governed the concrete backfill are stated in the following sections:

a. Section 3.8.6.1.2.2 Mix proportioning

b. Section 3.8.6.1.4.2 Mixing and delivery

c. Section 3.8.6.1.4.3 Placing

d. Section 3.8.6.1.4.4 Consolidation

e. Section 3.8.6.1.5 Construction testing Field tests of Type I random fill, select granular backfill, cementitious backfill and concrete backfill used in conjunction with seismic Category I structures, electrical duct banks, manholes, pipelines and valve pits were performed by Quality Control. The test data are available at the site for inspection and review.

2.5.4.5.5 Miscellaneous Category I Facilities - Excavation and Backfill Seismic Category I facilities not founded on unweathered bedrock include part of the spray pond, portions of the underground piping and electrical ducts, oil tanks, and valve pits. The spray pond is discussed separately in Section 2.5.4.5.3. Portions of these Category I facilities are founded on weathered rock, natural soil, or fills. The fills are discussed in Section 2.5.4.5.4.

Underground piping was installed in trenches excavated to a minimum of 6 inches below the pipe.

Soft spots and unsuitable material found at the bottom of the trenches were removed and replaced with select granular backfill, cementitious backfill, or concrete. Select granular backfill, cementitious backfill, or concrete was placed at least 6 inches below and on each side of the pipe to a minimum of 12 inches above the pipe. The remainder of the trench was backfilled with Type I fill, select granular backfill, cementitious backfill, or concrete. All Category I piping was buried with adequate cover for missile protection.

The plans, profiles, and sections showing the detailed relationship of the Category I piping to subsurface soil, fill, and rock materials are shown on Figure 2.5-37.

The diesel oil tanks and Category I electrical ducts were buried with adequate cover for missile protection. The Category I valve pits were buried, except the roof slabs, which are missile and tornado resistent and are exposed above ground. Soft spots and unsuitable material found at the bottom of the excavations for these structures were removed and replaced with cementitious CHAPTER 02 2.5-64 REV. 13, SEPTEMBER 2006

LGS UFSAR backfill or concrete. Where overexcavation occurred below these structures, the select granular, cementitious or concrete backfilling materials were used. Cementitious backfill was placed at least 2 feet below and on each side of the diesel oil tanks, to a minimum of 12 inches above the tanks.

The remaining backfill to finish grade was placed using select granular backfill. The sides of the valve pits were backfilled with cementitious backfill or Type I fills. The electrical duct banks were completely encased in concrete with a minimum of 3 inches of concrete cover on all four sides.

The remaining trench excavation was backfilled to finish grade with Type I or Type II fills. Section 3.8.4.1.6 contains additional discussions on these miscellaneous structures. The plans, profiles, and sections showing the detailed relationship of the Category I electrical duct banks to subsurface soil, fill, and rock materials are shown on Figure 2.5-37.

2.5.4.6 Groundwater Conditions A detailed groundwater study of the site is presented in Section 2.4.13. Groundwater occurs at the site in the Brunswick lithofacies which consist of bedded siltstone, sandstone, and shale.

Groundwater flows primarily through joints, fractures, and other secondary openings in the consolidated rock. The water table is 15-95 feet below land surface at the site. A map of the potentiometric surface, determined from water levels measured in May, 1979, indicate the groundwater levels range from el 250' east of the spray pond to el 120' southwest of the radwaste enclosure. Fluctuation of water levels in observation wells are indicated by the hydrographs in Figure 2.4-18.

Groundwater studies conducted for the spray pond include installation of permanent observation wells and the performance of 41 permeability tests. Permeability values obtained from the field tests at the spray pond are given in Table 2.4-18. The average permeabilities for various materials are as follows:

Material Permeability (ft/yr)

Overburden 3.5 Contact Zone 14.0 Rock 214 2.5.4.6.1 Spray Pond Seepage Analysis The spray pond makeup system has sufficient capacity to replace estimated seepage losses during normal operation. Moreover, the total volume of water in the pond itself is sufficient to accommodate estimated seepage during the 30 day transient period throughout which no makeup to the pond is assumed to be available.

Seepage losses from the pond migrate toward the Schuylkill River and to the north, as shown on Figure 2.5-23. As detailed below, estimated seepage losses from an unlined spray pond indicate that seepage from the pond would not adversely affect the safety and performance of the ultimate heat sink, nor significantly effect groundwater levels beneath the site. Nevertheless, a liner is provided.

Estimated seepage losses from an unlined pond were calculated by subtracting the natural, preconstruction groundwater underflow from the total underflow expected after the spray pond is constructed. Flows toward two discharge areas were analyzed separately because of the difference in differential heads. Two methods were used to calculate total underflow using Darcy's CHAPTER 02 2.5-65 REV. 13, SEPTEMBER 2006

LGS UFSAR Law: construction and analysis of a flow net, and computation of underflow through a peripheral cross-section.

In the first analysis, a flow net method was used in which a plan flow net was constructed as shown on Figure 2.5-23. Total flow was calculated using the equation (from Reference 2.5-42):

Q= ns KDH nd (EQ. 2.5-2) where:

Q = quantity of underflow, ft3/yr ns = number of stream tubes nd = number of equipotential drops K = permeability, ft/yr H = differential head, ft D = aquifer thickness, ft The differential head between the spray pond surface and the Schuylkill River is 141 feet. The differential head between the spray pond surface and the northern discharge area at el 200' is 51 feet. An effective aquifer thickness of 140 feet is used because of the decrease in number and size of fractures at that approximate depth as observed in the core holes. A permeability of 200 ft/yr is used as an effective value for the residual soils and bedrock materials (Section 2.4.13.2.5).

Underflows were determined to be 5.3x106 ft3/yr towards the Schuylkill River, and 1.6x106 ft3/yr toward the north, giving a total underflow of 6.9x106 ft3/yr.

The second method of analysis, a cross-sectional area method, uses the following form of Darcy's Law:

Q = KIA (EQ. 2.5-3) where:

Q = quantity of underflow, ft3/yr K = permeability, ft/yr I = hydraulic gradient (ratio)

A = cross-sectional area of underflow, ft2 Using an aquifer thickness of 140 feet, the cross-sectional area through which both natural underflow and seepage from the pond is flowing toward the Schuylkill River is approximately 224,000 ft2; the cross-sectional area through which water is flowing toward the north is approximately 168,000 ft2. The hydraulic gradient is approximately 0.1 toward the Schuylkill River CHAPTER 02 2.5-66 REV. 13, SEPTEMBER 2006

LGS UFSAR and 0.05 toward the north. The permeability is 200 ft/yr. The rates of underflow determined by this method are 4.5x106 ft3/yr toward the Schuylkill River and 1.7x106 ft3/yr toward the north, giving a total underflow of 6.2x106 ft3/yr.

Preconstruction natural underflow was calculated using equation 2.5-3. The hydraulic gradient (I) was determined from equipotential contours of the groundwater table measured on June 24, 1974, shown on Figure 2.5-23. The hydraulic gradient is 0.08 for flow toward the Schuylkill River, and 0.02 for flow toward the northern discharge area. The cross-sectional areas of natural underflow (A) are based on a saturated aquifer thickness of 110 feet. The permeability (K) is 200 ft/yr, as described above. Based on these parameters, natural underflow beneath the pond is estimated to be 2.74x106 ft3/yr toward the Schuylkill River and 0.54x106 ft3/yr toward the north. Total preconstruction (natural) underflow, then, is estimated to be 3.3x106 ft3/yr, the sum of these flows.

Therefore, the estimated spray pond seepage loss from an unlined pond is:

(6.9x106) - (3.3x106) = 3.6x106 ft3/yr (Flow net method) or (6.2x106) - (3.3x106) = 2.9x106 ft3/yr (Cross-sectional area method)

These calculated losses would cause a decline of 0.6-0.7 feet per month in the water level of an unlined spray pond.

In both methods of analysis, the average permeability is assumed to be approximately that of rock (200 ft/yr). Because only 60% of the pond bottom is exposed to rock, and the balance is exposed to residual soils of markedly lower permeabilities, these estimates of total seepage loss from an unlined pond are probably high.

The permeability of the materials exposed within the pond is an important parameter in the seepage analysis. Bedrock, the most permeable material underlying the pond, underlies approximately 60% of the pond. Most of this rock is now covered with concrete which provides foundation and support for the spray pond piping network.

The balance of pond is underlain by overburden and the bedrock-overburden contact zone. As presented in Table 2.4-18, the 33 measured permeabilities of the bedrock (Brunswick Formation) range from 1 ft/yr to 1247 ft/yr, of which the arithmetic mean is 214 ft/yr. The location of wells where tests were performed is shown on Figure 2.4-15. The median value is 71 ft/yr, and fully 90%

of the measured permeabilities in rock are less than 500 ft/yr. Only two of the measured permeabilities are more than 1000 ft/yr. The measured permeabilities from eight tests in the overburden and contact zone range from less than 1 ft/yr to 21 ft/yr (Table 2.4-18). The mean value is 7.5 ft/yr and the median is 3 ft/yr.

Section 2.4.13.1.1 points out that most groundwater flow occurs through secondary openings in the rock (fractures and joint), and the higher permeability measurements are attributed to test performed in relatively more fractured zones that are of limited extent. The 200 ft/yr used in the seepage analysis is quite conservative because: 1) it is approximately equal to the arithmetic mean of all rock permeability measurements even though this average is biased by the inclusion of the abnormally high (greater than 1000 ft/yr) permeability measurements; 2) almost half (40%) of the spray pond area is underlain by overburden or contact zone material that has a significantly lower permeability than the bedrock (average value of 7.5 ft/yr). More realistic, although still CHAPTER 02 2.5-67 REV. 13, SEPTEMBER 2006

LGS UFSAR conservative, estimates of the average effective permeability of the spray pond area might be 132 ft/yr (60% at 214 ft/yr and 40% at 7.5 ft/yr), or 173 ft/yr (the average of all permeability tests).

The spray pond includes a soil-bentonite liner on the bottom and on soil slopes, and shotcrete on rock slopes. The soil-bentonite liner is one foot thick and has a permeability of less than one ft/yr (Figure 2.5-24 and Section 2.5.5.4). The seepage loss for the lined pond is calculated to be 1.83x106 gal/month or 2.94x106 ft3/yr (Section 9.2.6.4). The liner ensures that the actual seepage loss is acceptable by preventing higher rates of seepage through localized fracture zones in the event such conditions were found to exist.

A seepage test was performed to ensure that the design basis seepage rate assumptions were not exceeded. The test results show that, based on weekly measurements of seepage losses over a 23 week period, the average seepage loss for the entire period is between 11% and 21% of the allowable design value, depending on how evaporation losses are measured. This is equivalent to a seepage rate of about 5-9 gpm for the whole pond. Considering possible error contributions from all measurements, the upper limit of this seepage loss is estimated to be approximately 40% of allowable design value, whereas the lower limit is effectively zero. Therefore, the actual seepage loss from the spray pond is well below the allowable design value. Additional information is contained in Section 2.5.5.4.3 and Reference 2.5-145.

2.5.4.6.2 Dewatering During Construction Groundwater presented no problem during excavation and construction in the main power block area. The water table was below most of the excavation during construction. The rock around the power block has low permeability and did not transmit significant quantities of water into the excavation. Only small amounts of seepage occurred along the walls of the radwaste enclosure excavation during construction.

This same low permeability of foundation rock caused surface run-off to collect in the foundation excavations, especially during periods of construction inactivity. For example, the rock in the excavation was more or less continuously covered with water from early 1972 to late August. In March 1972, water was standing in the excavation to about el 170', inundating the radwaste foundation area and lower parts of the reactor foundations. On July 6, 1972, rain from Hurricane Agnes raised the water level in the excavations to about el 175'; 22 days later, the water level had not noticeably declined. The run-off water was pumped out of the excavation following resumption of construction activities. These conditions further indicate the low permeability of the foundation rock.

A subdrainage system was installed for dewatering of the main power block region during construction. This system will remain permanent following construction to minimize water table height, thereby minimizing hydrostatic loading on subsurface walls. Any lowering of groundwater elevation by the subdrainage system has not been taken into consideration for design of subsurface portions of power block structures. The maximum expected water table elevations, defined in Section 2.4.13.5, were used for design.

2.5.4.6.3 Groundwater Monitoring The water level in observation wells at the spray pond and power block area were monitored as discussed in Sections 2.4.13.2.4 and 2.4.13.4. The data obtained were used in establishing the potentiometric surface and the direction of groundwater flow. This water level monitoring, CHAPTER 02 2.5-68 REV. 13, SEPTEMBER 2006

LGS UFSAR throughout plant construction and site grading/surfacing, has indicated no changes in the groundwater flow direction. Changes in the potentiometric surface are within the plant design criteria for hydrostatic loading. Hence, the hydrologic condition is expected to remain within the design criteria during the operating life of the plant.

2.5.4.7 Response of Soil and Rock to Dynamic Loading The responses of soil and rock to dynamic and seismic loading conditions are discussed in Section 2.5.2. Further discussion of response characteristics of soil at the spray pond site is contained in Sections 2.5.4.8 and 2.5.5.2. Soil-structure interaction considerations are discussed in Sections 3.7.1.4 and 3.7.2.4.

2.5.4.7.1 Response of Soil Along Pipelines A soil response study was performed to determine the characteristics of ground motion induced by the SSE. This ground motion is required to evaluate the response of buried seismic Category I piping.

The stratigraphic profile along seismic Category I piping is shown on Figure 2.5-37. The pipe support conditions vary from rock to in situ soil to as much as 24 ft of Type I fill beneath the pipe.

The soil stratigraphy and properties used in the analyses are shown in Table 2.5-10. The assumed properties for Type I fill are given in Section 2.5.4.2.2.5. The shear modulus of the rock is presented in Table 2.5-11.

Soil response studies were performed using the computer program SHAKE, Reference 2.5-143.

The design SSE time history with a peak acceleration of 0.159 (Section 3.7.1.2) was input at the surface of the rock. The shear modulus and damping factors were varied with strain as shown on Figures 2.5-40 and 2.5-41, in general accordance with Reference 2.5-144. The results of the analyses along with solutions for higher and lower moduli for Type I fill are given in Table 2.5-10.

The results presented include peak acceleration at the level of the pipe, peak particle velocity at the level of the pipe, and natural frequency of the soil column.

2.5.4.8 Liquefaction Potential The soil at the seismic Category I spray pond was analyzed for liquefaction potential. The soils at other seismic Category I facilities were not analyzed since these soils are not saturated and the potential for becoming saturated is negligible.

The liquefaction potential of soil in the spray pond site was analyzed for a maximum ground acceleration of 0.15 g. Because of the shallow depth of soil, the maximum induced shear stress was calculated assuming that the soil mass behaves as a rigid body, and the average equivalent shear stress was taken as 65% of the maximum induced shear stress (Reference 2.5-43).

The dynamic strength of soil was determined from the cyclic triaxial test results included in appendix J of the PSAR and is shown graphically in Figure 2.5-19. The equivalent number of uniform stress cycles is taken as five.

The soil profile below the pond bottom used in the analysis represents the most critical section.

The soil profile includes a 12 inch layer of protective soil cover and a 12 inch layer of soil-bentonite CHAPTER 02 2.5-69 REV. 13, SEPTEMBER 2006

LGS UFSAR liner on top of 9 feet of in situ soil. The pond bottom and the bedrock are at el 241' and el 230',

respectively.

The average induced shear stress was calculated as follows:

ave = 0.65 h amax d (EQ. 2.5-4) g where:

ave = average induced shear stress

= saturated unit weight of the material. The values used in the analysis were 123.8 pcf, 119.0 pcf, and 126.4 pcf for the soil cover, soil-bentonite liner, and the in situ soil respectively (Table 2.5-5) h = depth where the induced stress is to be computed amax = 0.15 g

d = correction factor (0.98 to 0.99 for shallow soil profile)

The shear strength was calculated based on the results of cyclic triaxial shear tests and equals 0.37o, which was obtained by multiplying the design cyclic stress ratio of 0.61 (Figure 2.5-19), the effective overburden pressure o, and a correction factor of 0.60 (Reference 2.5-43).

The factor of safety was obtained by dividing the shear strength by the average induced shear stress. Since both the shear strength of the soils and the induced shear stresses are dependent on depth below ground surface, determinations of the factor of safety against liquefaction were made at various depths. The results of this analysis are summarized in Figure 2.5-25. The minimum factor of safety was computed to be 1.9.

2.5.4.9 Earthquake Design Basis Derivation of the OBE and SSE are discussed in Section 2.5.2. The liquefaction potential and slope stability of the spray pond are analyzed for the SSE event.

2.5.4.10 Static Stability The reactor enclosures, control structure, diesel generator enclosure, spray pond pump house, spray networks, turbine enclosures, and radwaste enclosure are founded on sound, unweathered bedrock. Seismic Category I facilities not founded completely on unweathered bedrock include the spray pond, underground piping, diesel oil tanks, valve pits and electrical ducts. Portions of these facilities not founded on rock are founded on natural soil and/or manmade fills.

The strength of the unweathered bedrock amply accommodates the loads of the plant, providing highly stable foundation conditions. As measured by seismic refraction surveys in the area of the principal plant structures, compressional wave velocities range from 7000-20,000 fps, averaging about 12,500 fps; shear-wave velocities range between 5800 fps and 6100 fps. An up-hole survey CHAPTER 02 2.5-70 REV. 13, SEPTEMBER 2006

LGS UFSAR (Figure 2.5-21) measured a compressional wave velocity of 12,600 fps in the siltstone beneath the site. Unconfined compression test results on rock core samples (Table 2.5-3) range from 460-1760 tons/ft2, with an average of about 1140 tons/ft2. Poisson's ratio is calculated to be about 0.3 (Table 2.5-11). Static moduli derived from additional compression tests on rock cores range from 1.2x106 psi to 8.3x106 psi, averaging 4.1x106 psi; compressive strengths range from about 580-2370 tons/ft2, averaging 1230 tons/ft2 (Table 2.5-12). Unconfined compressive strength and elastic modulus determinations substantially followed ASTM D2938-71 and ASTM D3148-72, respectively, which were adopted after these tests were completed.

Plate bearing tests (ASTM D1194-57) were run by Dames and Moore at the site (Reference 2.5-50); the results are quite variable. Values of the Secant Modulus of Deformation at first loading, which includes plastic and elastic deformation and also reflects the closing of joints and fractures, ranges from 30,000 psi to 200,000 psi, with an average of 85,000 psi. The Secant Modulus of Elasticity at second loading is much higher, with an average value of 356,667 psi.

A bearing capacity of 30 tons/ft2 (60 ksf) for static and frequently applied live loads on sound rock is used for design, following recommendations by Dames and Moore (Reference 2.5-51). Actual loads induced by the plant structures founded on bedrock are much less than the allowable bearing pressure of the foundation rock, and they are far below the ultimate bearing capacity. The structural loads produce no significant total or differential settlement of the foundations.

The design lateral earth pressure acting on subsurface walls of seismic Category I structures was computed assuming granular backfill having the properties stated in Section 2.5.4.5.4. The coefficient of earth pressure "at-rest" was used. In addition, the walls were designed for surcharge loadings and dynamic soil pressures as appropriate. The typical pressure diagrams and combinations are shown on Figure 2.5-39.

2.5.4.10.1 Static Stability of Safety-Related Structures on Rock The following sections contain information regarding static and dynamic lateral earth pressures and groundwater loads on the reactor enclosure, control structure, diesel generator enclosure including pipe tunnel, and spray pond pumphouse, which are all founded on bedrock. Table 2.5-9 includes safety-related structures, dimensions of foundations, approximate bearing elevation, design bearing pressure, and hydrostatic pressure. Seismic Category I structures not founded on rock are discussed in Section 2.5.4.10.2.

2.5.4.10.1.1 Diesel Generator Enclosure Including Pipe Tunnel The exterior and interior foundation walls of the diesel generator enclosure are founded on bedrock (Figure 3.8-61). The interior walls support the base slab at el 217'. The space between the bedrock and the bottom of the base slab is backfilled with fillcrete. Concrete backfill surrounds all subsurface walls and extends to the rock profile such that there will be no transmissibility of lateral pressures to the walls.

The pipe tunnel is a concrete box section with the base slab founded on bedrock. The north wall lies parallel to the adjacent reactor enclosure wall and is separated by a 1 inch seismic gap. The west and east tunnel walls are separated by 1 inch seismic gaps from the adjacent radwaste enclosure and auxiliary boiler enclosure. The south tunnel wall was designed as a restrained retaining wall to resist the lateral earth pressure due to backfill which has a saturated unit weight of CHAPTER 02 2.5-71 REV. 13, SEPTEMBER 2006

LGS UFSAR 140 pcf, an at rest earth pressure coefficient of 0.7, a surcharge of 250 psf due to AASHO H-20 truck loading, and a dynamic lateral force due to seismic loading as shown on Figure 2.5-39.

The box section was also designed for the lateral earth pressure resulting from the Cooper E-72 railroad loading, the live load on the roof, and the seismic load.

Because the high-water table elevation is below that of the foundation in the power block region, there are no groundwater loads acting on the foundations of the diesel generator enclosure.

2.5.4.10.1.2 Reactor Enclosure and Control Structure The reactor enclosure and control structure are separated from the surrounding Type I fill by adjacent structures or pipe tunnels.

The reactor enclosure and control structure are bounded to the north and west by two nonseismic Category I structures (the turbine enclosure and the radwaste enclosure). The south and east sides of the reactor enclosure and control structure are bounded by pipe tunnels. The adjacent structures and pipe tunnels, founded on competent bedrock, extend up to or above plant grade el

+/-217'. Those subsurface exterior walls of the reactor enclosure and control structure, lower than the foundation grade of adjacent structures, are placed adjacent to excavated rock slope with Class A concrete or fillcrete backfilled between the face of the wall and the face of the rock slope.

Because of the conditions described above, lateral earth pressure is not considered in the design of the exterior walls.

The walls of the control structure and reactor enclosure have been designed for a hydrostatic pressure up to el 195', which is the expected maximum water table elevation in this region.

2.5.4.10.1.3 Spray Pond Pumphouse The foundation mat and walls of the spray pond pumphouse are founded on bedrock (Figure 3.8-62). Exterior walls along the east, west, and south sides are placed adjacent to excavated rock slopes with Class A concrete or fillcrete backfilled between the face of the wall and the face of the rock slope. The exterior wall along the north side extends to the bottom of the spray pond with no embedment. Because of the conditions discussed above, lateral earth pressure is not considered in the design of the exterior walls.

The north wall of the water pit area has been designed to resist hydrostatic pressure (from el 236' to el 267') and lateral seismic loads. The foundation mat has been designed for the same hydrostatic pressure as the north wall in combination with other concurrent loads.

2.5.4.10.1.4 Diesel Oil Tanks These tanks are located on a base slab founded on weathered rock or cementitious backfill bearing on weathered rock as shown in Figure 2.5-37. The associated valve pits located on top of the tanks are founded on cementitious backfill, which also acts as backfill around the tanks. The cementitious backfill has a minimum compressive strength of 80 psi. The remaining portion is backfilled with select granular fill placed and compacted as discussed in Section 2.5.4.5.4.

Railroad tracks were located on top of select granular fill. The base slab bearing pressure is 5410 psf, which includes all dead and live loads and the Cooper E-80 railroad loading. The bearing CHAPTER 02 2.5-72 REV. 13, SEPTEMBER 2006

LGS UFSAR pressure on the foundation of the valve pits, including dead loads and AASHO H-20 truck loading, is 2830 psf. The allowable bearing capacities of the rock and cementitious backfill are not exceeded by these pressures.

The walls of the valve pits are designed to resist lateral loads due to backfill having a saturated unit weight of 140 pcf and an at rest earth pressure coefficient of 0.7, AASHO H-20 truck surcharge, and dynamic lateral loading due to a seismic event. The roofs of the valve pits are adequately designed to resist AASHO H-20 truck loading and tornado depressurization or missile impact.

Because the tanks are founded on bedrock or dense natural soil, the amount of settlement is considered to be insignificant.

As an additional protection against flotation, the tanks are adequately tied to the base slab by holddown straps. For this purpose, the tanks are assumed to be submerged completely in water.

2.5.4.10.2 Static Stability of Safety-Related Structures on Soil The following sections discuss seismic Category I facilities not founded completely on unweathered bedrock. There are no groundwater loads acting on the foundations of the spray pond, underground piping, diesel oil tanks, valve pits, or electrical ducts because the high water table elevation is below these foundations. Table 2.5-9 includes safety-related structures, dimensions of foundations, approximate bearing elevation, design bearing pressure, and hydrostatic pressure.

2.5.4.10.2.1 Spray Pond The sustained load from the spray pond is less than the weight of overburden removed; therefore, there is an adequate factor of safety against overstressing the underlying soil (Figures 3.8-55, 3.8-56, and 3.8-57). Soil rebound during excavation for the spray pond is insignificant. Section 2.5.5 contains a discussion of slope stability under static and seismic conditions, including the design parameters and test results of soil exploration.

2.5.4.10.2.2 Underground Piping The method used for installation of underground piping is discussed in Section 2.5.4.5.5. The placement of backfill is discussed in Section 2.5.4.5.4. All buried pipes satisfy the diameter-to-thickness ratio (<300) requirement in accordance with Reference 2.5-52. Therefore, piping deflection due to earth load will not exceed the allowable. In addition, in accordance with table 1 of Reference 2.5-52, deflections due to AASHO HS-20 loading with the minimum required cover of 4 ft are less than the allowable deflections. Process piping located under railroad tracks that are not encased in concrete are approximately 12 ft below-grade. Therefore, the deflection due to Cooper E-80 railroad loading will not exceed the allowable. The more conservative E-80 loading was used in design for required railroad loadings outside of safety-related structures.

2.5.4.10.2.3 Valve Pits The valve pits for the RHR and ESW piping (Figure 3.8-64), which are not founded directly on bedrock, fall into the following categories:

a. Valve pits founded on concrete backfill with a minimum compressive strength of 2000 psi, which bears on bedrock CHAPTER 02 2.5-73 REV. 13, SEPTEMBER 2006

LGS UFSAR

b. Valve pits for diesel oil storage tanks shown on Figure 2.5-37 are founded on cementitious backfill with a minimum compressive strength of 80 psi, which bears on weathered rock.

c. One valve pit (located near RHR and ESW piping for Unit 2 as shown in Figure 2.5-37, coordinates N5819.5, E4205.0) is supported on cementitious backfill which bears on natural soil, with minimum bearing capacity of 6000 psf.

The valve pits are designed for AASHO HS-20 truck loading. The maximum calculated pressure under the base slabs is 2440 psf.

The walls of the valve pits are designed to resist lateral loads due to backfill having a saturated unit weight of 140 pcf and an at rest earth pressure coefficient of 0.7, and a surcharge of 250 psf due to AASHO HS-20 truck loading and lateral force increment due to seismic loading as shown on Figure 2.5-39. The roofs of the valve pits are adequately designed to resist AASHO HS-20 truck loading, tornado depressurization or missile impact. Because the valve pits are founded on concrete or cementitious backfill, the amount of settlement is considered to be insignificant.

2.5.4.10.2.4 Electrical Ducts Electrical ducts are encased in Class A concrete having a minimum design strength of 2000 psi.

The ducts are buried a minimum of 4 ft below finished grade with Type I or II fill placed on top and compacted as described in Section 2.5.4.5.4. The duct banks are founded on either bedrock, weathered rock, dense natural soil or compacted Type I fill. Where the bottom of the trenches were overexcavated, they were backfilled under the ducts with a minimum of 6 inches of either select granular, cementitious, or concrete backfill.

Select granular, cementitious, and concrete backfill are described in Section 2.5.4.5.4.

All Class I electrical ducts have a minimum 4 ft of backfill on top, which has been found adequate for Cooper E-80 loading without causing significant settlement or loading of ducts or foundation.

2.5.4.11 Design Criteria 2.5.4.11.1 Design Criteria For Safety-Related Structures on Rock The plant structures founded on rock are designed for a maximum acceleration of 0.15 g from an occurrence of the SSE event. From consideration of its engineering properties, it is evident that the foundation rock would not be measurably affected by seismic loadings, and negligible additional foundation settlement would accompany these maximum potential dynamic loads. The maximum contemplated total static and dynamic loads are only a fraction of the bearing capacity of the rock, thus ensuring an ample margin of safety.

2.5.4.11.2 Design Criteria For Safety-Related Structures on Soil The design criteria and methods of design concerning the liquefaction potential of soil at the spray pond are discussed in Section 2.5.4.8. The design criteria and stability analyses of the spray pond slopes are discussed in Section 2.5.5.2.

CHAPTER 02 2.5-74 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5.4.12 Techniques to Improve Subsurface Conditions 2.5.4.12.1 Treatment of Fracture Zones and Clay Seams - Main Power Block Area During excavation of the main power block area, three fracture zones and two clay seams were encountered in the foundation rock. The locations of these zones are shown on Figure 2.5-13. The zones are described in Section 2.5.1.2. Treatment of these zones during construction is described below.

2.5.4.12.1.1 Criteria for Treatment of Fracture Zones and Clay Seams As outlined in the PSAR, the criteria for treatment of the steeply dipping fracture zones are to:

excavate soft or excessively fractured material in the zones under footings to a depth below foundation grade that is at least equal to the width of the undesirable material at foundation grade; slope the sides of these excavations so that they become narrower downward; extend this excavation beyond the edges of footings crossing the fracture zones; and replace the excavated material with concrete. This dental treatment is designed to replace the compressible material under footings with a wedge of concrete that transfers vertical loads laterally to the adjacent sound rock on each side of the zone, to confine this adjacent rock, and to reduce unit loads by extending the bearing area beyond the edges of walls and columns.

The criteria for treatment of a clay seam, except under column footings, are to determine the thickness of soft, compressible material in the seam at its intersection with foundation grade and to treat the seam only if the compressive material is 3/4 inch or more thick. Under wall footings, treatment consisted of removing the compressible material in the seam and the rock above it, to provide a minimum of 3 feet of sound rock between the bottom of the footing and the seam. The material removed was replaced with concrete. Isolated column footings do not span the intersection of the clay seam with foundation grade; where this situation could occur, the clay seam and all rock above it were removed under the column footing and replaced with concrete so that all of the footing is founded on rock below the seam. Where the clay seam was not treated, compressible material in the seam is thin (3/4 inch or less). This material commonly contains hard rock fragments; thus, the total consolidation resulting from it is very slight. Also, the consolidation occurs rapidly after loading and is completed before the structures are finished.

2.5.4.12.1.2 Treatment of Fracture Zone A The location, attitude, and width of Zone A at final foundation grade are shown on Figure 2.5-13.

The zone is described in detail in Section 2.5.1.2. Fracture Zone A trends generally N 40 to 45 E and dips from 70 to vertical.

Zone A is locally over 10 feet wide at ground surface, but narrows rapidly with depth. Weathering along the zone decreases with depth, and the fractures often converge with depth. At foundation grade, weathered, soft material between fracture surfaces is usually thin, or absent entirely. As a result, Zone A usually is minor in extent at foundation grade, and does not normally require treatment under thick, stiff, reinforced footings, since such footings could span over a narrow, steeply dipping zone. It was treated under wall footings of Category I structures, however,in accordance with the criteria presented in the PSAR.

Photo 1 (figure 13A) of Figure 2.5-13 shows a typical section of Fracture Zone A at foundation grade. Photo 2 (figure 13B) of Figure 2.5-13 shows the type of excavation performed for dental concrete placement. Photo 3 (figure 13C) of Figure 2.5-13 shows placement of dental concrete at "Mh" and "N" lines.

CHAPTER 02 2.5-75 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5.4.12.1.3 Treatment of Fracture Zone B Fracture Zone B occurs at the southeast corner of Unit 2 reactor enclosure, as shown on Figure 2.5-13. The zone trends N 30 E and is nearly vertical. It contains from 2-10 inches of crushed, weathered rock and plastic clay. The rock on the southeast side of the zone shows an apparent displacement of 8-10 inches downward relative to that on the northwest side. Zone B is treated in the same manner as Zone A. Dental excavation for Zone B at "D" line is shown on Photo 4 (figure 13D) of Figure 2.5-13. Photos 5 and 6 (figures 13E and 13F) of Figure 2.5-13 show Zone B before and after dental excavation for the column footing at the intersection of lines F and 31.9.

Treatment excavation is performed using picks, shovels, and a backhoe with a hydraulic breaker.

The excavation is deeper and wider than the actual fracture zone in order to obtain a wedge effect into the zone. This wedge was created as shown on the photographs of Figure 2.5-13. The excavation continues past the column footing in order to increase the bearing area under the footing.

2.5.4.12.1.4 Other Fracture Zones Another less significant fracture zone was exposed in the excavation between Zones A and B and north of "J" line. Its location and description are shown on Figure 2.5-13. This zone contains from less than 1 inch to a maximum of about 4 inches of soft material, and they are nearly vertical; consequently, they required no special foundation treatment. The adjacent rock is hard and strong, although sometimes closely jointed, and can easily support the loads imposed by the structures.

As noted on Figure 2.5-13, this zone shows slight offsets of several inches.

2.5.4.12.1.5 Treatment of Clay Seams The clay seam in Unit 1, described in Section 2.5.1.2, is hereafter referred to as Seam 1. A second clay seam, hereafter referred to as Seam 2, is stratigraphically lower than Seam 1 and intersects foundation grade in part of Unit 2. These seams occur along shaly beds that are relatively softer than the adjacent hard siltstone. As mentioned in Section 2.5.1.2, there has apparently been some shearing along these seams, since they show slickensides and some crushing.

Seam 1 has a maximum thickness of about 10 inches near the top of rock south of Unit 1, but its thickness decreases rapidly downdip, so that under Unit 1, it is generally less than about 11/2 inches thick, and in places, it is barely discernible, Where the seam contains clay, it also usually contains rock fragments.

As described in the PSAR, three large diameter core holes were drilled downdip from the seam's exposure on excavated rock surfaces to verify that the seam continues to become thinner downdip.

These holes show the seam to be no more than 3/4 inch thick.

Figure 2.5-13 shows the approximate trace of Seam 1 on the excavated rock surface in the radwaste building and in Unit 1. Normal excavation extends below this seam in Unit 2, as the rock tends to break naturally to the seam, and cleanup and foundation preparation removed the seam northward to the "J" line. Thus the trace of the seam is at the toe of the excavated slope near the "J" line.

Under the mat for Unit 1, clay in Seam 1 has a maximum, thickness of less than 1/2 inch. Where this seam intersects grade, the thin silver of rock above the seam tends to break naturally to the CHAPTER 02 2.5-76 REV. 13, SEPTEMBER 2006

LGS UFSAR seam and was removed during the normal foundation preparation. Special treatment for Seam 1 was done in only one area, under the footing for column F-15.5. Here, all the rock above the seam was removed under that footing. Elsewhere, Seam 1 is so thin that no treatment was required.

The second clay seam, Seam 2, occurs along a relatively soft, thin shaly bed that is stratigraphically lower than Seam 1.

Soft material in Seam 2 is usually only about 1/4-1/2 inch thick where exposed in foundations or excavated slopes, and it consists of plastic clay with shaly, sheared rock fragments. Seam 2 was removed under several column footings in Unit 2. At columns F-31.9, H-31.9, F-30.5, and H-30.5, all the rock above Seam 2 was removed (the footings rest on rock below the seam).

2.5.4.12.1.6 Summary and Conclusions on Treatment of Fracture Zones and Clay Seams The fracture zones, although narrow and steeply dipping, have been conservatively treated to preclude the occurrence of any significant settlements due to these zones. They pose no threat to the stability of slopes after the structures are completed. The stresses that caused the fracture zones, and the other fractures and joints in the foundation rock, no longer exist and have not existed for millions of years. The faults near the site have been inactive for millions of years, as documented in the PSAR and the geologic report prepared by Dames and Moore in 1974 (Reference 2.5-1).

The two thin clayey seams along bedding were carefully evaluated in the foundation rock and they were treated under some column footings. Settlement due to consolidation of clay in untreated seams is negligible, since the clay is so thin. Any consolidation of the clay will have occurred before the structures are completed.

2.5.4.12.2 Treatment of Fracture Zones and Clay Seams - Spray Pond and Other Areas No special treatment of foundation rock was required at the spray pond. The foundations for the spray pond pumphouse, overflow structure, and spray network pipe supports were excavated to unweathered bedrock.

Two bedding-plane clay seams, generally not exceeding 1 inch in thickness, were encountered one of which trends across the bottom of pond excavation (Figure 2.5-42). Where exposed in the excavations for the spray networks, the seam was excavated below foundation grade whenever it was overlain by less than one or two feet of firm rock. At the eastern spray network (network "B")

the seam was slightly thicker and additional overlying rock was removed beneath the pipe support footings (Figure 2.5-42). At the spray pond pumphouse excavation, this seam was not exposed on bearing rock except in the western abutment trench, in which case all rock overlying the seam was removed.

A considerable amount of soil and soft, weathered rock was overexcavated at the western spray network (network "C"), and replaced by up to 14 feet of backfill concrete (section A-2 of Figure 2.5-43). Small amounts of broken or weathered rock were also removed as locally required along minor east-west trending fracture zones (Figure 2.5-42).

The entire bottom of the spray pond whether underlain by soil or rock, and the soil slopes below the elevation of the perimeter service road, are covered with a soil-bentonite liner (Section 2.5.5.4).

The spray pond rock-cut slopes are covered with shotcrete. The shotcrete extends below the soil-CHAPTER 02 2.5-77 REV. 13, SEPTEMBER 2006

LGS UFSAR bentonite liner to provide a suitable transition zone. A plan and typical sections are shown on Figure 3.8-57.

2.5.4.13 Subsurface Instrumentation Subsurface instrumentation is not required at the site. Load- bearing safety-related structures are founded on competent bedrock. The properties of the bedrock are presented in Section 2.5.4.2.1.

Instrumentation for surveillance of rock foundations was not installed for the reasons described in Section 2.5.4.5.

2.5.5 STABILITY OF SLOPES Natural slopes at the site are shown on the Site Plan, Figure 2.1-3.

There are no steep or unstable natural rock slopes within the plant boundaries that would have nay adverse effect on the safety-related operation of the LGS plant. Although excavated rock slopes could be cut vertically, the rock would eventually weather and ravel to a flatter slope with time.

Therefore, all permanent excavated rock slopes are either cut on a gradient no steeper than 1 horizontal to 1 vertical, or they have engineered backfill placed against them to provide additional stability. The only permanent exposed rock slopes that could affect safety-related structures are those in the spray pond excavation.

The soil slopes considered are the excavated soil slopes in the spray pond. In addition, the dikes surrounding storage tanks are discussed in Section 2.5.5.5.

2.5.5.1 Slope Characteristics The spray pond was constructed primarily by excavation. The spray pond is designed for the water level to be at el 251' and for the bottom of the pond to be at el 241'. Because of the sloping bedrock surface, the bottom of the pond is located in both soil and rock. The top of the rock is shown on Figure 2.5-9. Cut slopes are 4 horizontal to 1 vertical in soil and generally 1 horizontal to 1 vertical in rock, with a bench at the soil-rock interface. A short section of weathered rock is excavated at 2 horizontal to 1 vertical.

The bottom of the spray pond, excavated in both soil and rock, and soil slopes below the perimeter service road (at el 255') are covered with a 12 inch thick liner of a soil-bentonite mixture (Figure 2.5-24). The liner is protected by a 12 inch thick soil cover to reduce the possibility of the liner drying and cracking and to prevent damage from equipment during construction and maintenance.

The design and construction of the soil-bentonite liner and soil cover are discussed in Section 2.5.5.4.1. In order to prevent erosion due to wave action, the liner and soil cover at the soil slopes are protected by a 12 inch thick layer of riprap bedding and an 18 inch thick layer of riprap. The gradation and placement of the riprap and riprap bedding are discussed in Section 2.5.5.4.2. Rock slopes are covered by shotcrete, which extends below the soil-bentonite liner in the bottom of the pond. A plan and typical sections of the slopes are shown on Figures 2.5-48 and 2.5-49.

2.5.5.1.1 Geologic Conditions Rock at the site consists of well-indurated, gently dipping sandstone, shale, and siltstone of Triassic-age. Bedding dips 8 to 20 to the north, with only minor local variations. Two major joint systems are prevalent in the area. Both are vertical or nearly vertical; they strike approximately N CHAPTER 02 2.5-78 REV. 13, SEPTEMBER 2006

LGS UFSAR 20 to 50 E and N 50 to 60 W. Three fracture zones and two clay seams were encountered in foundation excavations in the main power block area; they are described in Section 2.5.1.2.5.

Bedrock is overlain by from 0 feet to about 40 feet of residual soil, developed in situ by the gradual decomposition of the parent rock. No clearly defined boundary between soil and rock exists, as the soil grades into fresh, unweathered rock.

The spray pond is underlain mostly by Lockatong lithofacies beds, except at the southeastern and northwestern margins of the pond (Figure 2.5-42). The Lockatong rocks are characterized by black shale and associated fine-grained, variably calcareous beds deposited in a lacustrine or swampy environment, in contrast to the sandier, typically red beds of the Brunswick Formation. Two prominent black shale or claystone beds, neither exceeding about 4 feet in thickness, were exposed and mapped in the spray pond excavation (Figure 2.5-42). These same beds had been recognized and logged from rock cores during the initial 1969-1970 site exploration by Dames and Moore (see log for Boring 150, Figure 2.5-22).

Joints exposed at the spray pond trend northeast and northwest, the former being dominant. A third prominent joint set trending approximately east-west in the Brunswick sandstone beds tends to be more widely spaced and shows irregular fracturing in the Lockatong (Figure 2.5-42). Major north-south jointing is locally prominent near the middle of the excavation (spray network "D").

Generally, any particular joint was found to be confined vertically to a single bed or series of lithologically similar beds.

Only a few minor offsets were noted in the excavation, along near-vertical joints and fractures.

They are described in detail in Section 2.5.1.2.5.

Slickensides are locally abundant in the spray pond excavation. They occur almost without exception as dip-slip striations on curved, gently-to-moderately dipping joint surfaces, in beds that overlie claystone or shale units in the Lockatong. Where observable in profile at vertical cuts within the excavation, these curved joint planes wee seen to terminate downward at the contact with the underlying shale. It is therefore concluded that the slickensides arise from slight differential movement related to compaction of the shales during diagenesis and have no tectonic significance.

Figure 2.5-42 shows typical occurrences of these slickensides.

2.5.5.1.2 Exploration - Emergency Spray Pond The exploration for the site is presented in Section 2.5.4.3. Exploration for the emergency spray pond is discussed detail in this section.

2.5.5.1.2.1 General The exploration for the emergency spray pond was conducted in three phases. The first phase, conducted during May and June, 1973, consisted of 5 NX core holes and 43 auger holes to define the bedrock surface. The second phase, carried out in April and May, 1974, consisted of 25 borings and 4 test pits to obtain undisturbed soil samples in the spray pond area. The third phase, conducted in March, 1977, consisted of 13 test pits to investigate the soils for use in constructing the soil-bentonite liner.

2.5.5.1.2.2 Phase 1 Exploration Borings in the first phase were made by American Drilling Company of East Providence, Rhode Island. In order to define the bedrock surface, 43 auger holes were drilled in the spray pond area.

CHAPTER 02 2.5-79 REV. 13, SEPTEMBER 2006

LGS UFSAR The contours of the top of the bedrock are shown on Figure 2.5-9. Five NX core holes were drilled to determine the bedrock lithology, structure, and foundation conditions at the intake structure.

2.5.5.1.2.3 Phase 2 Exploration The Phase 2 exploratory borings were also made by American Drilling Company. Block samples from test pits were taken by Geotechnical Engineers, Inc. of Winchester, Massachusetts.

Twenty-five borings were made around the perimeter of the spray pond. The locations of these borings are shown on Figure 2.5-26. Both disturbed and undisturbed samples were obtained in all holes. Disturbed samples were obtained using a standard split spoon sampler (ASTM D1486).

Undisturbed samples were obtained using Shelby tubes and a Denison sampler equipped with a 3-7/8 inch ID, 24 inch long brass tube liner. Four test pits were dug to obtain undisturbed block samples. Their locations are shown on Figure 2.5-26.

Laboratory testing of soil samples was carried out by Geotechnical Engineers, Inc. All test results are given in a soil testing report from Geotechnical Engineers, Inc (Reference 2.5-39) and they are briefly discussed in Section 2.5.4.2.2.

Figures 2.5-27 through 2.5-29 show typical subsurface cross-sections through the spray pond.

2.5.5.1.2.4 Phase 3 Exploration Thirteen test pits were excavated at the spray pond site in March, 1977 to investigate the soils for use in constructing the soil-bentonite liner,. The locations of the test pits and the generalized soil profiles are shown on Figure 2.5-30. Bulk samples of soils suitable for constructing the soil-bentonite liner were taken from each of the test pits, except for one pit where suitable soil material was not encountered. A more detailed discussion on exploration and sampling is given in Section 2.5.5.4.1.1.

Laboratory testing of soil samples and soil-bentonite mixtures was carried out by Geotechnical Engineers, Inc. All test results are given in a July, 1977 report on soil-bentonite mixtures that was prepared by Geotechnical Engineers, Inc. (Reference 2.5-44).

2.5.5.1.3 Laboratory Testing of Soils The results of laboratory testing of undisturbed samples of soil from the spray pond site are presented in Section 2.5.4.2. The results of laboratory testing of soil-bentonite mixtures for the pond lining are presented in Section 2.5.5.4.

2.5.5.2 Design Criteria and Analyses The design analyses for the spray pond include seepage analyses, liquefaction potential evaluation of the soils in the pond, and stability of the side slopes. The seepage and the liquefaction evaluation are presented in Sections 2.5.4.6 and 2.5.5.8, respectively. The analysis of the stability of the pond side slopes is provided in this section.

2.5.5.2.1 Stability of Rock Slope Rock slopes at the spray pond are cut no steeper than 1 to 1. The slope in partly weathered rock west of the spray pond pumphouse is at 2H to 1V. Weathered bedrock elsewhere is cut at the CHAPTER 02 2.5-80 REV. 13, SEPTEMBER 2006

LGS UFSAR same slope as the soil, 4H to 1V. The height of the 1:1 rock slope varies from zero to about 20 feet. The bench at the top of the 2:1 rock slope is about 13 feet above bottom of pond excavation (11 feet above the top of the soil liner). The location of rock slopes at the spray pond are shown on Figure 2.5-48; Figure 2.5-49 shows representative rock slope profiles.

Stability of rock slopes is influenced primarily by the orientation of bedding attitudes with respect to that of the slope. Bedding dips quite uniformly at about 10 (range 8 to 13) to the northwest (N 20 W; strike N 70 E). Major joint trends are nearly east-west and NE-SW. Minor or locally prominent joints trend northwest and north-south. The great majority of these joints are nearly vertical.

Major consideration was given to stability of slopes on the south side of the pond, where beds dip uniformly out of the slope at about 10 to 11. Because the strike of beds is nearly parallel to the slope, potential failure would occur as a sliding block; prominent east-west joints, together with additional joint sets described above, would provide cohesionless release surfaces for the slide.

Simple single-plane, two-dimensional static analysis is therefore appropriate. Seismic loading is conservatively represented as the design SSE acceleration (0.15 g) applied to the entire mass of the assumed failure block, parallel to the maximum dip direction. Such a representative overestimates actual earthquake loading of the slopes and is therefore conservative.

1 to 1 Slope East of Spray Pond Pumphouse. Rock along the slope east of the spray pond pumphouse consists of highly competent, indurated (Calcareous) siltstone and fine-grained sandstone (section E of Figure 2.5-49). The entire slope required excavation by blasting; the in situ rock here resisted attempts at removal by heavy, single-toothed ripping equipment. It is likely that its strength exceeds typical values of unweathered foundation rock (Table 2.5-11). Prominent but widely spaced east-west joints, together with relatively thick bedding, resulted in a steplike slope profile; the average width of the steps which form due to breaking along bedding-planes is more than 1 foot. The rock, thoroughly examined for evidence of clayey or shaly interbeds or partings, displays no evidence of these features.

An assessment of the stability of this slope was made by calculating the angle of internal friction required along bedding-planes to achieve stability, assuming no cohesion on these planes.

Although the water table, presently well below the bottom of the pond, is not expected to be significantly affected by any seepage from the pond during operation of the plant, the slope behind the pond was conservatively assumed to be saturated (water table at el 251') for purpose of calculation (actually the weight of water in the pond would enhance the stability of the unsaturated slope). During rapid drawdown under static conditions, the water level behind the slope is assumed to remain at that elevation, and the toe is assumed to be undrained. Under assumed maximum seismic loading, a friction angle of only 25 or less would be sufficient to maintain stability at a safety factor of 1.25 if cohesion is assumed to be zero. This value is much lower than any reasonable coefficient of friction to be expected for the strength of bedding in this rock. In addition, significant cohesive strength is present, for otherwise beds could have been excavated by ripping.

Based on the above analysis, the slope east of the spray pond pumphouse is considered to be highly stable.

2 to 1 Slope West of the Spray Pond Pumphouse. The rock in the 2 to 1 slope west of the spray pond pumphouse (section F of Figure 2.5-49) is slightly to moderately weathered, most generally in the upper part of the slope. Jointing is generally well developed, parallel to the slope and in NW-SE and NE-SW orientations. At two intervals of strata exposed in the slope, bedding-planes with brown, thin clayey films or partings about one millimeter thick on them occur. (X-ray analysis of CHAPTER 02 2.5-81 REV. 13, SEPTEMBER 2006

LGS UFSAR similar brown clay from partings in rock exposed in the spray bank area indicate that kaolin constitutes most of the clay mineral content.) Bedding-planes are not perfectly planar but undulate on the order of several millimeters to several centimeters. The amplitude of these asperities with respect to the thickness of the clay film should impart a relatively high initial coefficient of friction against sliding, but it is difficult to quantify this effect. For stability calculations, initial friction angle was conservatively estimated to be 20 along the bedding-planes; for additional conservatism, the planes were assumed to be cohesionless although some cohesion occurs along the bedding (all detached blocks of rock were removed during cleaning of the slopes). Maximum weight of the sliding mass is calculated by assuming failure at the toe of the slope (el 239', below pond liner), the failure plane extending back under the 4:1 slope in soil above the rock. Factors of safety were computed for the following conditions:

CONDITION POND SLOPE LOADING

a. rapid drawdown empty saturated static

b. operation full saturated seismic

c. operation full drained seismic Water table levels and hydrostatic pressures are assumed to be the same as described above for the 1:1 slope. The calculations indicate that if rock bolts were not installed and shotcrete was not applied, a portion of the 2:1 slope would be unstable under rapid drawdown conditions and under seismic loading, provided that all of the numerous conservative assumptions are correct. In general, the weak bedding-planes described are exposed in the middle or upper part of the slope, whereas stability calculations assumed that the weak plane occurs at the toe of the slope.

Two lines of rock bolts were installed in the 2:1 rock slope to stabilize the two bedding-planes prior to placing shotcrete. Bolts were tensioned to 25 kips and spaced an average of 5 feet apart on the slope. A total of 38 rock bolts, 10 feet long, were installed.

The shotcrete on the rock has a significant stabilizing effect against sliding failure on weak bedding-planes because of its bond strength to rock and its shear strength. Slope failure would initiate along a bedding-plane where the plane forms a bench in the overall slope (Figure 2.5-49).

The shotcrete will be thickest at this location of the potential site of movement, at the intersection of a vertical rock face and a bench (a leveling course of shotcrete was applied to fill in rock re-entrants and overhangs). It is calculated that about 45 kips force per linear foot of slope would be required to shear the 3000 psi shotcrete, minimum 4 inches thick, assuming shear strength at 600 psi.

Shotcrete bond strength to rock of at least 100 psi or 14 kips per linear foot of slope is calculated to be resisting slope failure where there is a bench width of 1 foot in front of the failure plane; this is somewhat less than the average width measured in the field. It is important to note that the weakest and most prominent bedding-planes consistently form the widest benches in the slope, as would be expected. Good rock-to-shotcrete bond was assured during preparation of the slope for shotcreting by air-water jetting to remove loose and clayey material from the rock surface.

Table 2.5-7 shows the factors of safety against sliding under various conditions, considering the effect of rock bolts, and of rock bolts and shotcrete; the latter is the present condition of the pond.

The minimum factor of safety computed for the conservative assumptions made for the stability CHAPTER 02 2.5-82 REV. 13, SEPTEMBER 2006

LGS UFSAR analyses is 4.5. If cohesion of only 5 psi is assumed along the failure plane, the minimum factor of safety is increased to over 10.

The weakest bedding-plane in the excavation is at the eastern end of the 2:1 slope. A soft, black seam of clay about 1 inch thick was barely exposed (no more than 2 inches above bottom of pond) at the extreme northwestern corner of the spray pond pumphouse excavation; the plane dips below bottom of the excavation slope in a westward direction along the toe of the slope. Rock surfaces are exceptionally smooth and planar on each side of the seam. The clay fraction is mostly smectite (expandable clay) as determined by x-ray analysis.

Potential slope instability at this location is effectively eliminated by the buttressing provided by the header pipe footing (section G of Figure 2.5-49). The excavation for the footing, approximately 15x15 feet, was extended to hard, competent rock below the soft clay seam. The surface was thoroughly jetted with water and cleaned prior to placement of Class A concrete backfill. The footing and backfill concrete along with the rock below the bottom of the pond act as a buttress against possible sliding by virtue of the bond strength between concrete and competent rock and the passive resistance of the rock against which the Class A concrete was placed. Typical bond strength between concrete and the rock underlying the footing is more than 200 psi. Assuming a bond strength of only 100 psi, the footing affords a resistance of over 3,000 kips against slope failure, not including the frictional effect of load on the footing or the passive resistance of the rock at the north edge of the footing. Furthermore, substantial resistance against sliding is provided by Class A concrete backfill between the edge of the pumphouse excavation and the pumphouse wall (Figure 2.5-48). This concrete backfill, placed against the irregular edge of rock above the clay seam, helps to lock the rock mass in place.

A calculation was made of the force required to stabilize the slope against sliding on the soft clay seam at the extreme eastern end of the 2:1 slope, neglecting the stabilizing effects of the footing and concrete backfills. Assumptions are the same as described above, except that the angle of friction of the clay seam is assumed to be 8, with cohesion being zero. The section of slope considered is about 35 feet long, the portion where there is less than 2 feet of competent rock covering the plane at the toe of the slope (the shear strength of this rock is not considered in the calculation). Maximum force required to achieve stability is about 12 kips per foot, or a total of approximately 430 kips, compared to the resisting strength of the header pipe footing of over 3000 kips.

On the north side of the spray pond, beds dip into the slope producing an inherently stable slope configuration. The east slope of the pond bedding has a shallow apparent dip out of the slope only in the southeast segment where the competent rock, which required blasting, is exposed. Thus, slope instability along bedding is not a concern in these areas.

Consideration was given to toppling failure of portions of the slopes where vertical joints isolate columns of rock. The shallow angle of the excavated slopes is not conducive to such failure, and the continuous layer of shotcrete and wire mesh ties the rock together and eliminates any concern for failure due to toppling.

CHAPTER 02 2.5-83 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5.5.2.2 Design Criteria for Soil Slopes The soil slopes of the spray pond must be stable against sliding under normal conditions and also during and following an SSE event. The minimum acceptable factors of safety are:

End of construction 1.3 Long-term static condition 1.5 Long-term seismic condition 1.1 Rapid drawdown 1.25 The design value used for the maximum horizontal ground acceleration during the SSE is 0.15 g (Section 2.5.2.6).

2.5.5.2.3 Methods of Slope Stability Analyses The stability of the soil slopes was analyzed by the Simplified Bishop Method (Reference 2.5-45) using the computer program SLOPE (Reference 2.5-46) and by the wedge method of analysis described in the U.S. Navy Design Manual DM-7 (Reference 2.5-47) and in the U.S. Army Corps of Engineers Manual (Reference 2.5-48).

The wedge analysis was used to determine the sliding stability along the interface of the soil-bentonite liner and the in situ soil underneath. The Simplified Bishop method was used to determine the critical arc through the in situ soil and the corresponding factor of safety.

The wedge analysis considers the stability of the mass of soil above the bottom of the soil-bentonite layer, as shown on Figure 2.5-31. The tendency for this mass to move downslope is resisted by the passive earth pressure at the toe of the slope and the total shear strength developed along the bottom of the wedge. The sum of these resisting forces must be greater than the gravitational force that tends to cause downward movement. The factor of safety against sliding is equal to the ratio between the sum of the resisting forces and the gravitational driving force.

The Simplified Bishop method for evaluating the stability of a slope assumes a trial failure arc; the mass above the arc is broken up into a series of vertical slices, and the equilibrium of each of these slices is considered. It is assumed that the forces acting on the sides of any slice have zero resultant in the vertical direction (Reference 2.5-49). A factor of safety is obtained for the assumed trial failure arc by determining the ratio of the resisting forces to the driving forces. These procedures are repeated for various locations of the center of arc and different lengths of radius until a minimum factor of safety is obtained.

The cross-section used in all circular arc analyses is based on the subsurface conditions at the northwest end of the spray pond (Figures 2.5-27 and 2.5-28). This profile was selected because the depth to rock is the greatest and it, therefore, represents the most critical conditions encountered.

2.5.5.2.4 Design Parameters Design parameters required for the slope stability analysis include those for the in situ soils, compacted soil-bentonite mixture, compacted soils, riprap, and riprap bedding. Detailed discussions of the engineering properties of these materials are presented in various sections:

Section 2.5.4.2 for in situ soils; Section 2.5.5.4.1 for compacted soil-bentonite mixture and CHAPTER 02 2.5-84 REV. 13, SEPTEMBER 2006

LGS UFSAR compacted soil; and Section 2.5.5.2.4.2 for riprap and riprap bedding. Design parameters are summarized in Table 2.5-5 and given in the subsequent sections under the headings of various types of materials.

2.5.5.2.4.1 In Situ Soil

a. Unit Weight The total and saturated unit weights of in situ soils were determined in the laboratory; the average values chosen for design analyses are 122 lb/ft3 and 126.4 lb/ft3, respectively.

b. Static Shear Strength Effective shear strength parameters based on the results of consolidated-undrained triaxial tests are o = 33.5, and c = 0. The undrained shear strength of the material is 1.2 P o, where P o is the initial mean effective principal stress (Section 2.5.4.2).

c. Dynamic Shear Strength The dynamic shear strength of the soil is determined by multiplying the cyclic stress ratio required to cause 5% double amplitude strain in 5 cycles by the average effective confining pressure. This results in a design dynamic shear strength of 0.61o where o is the effective overburden pressure (Section 2.5.4.2.2.3).

2.5.5.2.4.2 Soil-Bentonite Mixture

a. Unit Weight The total and saturated unit weights of the soil-bentonite mixture (selected from laboratory test results) are 113 lb/ft3 and 119 lb/ft3, respectively. The mixture is compacted at optimum moisture content to 95% of maximum dry density as determined in ASTM D698.

b. Static Shear Strength The static effective shear strength parameters of the soil-bentonite mixture are based on the results of consolidated-undrained triaxial tests with pore pressure measurements (R). The design values of o = 29 and c = 0 are chosen from the test results shown on Figure 2.5-32, based on the U.S. Army Corps of Engineers recommendation that two-thirds of the test values should exceed the design value (Reference 2.5-48). The undrained shear strength of the mixture is 0.95 P o, where P o is the initial mean effective principal stress (Section 2.5.5.4.1.2.3).

c. Dynamic Shear Strength The dynamic shear strength of the soil-bentonite mixture is based on test results of undrained triaxial tests after cyclic loading (C R - R ). The test is described in Section 2.5.5.4.1.2.3, and the results are shown on Figure 2.5-33. The total shear CHAPTER 02 2.5-85 REV. 13, SEPTEMBER 2006

LGS UFSAR strength design parameter values o = 20 and c = 225 lb/ft2 are based on the U.S. Army Corps of Engineers recommendation that two-thirds of the test values should exceed the design value (Reference 2.5-48).

2.5.5.2.4.3 Compacted Soil The total and saturated unit weights determined from laboratory test results are 120.5 lb/ft3 and 123.8 lb/ft3, respectively. The values were determined at 95% of maximum dry density as determined in ASTM D698. The shear strength of compacted soil is assumed to be equal to the shear strength of the in situ soil.

2.5.5.2.4.4 Riprap and Riprap Bedding A unit weight of 130 lb/ft3 assumed for both the riprap and the riprap bedding material. The effective angle of internal friction is assumed to be 45for the riprap and 40 for the riprap bedding.

2.5.5.2.5 End of Construction Under Static Conditions The stability of the slope at the end of construction was analyzed using a total stress analysis.

The in situ soil is divided into sublayers, and the shear strength assigned to each layer determined from the undrained shear strength relationship of q = 1.2 P o (Section 2.5.5.2.4.1),

where P o is the effective average stress at the mid-depth of each sublayer. The minimum factor of safety of a circle is found to be 3.8, and the factor of safety of a wedge with a potential sliding surface along the bottom of the liner is found to be 5.7. Both factors of safety are considerably higher than the minimum value of 1.3 required by the design criteria. Figure 2.5-31 shows the sections analyzed and the critical arcs.

2.5.5.2.6 Rapid Drawdown Under Static Conditions Drawdown is assumed to be instantaneous from the normal pool level (el 251') to the bottom of the pond (el 241'). The stability of effective stress analysis, with o = 29 and c = 0 for the soil-bentonite liner, and c = 0 and o = 33.5 for the in situ soil and compacted soil. The minimum factor of safety of a sliding wedge is found to be 1.7, and the minimum factor of safety for the critical deep-seated circle is found to be 2.1. All factors of safety are higher than the minimum values required by the design criteria. The results are shown on Figure 2.5-31.

2.5.5.2.7 Long-Term Stability Under Static Conditions The long-term stability of the spray pond slope is calculated using an effective stress analyses.

The water level is assumed to have risen to its normal level of el 251' in the pond. The minimum factor of safety is determined using the infinite slope procedure and using effective shear strength parameters of the material. Under static conditions, the factor of safety is given by:

tan F.S. = (EQ. 2.5-5) tan where:

CHAPTER 02 2.5-86 REV. 13, SEPTEMBER 2006

LGS UFSAR

= effective friction angle

= angle of inclination of the slope For = 14 (4 horizontal to 1 vertical), the minimum factor of safety is 2.2 for a potential sliding along the soil-bentonite liner with a o = 29. The factor of safety of a wedge with a potential sliding surface through the interface between the soil-bentonite liner and the soil underneath is 2.6, and the minimum factor of safety for a critical circle is 2.4. Both the minimum factors of safety resulting from the infinite slope analyses, and that determined by the sliding wedge and circular arc analysis, are higher than the minimum values required by the design criteria. The results of these analyses are shown on Figure 2.5-31.

2.5.5.2.8 Long-Term Stability Under Seismic Conditions Two analyses were made for the stability of soil slopes under seismic conditions. A pseudostatic analysis using effective shear strength evaluated the stability of the slope during an earthquake, and a total stress analysis evaluated the stability of the slope immediately after the earthquake.

The pseudostatic method of analysis involves the computation of the minimum factor of safety against sliding either along a circular arc or along a plane surface or series of plane surfaces for a sliding wedge. A horizontal force is added to account for seismic forces. The entire analysis is handled as a static effective stress analysis. The added horizontal force is equal to the total weight of the sliding mass times the seismic coefficient. The total weight of the sliding mass is calculated using the moist unit weight of the soil above the phreatic surface and the saturated unit weight of the soil below the phreatic surface. For this analysis, it is conservatively assumed that the seismic coefficient used equals the design maximum ground acceleration (0.15g).

In the pseudostatic analysis, the strength used for the soil-bentonite mixture is o = 29 and c = 0 (Section 2.5.5.2.4.2), and = 33.5, c = 0 for the in situ soil (Section 2.5.5.2.4.1). The minimum safety factor obtained is 1.1 for both the circular arc and a sliding wedge through the soil-bentonite lining. All computed safety factors meet the minimum values required by the design criteria. The results are summarized on Figure 2.5-31.

Immediately after an earthquake, the pore pressures built up in the soil and soil-bentonite mixture during the earthquake reach their maximum values. Laboratory tests evaluated the undrained shear strength of both the soil-bentonite mixture and the in situ soil, without permitting any dissipation of pore pressure built up during cyclic stress loading. The total shear strength parameters representing the total strength of the soil-bentonite mixture and the in situ soil at the end of 5 uniform stress cycles are o = 20, c = 225 lb/ft2, and c = 0.61o respectively (Section 2.5.5.2.4.2 and Section 2.5.5.2.4.1). The minimum factor of safety for the critical circle and for the sliding wedge procedure are 2.3 and 4.3, respectively. Both values are higher than the minimum value of 1.1 required by the design criteria.

2.5.5.3 Logs of Borings The logs of borings and test pits from the spray pond exploration are presented in Figure 2.5-22.

Drilling and sampling procedures are discussed in Sections 2.5.4.3 and 2.5.5.1.2.

CHAPTER 02 2.5-87 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5.5.4 Compacted Fill The spray pond is primarily constructed by excavation (Figures 3.8-56 and 3.8-57). However, a soil-bentonite liner is placed on all soil slopes, as well as the pond bottom. The liner in turn is covered by a layer of compacted soil, riprap, and riprap bedding. The development of design and construction procedures for the development of a design mix for the compacted soil-bentonite mixture and the requirements for the compacted soils are covered in Section 2.5.5.4.1. The riprap and riprap bedding are discussed in Section 2.5.5.4.2.

2.5.5.4.1 Soil-Bentonite Liner and Soil Cover The following sections present the results of the investigations and studies made in designing a soil-bentonite liner for the emergency spray pond. These consist of a soil investigation of the spray pond site, laboratory testing and design of a suitable soil-bentonite mixture, slope stability analyses, and liner design. The following sections also briefly describe the construction procedures for the soil-bentonite liner and soil cover.

The design criteria for the soil-bentonite liner are:

a. The permeability of the soil-bentonite liner is less than 1 ft/yr (1x10-6 cm/sec).

b. The liner is protected to prevent drying and cracking, erosion from wave action, damage from equipment during construction and maintenance.

c. The slope of the pond with a liner is stable against sliding under static condition and during an SSE event. The minimum acceptable factors of safety against sliding are given in Section 2.5.5.2.2.

2.5.5.4.1.1 Soil Investigation and Sampling The purpose of the field exploration program was to investigate the soils to determine their suitability for use in constructing the soil-bentonite liner. Because the pond is constructed by removing more than 10 feet of material, the primary interest was to investigate whether there is sufficient suitable soil for soil-bentonite mixture within the limits of the pond excavation.

Thirteen test pits were made at the site in March 1977. These test pits were dug to depths ranging from 5 feet to 11.5 feet. A layer of sandy, clayey silt, which is suitable for a soil-bentonite liner, was found to be consistently located between the topsoil and the weathered siltstone/shale bedrock in all test pits. The thickness of this layer varies from less than 2 feet at the southeast end of the pond to more than 6 feet near the northwest end. At the northwest corner, the bedrock was not encountered in Test Pits 1 and 13, because the soils are thicker there. It was determined that suitable soil for the soil-bentonite mixture is available in amounts estimated to be about three times the quantity required for the liner.

At the time of the investigation, the site had been covered with approximately 3 feet of shot rock-fill.

The area was being used as a lay-down area for plant construction. The locations of the test pits and soil profiles are shown on Figure 2.5-30. Logs of test pits are included in Figure 2.5-22.

CHAPTER 02 2.5-88 REV. 13, SEPTEMBER 2006

LGS UFSAR Bulk samples of soils suitable for the soil-bentonite liner were taken from each of the test pits except TP-8, where suitable soil material was not encountered. In addition to the bulk samples, jar samples were taken at each test pit to determine the natural moisture content.

2.5.5.4.1.2 Laboratory Testing and Soil-Bentonite Mix Design All laboratory testing was performed in the laboratory of Geotechnical Engineers, Inc. of Winchester, Massachusetts. Soil samples obtained during the field investigation, along with the bentonite to be used in the tests, were shipped to the laboratory.

The laboratory testing proceeded in three stages. In the first stage, the onsite soils were classified.

In the second stage, moisture-density relationships were established for soil-bentonite mixtures, and the bentonite content to meet permeability requirements was determined. In the final stage, the static and dynamic engineering properties were determined for the soil-bentonite mixture at the bentonite content established in the second stage.

The first stage was performed to obtain the index properties of the soil specifically used in testing for the soil-bentonite mix design, rather then using the properties previously obtained at the spray pond site (Section 2.5.4.2.2.1). As can be seen below, the results are similar.

The laboratory test procedures and the results of all tests are summarized in the July, 1977 report prepared by Geotechnical Engineers, Inc. (Reference 2.5-44).

2.5.5.4.1.2.1 Classification of Onsite Soils Soil samples were classified visually in the laboratory as well as by testing to determine grain-size distribution, natural moisture content, and Atterberg Limits. The test results are summarized in Table 2.5-6.

a. Grain-Size Distribution Grain-size determinations were made on all bulk samples according to ASTM D422, except that the sample preparation avoided air-drying of fines. During sieving of the moist soil through the 3/4 inch sieve and during washing through the No. 200 sieve, only finger pressure was applied to break clay lumps and weak particles. These procedures prevent breakdown of soil particles.

The range of grain-size curves is shown on Figure 2.5-34. The median grain-size (D50) of the soils varies between 0.006 mm and 2.38 mm, with an average of 0.20 mm. The percent by weight passing the No. 200 sieve varies from 41% to 88%,

with an average of 71%.

b. In Situ Moisture Content Moisture contents of all jar and bulk samples were determined in accordance with ASTM D2216. The moisture contents, determined from the samples kept in water-tight jars, are believed to be representative of in situ moisture contents, and are included in Table 2.5-6. The range of in situ moisture contents for the soils varies between 15.0% and 30.7%, with an average of 23.6%.

c. Atterberg Limits CHAPTER 02 2.5-89 REV. 13, SEPTEMBER 2006

LGS UFSAR Atterberg Limits were determined for the 12 bulk samples in accordance with ASTM D423 and ASTM D424, except that the soil was not air-dried before testing, and the grooving tool used was of the Casagrande-type. The results are summarized in Table 2.5-6 and are plotted on Figure 2.5-35.

2.5.5.4.1.2.2 Soil-Bentonite Mix Design Three soils, which represent the entire range of grain-size distribution for onsite soils were selected for the soil-bentonite mixture design. These soils were designated as Soils A, B, and C, and each was obtained by combining two bulk samples. Their grain-size distributions are shown on Figure 2.5-34. Soils A and C represent the finest and the coarsest soils, respectively, in the spray pond area, and Soil B is typical of the soils in the middle of the range. The bentonite used in the tests was a high swelling Wyoming-type bentonite. The bentonite was in granular form before mixing with soil and water (Reference 2.5-44).

a. Compaction Tests Compaction tests were made on Soils A, B, and C in accordance with ASTM D698.

For each soil, the amount of bentonite mixed with the soils varied from 0% to 20%

by weight.

The bentonite used in the test was in granular form, with the size of granules ranging from 0.1 mm to 1.0 mm. Mixtures of soil and bentonite before compaction had the appearance of granular soil. No difficulty was experienced during mixing and compacting of soil-bentonite mixtures as the percentage of bentonite was increased, because the moisture was kept at optimum.

Results of the compaction tests are shown on figures 6, 7, and 8 of Reference 2.5-44. As shown on these figures, the maximum dry density decreases as the bentonite content increases, with the maximum dry density dropping about 10%

when the bentonite content increases from 0% to 20%. However, the optimum moisture content did not vary significantly (less than 2%) when the bentonite content was varied from 0% to 20%.

b. Permeability Tests Permeability tests on Soils A, B, and C mixed with 0% to 20% bentonite were performed on compacted specimens in triaxial cells. The specimens had a diameter of approximately 2.8 inches and were about 1.0 inch high. The specimens were prepared at optimum moisture content and compacted to approximately 95%

of the maximum dry density determined by ASTM D698. A Harvard Miniature compaction device was used to compact the specimens.

After compaction, each specimen was placed in a triaxial cell and consolidated under an effective confining. All specimens were allowed to saturate and swell before being tested for permeability. Saturation was checked by determining the value of Skempton's "B" parameter (Reference 2.5-40). All specimens were saturated to a "B" parameter of 0.95 or higher. The swelling of the specimens was roughly proportional to the bentonite content. Specimens with 0% to 5% bentonite showed little swelling, while specimens with 20% bentonite swelled as much as 10%.

CHAPTER 02 2.5-90 REV. 13, SEPTEMBER 2006

LGS UFSAR The results of the permeability tests are shown on Figure 2.5-36. The permeability decreases from greater than 10-5 cm/sec to less than 10-7 cm/sec when the bentonite content increases from 0% to 20%.

The details of the permeability tests are given in Reference 2.5-44.

c. Selecting Design Mix of Soil and Bentonite Based on the results of permeability tests, a soil-bentonite mixture with 7.5% bentonite content by weight in terms of soil-bentonite mixture is selected for design, i.e., a mix with 7.5% bentonite and 92.5% soil by dry weight. When mixed with water at the optimum moisture content and compacted to 95% of maximum dry density as determined in ASTM D698, the compacted mixture has a permeability of less than 10-6 cm/sec (1 ft/yr) (Figure 2.5-36).

2.5.5.4.1.2.3 Engineering Properties Tests Static and dynamic triaxial tests were conducted to determine the strength characteristics of the design soil-bentonite mixture (Section 2.5.5.4.1.2.2).

Soils A and C, which envelop the range of the grain-size distribution of the onsite soils sample (Figure 2.5-34), were used in making the soil-bentonite specimens for the strength tests. The test specimens were approximately 6.1 inches high and approximately 2.8 inches in diameter.

a. Consolidated-Undrained Triaxial Tests with Pore Pressure Measurements ( R )

Six consolidated-undrained triaxial tests ( R ) were performed. Three specimens were mixed with Soil A, and another three with Soil C.

The specimens were consolidated isotropically at effective consolidation pressures of approximately 0.15 tons/ft2, 0.30 tons/ft2, and 0.60 tons/ft2. These consolidation pressures simulate actual field conditions. Specimens were saturated using a back pressure of 10 tons/ft2 to achieve a "B" parameter of 0.95 or higher (Reference 2.5-44). After saturation, the test specimens were axially loaded in compression, and pore pressures were monitored.

The results of the six R tests are summarized on Figure 2.5-32, which shows maximum shear stress at failure versus the mean effective principal stress at failure. From this drawing, the effective friction angle obtained is 29. Detailed test data and results are presented in table 4 and figures 14 and 15 of Reference 2.5-44.

The undrained shear strength of the soil-bentonite mixture is calculated from the effective stress friction angle and pore pressure parameter Af at failure, using the same procedure as for the in situ soil (Section 2.5.4.2.2.2). For the samples tested, Af varies from 0 to -0.25. Using the conservative values of Af = 0 and o = 29, the undrained shear strength is calculated to be 0.95 P o, where P o is the initial mean effective principal stress.

b. Cyclic Loading Followed by Undrained Triaxial Compression with Pore Pressure Measurements (C R - R ) Seven C R - R tests were made in which the specimens CHAPTER 02 2.5-91 REV. 13, SEPTEMBER 2006

LGS UFSAR were first subjected to cyclic loading, followed by static axial loading under undrained condition. The purpose of this test was to evaluate any reduction of the shear strength of the soil-bentonite mixtures following a seismic event. Three specimens were prepared with Soil A, and four with Soil C.

The consolidation and saturation of the C R - R specimens were similar to that described for R testing. After saturation, the specimens were subjected to five cycles of sinusoidal cyclic loading, with deviator stresses of approximately 0.015 tons/ft2, 0.03 tons/ft2, and 0.06 tons/ft2. Drainage was not permitted. The selection of five uniform stress cycles simulates the SSE of 0.15 g at the site, based on correlations of equivalent uniform stress cycles and earthquake time histories by Seed, et al (Reference 2.5-41).

The deviator stress during cyclic loading was chosen to be approximately equal to the average shear stress induced during an earthquake, which is approximately equal to one-tenth of the total overburden pressure when the maximum ground acceleration is 0.15 g, that is:

Deviator Stress = average induced shear stress

= 0.65 Ha g (EQ. 2.5-6)

= 0.65 (h) 0.15

= 0.1 (h)

Axial deformations and pore pressures were monitored. A typical test record is presented on figure 16 in Reference 2.5-44.

Upon completion of the cycling, time was permitted for pore pressure to equalize throughout the specimens; however, drainage was not permitted. The specimens were then subjected to undrained axial compressive loading. Pore pressures were monitored.

The results of the seven C R - R tests are summarized on Figure 2.5-33, which shows maximum shear stress versus the mean principal stress at the end of 5 uniform stress cycles. From this drawing the total dynamic shear strength parameters are calculated to be = 20 and c = 225 lbs/ft2. Detailed test data and results are presented in table 5 and on figures 16, 17, and 18 of Reference 2.5-44.

2.5.5.4.1.3 Design Section Based on the design criteria given in the previous section and the results of seepage and stability analyses (Sections 2.5.4.6 and 2.5.5.2, respectively), a 12 inch soil-bentonite liner with soil and rock protection was adopted as a precaution (Section 2.5.4.6). The section is shown on Figure 2.5-24 and will conform to these additional requirements:

a. The liner is placed on a 4 horizontal to 1 vertical slope.

b. The liner on the slope is protected by soil, riprap and riprap bedding to prevent erosion from wave action, drying and cracking. Design details of riprap and its bedding are given in Section 2.5.5.4.2.

CHAPTER 02 2.5-92 REV. 13, SEPTEMBER 2006

LGS UFSAR

c. Soil cover on the liner on the bottom of the pond is designed to protect the liner from damage by equipment during construction and maintenance.

2.5.5.4.2 Riprap and Riprap Bedding The riprap and bedding are 18 and 12 inches thick, respectively. The gradation of the riprap and bedding is as follows:

Riprap Particle Weight Fraction of Total Mix (pounds) (% by weight)

>500 0 300 - 500 25 10 - 300 45 - 75

<10 0 - 25 Riprap Bedding Sieve Size Fraction Passing (U.S. Standard Sieves) (% by weight) 2 inch 100 3/4 inch 52 - 100 3/8 inch 36 - 70 No. 4 24 - 50 No. 16 10 - 30 No. 200 0 - 10 The riprap is placed in a manner to insure that the larger rock fragments are uniformly distributed.

The smaller rock fragments serve to fill the spaces between the larger rock fragments, resulting in a well-keyed, densely placed, uniform layer of riprap.

Control of riprap placement consists of visual observation of the operation and of the completed product to assure that a dense, rough-surfaced layer of well-keyed and graded rock fragments of the required thickness is provided.

The riprap bedding material is handled and placed in such a manner as to prevent segregation.

The moisture content of the material during placement is controlled as required to minimize segregation. A uniform layer of bedding is formed by spreading with a bulldozer.

2.5.5.4.3 Construction Brief descriptions of materials, foundation preparation, and construction for the soil-bentonite liner and soil cover are given in this section.

The in situ soils from the excavation within the limits of the spray pond are used for the liner and soil cover. The gradation of soils is checked before mixing with bentonite to make sure they are within the range shown on Figure 2.5-34 and that sizes larger than 2 inches are removed. The bentonite is high swelling, Wyoming-type bentonite.

CHAPTER 02 2.5-93 REV. 13, SEPTEMBER 2006

LGS UFSAR Water used for adjusting the moisture content of the soil-bentonite mixture is clean and free from deleterious substances.

Before placement of the liner, all underlying soil surfaces are prepared by scarifying, moisture conditioning, and compacting as necessary to meet the required density.

The pond bottom is underlain with soil and rock. Where the pond bottom is overexcavated, it is backfilled to the bottom level of the soil-bentonite liner. The soil backfill, as well as soft spots in the in situ soil, underlying the liner will be compacted to 95% of the maximum dry density determined by ASTM D698.

The soil-bentonite mixture is mixed in a central plant to ensure uniformity of the mixture. The mixture is then transported from the mixing plant to the construction site in clean equipment. The soil-bentonite mixture and the soil cover are placed and compacted to 95% of the maximum dry density determined by ASTM D698.

The individual components of the soil-bentonite mixture are tested for conformance with the specification before mixing. The proportions of the mixture are inspected at the mixing plant for conformance to specified proportions. The plant includes metering and feeding equipment that can be calibrated to the required accuracy. During compaction, the density of the liner and soil cover is checked to confirm that the requirements of the specification are met.

2.5.5.4.4 Seepage Test Upon completion of construction of the spray pond, a seepage test was conducted to verify that the design basis seepage rate of 1.83x106 gallons per month would not be exceeded.

Filling of the spray pond began on October 11, 1982, and was completed on November 22, 1982, when the water surface reached el 251.0'. A seepage loss test was conducted for a period of 23 weeks from December 13, 1982 to May 21, 1983. Measurements of seepage loss were determined for each weekly interval. Depending upon the evaporation rate factor considered, the average seepage loss from the pond during this test was calculated to be between 2.1x105 gallons per month and 3.8x105 gallons per month, or between about 5 gpm and 9 gpm for the entire pond.

This is 11% to 21% of the design basis rate. An error factor was developed which considers the accuracy of measurements of pond water level, of flows into and out of the pond, and of precipitation. The maximum error contribution was estimated to be approximately +/-22% of the design basis rate. With error contribution, the upper bound of the average seepage loss is approximately 40% of the design value, and the lower bound is close to zero. Therefore, the test demonstrates that actual seepage will not exceed the design basis rate. Details of the seepage test and data analysis are contained in Reference 2.5-145.

2.5.5.5 Dikes The dikes surrounding the storage tanks as shown on Figure 3.8-58 are not seismic Category I, but are designed in accordance with seismic Category IIA requirements as discussed in Section 2.4.12 and 2.5.4. However, the dikes will be stable under both static and dynamic conditions as discussed below.

a. A cross-section of the dikes is shown in Figure 2.5-44. The dikes are designed to be 4 feet wide at the crest and minimum 6.5 feet high, with side slopes of 2.25 CHAPTER 02 2.5-94 REV. 13, SEPTEMBER 2006

LGS UFSAR horizontal to 1 vertical. The dikes are supported by Type I (Unit 1) or II (Unit 2) fill, which varies in thickness from a few feet to more than 30 feet. Between the plant fill and bedrock is natural soil with a thickness varying from a few feet to more than 10 feet. Bedrock is therefore at a depth varying from about 10 feet to more than 40 feet below the dikes. The subsurface conditions described above are based on the logs of boring No. 12 at the tank located southwest of Unit 1 and at borings Nos. 4 and 5 at the tank south of Unit 2. Figure 2.5-20 shows the locations of these borings. The logs of the borings are shown on Figure 2.5-22.

The dikes will be compacted to 95% of the maximum density determined by ASTM D698 (Standard Proctor). Shear strengths for the compacted soil in the dikes and the fill supporting the dikes are assumed to be similar to the shear strength for compacted soil at the spray pond (Section 2.5.5.2.4.3); the angle of internal friction is 33.5. The natural soil below the fill is dense to very dense sandy and clayey silt with rock fragments. Properties of the natural soil have been added to Table 2.5-4 and Section 2.5.4.2.2.4.

The crest and the inside slopes of the dikes will be protected by a bituminous surface.

b. For design, the groundwater level at the tanks is conservatively considered to be 15 feet above the water table contours shown on Figure 2.5-15. This means that it will be at el 145' at the location of the CST southwest of Unit 1 and at el 175' at the location of the CST south of Unit 2. Based on the logs of borings 4, 5, and 12, the design groundwater level will be in the dense natural soil below the bottom of the fill and just above bedrock. Therefore, there will be no water table within the retaining dikes and the fill supporting them so that the fill and dike cannot liquefy.

Because the groundwater level is below the dikes, the infinite slope analysis and the yield acceleration analysis by Seed and Goodman (Reference 2.5-160) are considered appropriate for evaluating the stability of the slopes.

For static conditions, the infinite slope analysis was used to determine the factor of safety of soil slopes:

F.S. = tan tan i (EQ. 2.5-7) where:

= friction angle i = inclination of slope Therefore, for = 33.5 and i = tan-1 (1/2.25) = 24, the factor of safety under static conditions is found to be 1.5.

For dynamic conditions, the yield acceleration analysis was used. The yield acceleration is defined as the acceleration at which sliding will begin to occur. The yield acceleration coefficient (ky) is defined as:

CHAPTER 02 2.5-95 REV. 13, SEPTEMBER 2006

LGS UFSAR ky g = sin ( - i) g where and i were defined in the previous paragraph. For = 33.5 and i = 24, (ky) is found to be 0.165. Compared to the SSE of 0.15 g, the factor of safety for the dynamic condition would be F.S. = 0.165 = 1.1 0.15 These stability values are equal to the minimum values required by the design criteria given in Section 2.5.5.2.2.

Therefore, the dike slopes will be stable under both static and dynamic conditions.

The slope of the fill is 2 horizontal to 1 vertical. With an angle of internal friction of 33.5, the yield acceleration for such a slope is 0.12 g. During an SSE event, the fill slope may revel back to a slope of 2.15 horizontal to 1 vertical. A probable failure surface is shown in Figure 2.5-44. This would mean a loss of about 4.0 feet at the top of a 25 foot high slope. The ravelling will not influence the stability of the retaining dikes because the minimum space between the toe of the dike and the top of the slope of the fill is more than 20 feet (at the southeast corner of Unit 2 dike).

c. Other than the logs of three borings (Nos. 4, 5, and 12) at the locations of the CST, no separate field borings and laboratory test programs were conducted for the evaluation of stability of the dikes. The shear strength ( = 33.5) used in the stability analysis mentioned above was assumed, based on data for similar soils from the spray pond compacted to the same degree (Table 2.5-4).

d. The construction procedure for the earth part of the dikes will be similar to that used for the spray pond construction. The main requirements will be as follows:

1. Soil gradation Sieve Size  % Passing (by weight) 2 inch 100 3/4 inch 75 - 100 No. 4 50 - 95 No. 50 35 - 90 No. 200 30 - 85

2. Loose lift thickness

-For machine compaction 8 inches maximum

-For hand compaction 6 inches maximum

3. Compaction Minimum 95% of the maximum density as determined by ASTM D698, method D.

CHAPTER 02 2.5-96 REV. 13, SEPTEMBER 2006

LGS UFSAR

4. Testing In-place density tests will be performed and documented with a minimum of one on each lift or 300 cubic yards of soil placed, whichever is the more frequent.

5. Quality Control The quality control program will contain requirements for surveillance, audits, inspection, testing, and corrective action for nonconforming conditions.

6. Inservice Inspection The dikes surrounding the storage tanks are passive water-retaining barriers for short-term water retention. The inspection requirements will conform with appropriate portions of Regulatory Guide 1.127 to assure short-term water retention when needed.

2.

5.6 REFERENCES

2.5-1 Dames and Moore, "Geologic Report, Limerick Generating Station", Limerick, Pennsylvania, (1974).

2.5-2 N.M. Fenneman, "Physiography of the Eastern United States", McGraw-Hill, New York, (1938).

2.5-3 N.M. Fenneman and D.W. Johnson, "Physiographic Divisions of the United States",

(Scale 1:3000,000), USGS, (1946).

2.5-4 W.D. Thornbury, "Regional Geomorphology of the United States", John Wiley and Sons, New York, (1969).

2.5-5 G.W. Fisher, "The Metamorphosed Sedimentary Rocks Along the Potomac River near Washington, D.C." in G.W. Fisher, et al (eds.), "Studies of Appalachian Geology, Central and Southern", Interscience Publishers, New York, (1970).

2.5-6 J.D. Glaser, "Provenance, Dispersal, and Depositional Environments of Triassic Sediments in the Newark-Gettysburg Basin", Pennsylvania Geological Survey (4th Ser.), General Geological Report G43, (1966).

2.5-7 E.B. Knopf, and A.I. Jonas, "Geology of the McCalls Ferry-Quarryville District, Pennsylvania", USGS Bulletin 799, (1929).

2.5-8 Ernest Cloos and Anna Hietanen, "Geology of the Martic Overthrust and the Glenarm Series in Pennsylvania and Maryland", Geological Society of America, Special Paper 35, (1974).

CHAPTER 02 2.5-97 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-9 D.U. Wise, "Multiple Deformation, Geosynclinal Transitions and the Martic Problem in Pennsylvania", in G.W. Fisher, et al (eds.), Studies of Appalachian Geology, Central and Southern, Interscience Publishers, New York, pp. 317-334, (1970).

2.5-10 Carlyle Gray, et al., "Geologic Map of Pennsylvania", Pennsylvania Geological Survey, 4th series, Map No. 1, (1960, reprinted 1975).

2.5-11 J.C. Reed, Jr., "Tectonic Map of the Central and Southern Appalachians", in G.W.

Fisher, et al (eds.), "Studies of Appalachian Geology, Central and Southern",

Interscience Publishers, New York, (1970).

2.5-12 G.W. Stose, "Structure of the Honeybrook Uplift, Pennsylvania", Geological Society of America Bulletin, 45, pp. 977-1000, (1937).

2.5-13 V.E. Gwinn, "Thin-Skinned Tectonics in the Plateau and Northwestern Valley and Ridge Provinces of the Central Appalachians", Geological Society of America Bulletin 75, pp. 863-900, (1964).

2.5-14 C.L. Drake and H.P. Woodward, "Appalachian Curvature, Wrench Faulting, and Offshore Structures", Transactions of the New York Academy of Science, Series II, 26, pp. 48-63, (1963).

2.5-15 P.B. King, "The Tectonics of North America", USGS Professional Paper 628, (1969).

2.5-16 V.E. Gwinn, "Kinematic Patterns and Estimates of Lateral Shortening, Valley and Ridge and Great Valley Provinces, Central Appalachians, South-Central Pennsylvania", in G.W. Fisher, et al (eds.), "Studies of Appalachian Geology, Central and Southern", Interscience Publishers, New York, (1970).

2.5-17 R.T. Faill, "Tectonic Development of the Triassic Newark-Gettysburg Basin in Pennsylvania", Geological Society of America Bulletin, 84, pp. 725-740, (1973).

2.5-18 U.S. Coast and Geodetic Survey, "Crustal Movement Map Showing Probable Vertical Movements of the Earth's Surface, (August 1972)".

2.5-19 USGS, "Earthquake History of the United States", Publication 41-1, revised edition through 1970 (1973) and Supplement 1971-76, Boulder, Colorado, (1979).

2.5-20 USGS, "Earthquakes in the United States - 1974", Circulares 723A-723D, U.S.

Department of the Interior, (1976).

2.5-21 USGS, "Preliminary Determination of Epicenters", Washington, D.C., (coverage of 1937 - January 1982).

2.5-22 Daniel Linehan, S.J., "A Reevaluation of the Intensity of the East Haddam, Connecticut, Earthquakes of May 16, 1791", prepared for Connecticut Yankee Atomic Power Company, (December 1964).

CHAPTER 02 2.5-98 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-23 F.L. Fox and C.T. Spiker, "Intensity Rating of the Attica (N.Y.) Earthquake of August 12, 1929, a Proposed Reclassification", Earthquake Notes, 48, p. 37, (1977).

2.5-24 L.D. Brown and J.C. Oliver, "Vertical Crustal Movements from Leveling Data and Their Relation to Geologic Structure in the Eastern United States" Review of Geophysics and Space Physics, 14, pp. 13-35, (1976).

2.5-25 J.P. Owens, "Post-Triassic Tectonic Movements in the Central and Southern Appalachians as Recorded by Sediments of the Atlantic Coastal Plain", in G.W.

Fisher, et al (eds.), "Studies of Appalachian Geology, Central and Southern",

Interscience Publishers, New York, (1970).

2.5-26 J.M. Bird and J.F. Dewey, "Lithosphere Plate-Continental Margin Tectonics and the Evolution of the Appalachian Origin", Geological Society of America Bulletin, 81, pp.1031-1060, (1970).

2.5-27 R.S. Naylor, "Age Provinces in the Northern Appalachians", Annual Review of Earth Planetary Sciences, 3, pp. 387-400, (1975).

2.5-28 R.P. Ballard and E. Uchupi, "Triassic Rift Structure in the Gulf of Maine", American Association of Petroleum Geologists Bulletin, 59 (7), pp. 1041-1072, (1975).

2.5-29 Atomic Safety and Licensing Appeal Board, "Decision in the Matter of Consolidated Edison Company of New York, Inc., and the Power Authority of the State of New York", NRC Docket Nos. 50-3, 50-247, and 50-286: ALAB-436, (October 12, 1977).

2.5-30 John Rogers, "The Tectonics of the Appalachians Wiley- Interscience", New York, (1970).

2.5-31 USGS, "Final Review, Geology and Seismology, Boston Edison Company, Pilgrim Station, Unit 2, Plymouth County, Massachusetts", NRC Docket No. 50-471, transmittal to E.G. Case (NRC), (August 16, 1977).

2.5-32 NRC, "Supplement No. 3 to the Safety Evaluation Report by the Office of Nuclear Reactor Regulation, U.S. Nuclear Regulatory Commission in the matter of Boston Edison Company, et al, Pilgrim Nuclear Generating Station, Unit 2", Docket No.

50-471: NUREG-0022 (Supplement No. 3 to NUREG 75/054), (September 6, 1977).

2.5-33 Boston Edison Company, "Summary Report Geologic and Seismologic Investigations, BE-SG7602, Rev. 1", Pilgrim Station Unit 2, Docket 50-471, (March 1977).

2.5-34 F. Neumann, "Earthquake Intensity and Related Ground Motion", University Press, Seattle, Washington, (1954).

2.5-35 M.D. Trifunac and A.G. Brady, "On the Correlation of Seismic Intensity Scales with the Peaks of Recorded Strong Ground Motion", Seismological Society of America Bulletin, 65, pp. 139-162, (1975).

CHAPTER 02 2.5-99 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-36 I. Zeitz, and C. Gray, "Geophysical and Geological Interpretation of Triassic Structure in Eastern Pennsylvania", USGS, Professional Paper 400-B, pp. 174-178, (1960).

2.5-37 J.E. Sanders, "Late Triassic Tectonic History of Northeastern United States",

American Journal Science, 261, pp. 501-524, (1963).

2.5-38 B. Willard, et al, "Geology and Mineral Resources of Bucks County, Pennsylvania",

Pennsylvania Geological Survey, 4th Series, Bulletin 9, (1959).

2.5-39 Geotechnical Engineers, Inc., "Report on Soil Testing, Limerick Nuclear Station Spray Pond", Winchester, Massachusetts, (September 1974).

2.5-40 A.W. Skempton, "The Pore Pressure Coefficients A and B", Geotechnique, 4, p.

143, (1954).

2.5-41 H.B. Seed, I.M. Idriss, F. Makdisi, and N. Banerjee, "Representation of Irregular Stress Time Histories by Equivalent Uniform Stress Series in Liquefaction Analyses, Report No. EERC 75-29", University of California, Berkeley, California, (October 1975).

2.5-42 S.N. Davis and R.J.H. DeWiest, "Hydrogeology", John Wiley and Sons, Inc., New York, (1966).

2.5-43 H.B. Seed and I.M. Idriss, "Simplified Procedure for Evaluating Soil Liquefaction",

Journal of the Soil Mechanics and Foundations Division ASCE, pp. 1249-1273, (1971).

2.5-44 Geotechnical Engineers, Inc., "Report on Laboratory Soil Testing, Soil-Bentonite Mixtures, Limerick Generating Station", Winchester, Massachusetts, (July 1977).

2.5-45 A.W. Bishop, "The Use of the Slip Circle in the Stability Analysis of Slopes",

Geotechnique, 5, pp. 7-77, (1955).

2.5-46 McDonnell Douglas Automatic Company, "Slope Stability Analysis System", User's Manual ICES Slope, St. Louis, Missouri, (1974).

2.5-47 U.S. Navy, Bureau of Yards and Docks, "Soil Mechanics, Foundations, and Earth Structures, Design Manual NAVDOCKS DM-7, (1962)".

2.5-48 U.S. Army Corps of Engineers, "Stability of Earth and Rock-fill Dams, Engineers Manual EM 11110-2-1902", (April 1970).

2.5-49 T.W. Lambe and R.V. Whitman, "Soil Mechanics", John Wiley and Sons, Inc., New York, (1969).

2.5-50 Dames and Moore, "Report, Plate Bearing Tests, Reactor Building Area, Limerick Generating Station", Limerick Township, Pennsylvania, (March 12, 1971).

CHAPTER 02 2.5-100 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-51 Dames and Moore, "Report, Foundation Investigation, Proposed Limerick Generating Station", Limerick Township, Pennsylvania, Philadelphia Electric Company, (October 5, 1970).

2.5-52 American Iron and Steel Institute, "Welded Steel Pipe", Steel Plate Engineering Data, Vol. 3, (1977).

2.5-53 Winkler, L., "Catalog of U.S. Earthquakes before the Year 1850", Bulletin, Seismological Society of America, v. 69, n. 2, pp. 569-602, (1979).

2.5-54 Winkler, L., "Catalog of Earthquakes Felt in the Eastern U.S. Megalopolis 1850-1930", Division of Health, Siting and Waste Management, Office of Nuclear Regulatory Research, NRC, Washington, D.C. NUREG/CR-2532, p. 27, (1982).

2.5-55 USGS, "United States Earthquakes, 1928-1979", Golden, Colorado, (1930-1981).

2.5-56 Northeastern U.S. Seismic Network, "Bulletins 1-23 of Seismicity of the Northeastern United States, October 1, 1975 - June 30, 1981". Weston Observatory, Boston College, Boston, Massachusetts, (1976-1981).

2.5-57 Southeastern U.S. Seismic Network, "Bulletins 1-9 of "Seismicity of the Southeastern United States, July 1, 1977 - December 31, 1981". Seismological Observatory, Virginia Polytechnic Institute and State University, Blacksburg, Virginia, (1978-1982).

2.5-58 Meyers, H. and C.A. von Hake, "Earthquake Data File Summary", National Geophysical and Solar Terrestrial Data Center, Boulder, Colorado, (1976).

2.5-59 W.J. Hall and V.M. Newmark, "Seismic Design Criteria For Pipelines and Facilities",

Journal of The Technical Councils of AISC, (November 1978).

2.5-60 T.M. Berg, et al. (compilers), "Geologic Map of Pennsylvania", Pennsylvania Dept.

Environmental Resources, Topographic and Geologic Survey Map No. 1, scale 1:250,000, (1980).

2.5-61 F.H. Jacobeen, Jr., "Seismic Evidence for High Angle Reverse Faulting in the Coastal Plain of Prince Georges and Charles County, Maryland", Maryland Geological Survey Information Circular No. 13, Baltimore, Maryland, (1972).

2.5-62 H.D. Ackerman, D.C. Campbell, and J.D. Phillips, "Geophysical Model of Upper-Crustal Structures near Charleston, South Carolina", Geological Society of America, Abstracts with Programs, 10 (4), p. 161, (1978).

2.5-63 Y.P. Aggarwal and L.R. Sykes, "Earthquakes, Faults, and Nuclear Power Plants in Southern New York and Northern New Jersey", Science, 200, pp. 425-429, (1978).

2.5-64 C.J. Ando, et al, "Crustal Geometry of the Appalachian Orogen from Seismic Reflection Studies", Geological Society of America, Abstracts with Programs, 14 (1), p. 2, (1982).

2.5-65 J.G. Arbruster and Leonardo Seeber, "Intraplate Seismicity in the Southeastern United States and the Appalachian Detachment", in J.E. Beavers, ed.,

"Earthquakes and Earthquake Engineering - Eastern United States", Volume 1, Ann Arbor Science Publishers, Inc., Ann Arbor, Michigan, pp. 375-396, (1981).

CHAPTER 02 2.5-101 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-66 P.J. Barosh, "Cause of Seismicity in the Eastern United States: A Preliminary Appraisal", in J.E. Beavers, ed., "Earthquakes and Earthquake Engineering -

Eastern United States", Volume 1, Ann Arbor Science Publishers, Inc., Ann Arbor, Michigan, pp. 397-418, (1981).

2.5-67 J.C. Behrendt, R.M. Hamilton and H.D. Ackerman, "Deep Crustal Seismic Reflection Study Offshore in the Area of the Charleston, South Carolina, 1886 Earthquake", EOS Transactions American Geophysical Union, 61, p. 1040, (1980).

2.5-68 J.C. Behrendt, et al, "Cenozoic Faulting in the Vicinity of the Charleston, South Carolina, 1886 Earthquake", Geology, 9, pp. 117-122, (1981).

2.5-69 J.C. Behrendt, et al, "Marine Multichannel Seismic Reflection Evidence for Cenozoic Faulting and Deep Crustal Structure near Charleston", in G.S. Gohn, ed.,

"Studies Related to the Charleston, South Carolina, Earthquake of 1886 - Tectonics and Seismicity (Collected Abstracts)", USGS, Open-File Report 82-134, pp. 19-20, (1982).

2.5-70 G.A. Bollinger, "The Giles County, Virginia, Seismic Zone - Configuration and Hazard Assessment", in J.E. Beavers, ed., "Earthquakes and Earthquake Engineering - Eastern United States", Volume 1, Ann Arbor Science Publishers, Inc., Ann Arbor, Michigan, pp. 277-308, (1981).

2.5-71 M.H.P. Bott and D.S. Dean, "Stress Systems at Young Continental Margins",

Nature, 235, pp. 23-25, (1972).

2.5-72 K.W. Bramlett and D.T. Secor, "Displacement on the Belair Fault Zone in South Carolina", Geological Society of America, Abstracts with Programs, 12 (4), pp.

171-172, (1980).

2.5-73 D. L. Campbell, "Stress-Concentration Mechanism for Earthquakes in the Charleston, South Carolina Area", Geological Society of America, Abstracts with Programs, 10 (4), p. 164, (1978).

2.5-74 G. P. Citron and L. D. Brown, "Recent Vertical Crustal Movement from Precise Leveling Survey in the Blue Ridge and Piedmont Provinces, North Carolina and Georgia", Tectonophysics, 52, pp. 223-236, (1979).

2.5-75 F. A. Cook and J. E. Oliver, "The Late Precambrian-Early Paleozoic Edge in the Appalachian Orogen", American Journal of Science, 281, pp. 993-1008, (1981).

2.5-76 F. A. Cook, L. D. Brown and J. E. Oliver, "The Southern Appalachians and the Growth of Continents", Scientific American, 243 (4), pp. 156-159, (1980).

2.5-77 F. A. Cook, et al, "Thin-Skinned Tectonics in the Crystalline Southern Appalachians, COCORP Seismic-Reflection Profiling of the Blue Ridge and Piedmont", Geology, 7, pp. 563-567, (1979).

CHAPTER 02 2.5-102 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-78 F. A. Cook, et al, "COCORP Seismic Profiling of the Appalachian Orogen beneath the Coastal Plain of Georgia", Geological Society of America, Bulletin, Part I, 92, pp.

738-748, (1981).

2.5-79 Duke Power Company, "Review and Evaluation of Recent Geologic Information with Specific Reference to the Charleston Epicentral Area", Appendix to Catawba Nuclear Station USAR, (December 1981).

2.5-80 D. Elliott, G. W. Fisher and S. Snelson, "A Restorable Cross-Section Through the Central Appalachians", Geological Society of America, Abstracts with Programs, 14 (7), p. 482, (1982).

2.5-81 B. B. Ellwood, J. A. Whitney and D. B. Wenner, "Age, Paleomagnetism, and Tectonic Significance of the Elberton Granite, Northeast Georgia Piedmont",

Journal of Geophysical Research, 85, pp. 6521-6533, (1980).

2.5-82 J. A. Fischer, "Capability of the Ramapo Fault System", in J. E. Beavers, ed.,

"Earthquakes and Earthquake Engineering - Eastern United States", Volume I, Ann Arbor Science Publishers, Inc., Ann Arbor, Michigan, pp. 441-456, (1981).

2.5-83 S. A. Guinn, "Earthquake Focal Mechanisms in the Southeastern United States",

Prepared for Division of Reactor Safety Research Office of Nuclear Regulatory Research, NUREG CR-1530, NRC, Washington D.C., (1980).

2.5-84 J. T. Hack, "Rock Control and Tectonism - Their Importance in Shaping the Appalachian Highlands", USGS Professional Paper 1126-B, (1979).

2.5-85 R. M. Hamilton, "Geologic Origin of Eastern United States Seismicity", in J. E.

Beavers, ed., "Earthquakes and Earthquakes Engineering - Eastern United States",

Volume 1, Ann Arbor Science Publishers, Inc., Ann Arbor, Michigan, pp. 3-23, (1981).

2.5-86 R. M. Hamilton, "Geologic Setting of Seismicity in Eastern North America",

Geological Society of America, Abstracts with Programs, 14, p. 22, (1982).

2.5-87 L. D. Harris, "Thin-Skinned Tectonics and Potential Hydrocarbon Traps - Illustrated by a Seismic Profile in the Valley and Ridge Province of Tennessee", USGS, Journal Research, 4 (4), pp. 379-386, (1976).

2.5-88 L. D. Harris and K. C. Bayer, "Sequential Development of the Appalachian Orogen above a Master Decollement - A Hypothesis", Geology, 7, pp. 568-572, (1979).

2.5-89 L. D. Harris and R. C. Milici, "Characteristics of Thin-Skinned Style of Deformation in the Southern Appalachians, and Potential Hydrocarbon Traps", USGS, Professional Paper 1018, (1977).

2.5-90 R. D. Hatcher, Jr., "Developmental Model for the Southern Appalachians",

Geological Society of America, Bulletin, 83, pp. 2735-2760, (1972).

CHAPTER 02 2.5-103 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-91 R. D. Hatcher, Jr., "Tectonics of the Western Piedmont and Blue Ridge, Southern Appalachians: Review and Speculations", American Journal of Science, 278, pp.

276-304, (1978).

2.5-92 R. D. Hatcher, Jr. and A. L. Odom, "Timing of Thrusting in the Southern Appalachians, U.S.A.: Model for Orogeny", Journal Geological Society, London, 137, pp. 321-327, (1980).

2.5-93 R. D. Hatcher, Jr. and Isidore Zietz, "Tectonic Implications of Regional Aeromagnetic and Gravity Data from the Southern Appalachians", in D. Waves, ed.,

"Proceedings of IGCP Project 27: Caledonide Orogeny, 1979 Meeting", Geology Science, Memoir 2, Virginia Polytechnic Institute, Blacksburg, Virginia, pp. 235-244, (1980).

2.5-94 R. D. Hatcher, Jr. and Harold Williams, "Timing of Large-Scale Displacements in the Appalachians", Geological Society of America, Abstracts with Programs, 14, p.

24, (1982).

2.5-95 P. L. Heller, C. M. Wentworth and C. W. Poag, "Episodic Post-Rift Subsidence of the United States Atlantic Continental Margin", Geological Society of America, Bulletin, 93, pp. 379-390, (1982).

2.5-96 D. R. Hutchinson and J. A. Grow, "New York Bight Fault", USGS, Open-File Report 82-208, (1982).

2.5-97 W. P. Iverson and S. B. Smithson, "Master Decollement Root Zone Beneath the Southern Appalachians and Crustal Balance", Geology, 10, pp. 241-245, (1982).

2.5-98 F. H. Jacobeen, Jr., "Seismic Evidence for High Angle Reverse Faulting in the Coastal Plain of Prince Georges and Charles County, Maryland", Maryland Geological Survey Information Circular 13, (1972).

2.5-99 M. F. Kane, "Correlation of Major Eastern Earthquake Centers with Mafic/Ultramafic Basement Masses", in D. W. Rankin, ed., "Studies Related to the Charleston, South Carolina, Earthquake of 1886 - A Preliminary Report", USGS Professional Paper 1028-0, (1977).

2.5-100 K. D. Klitgord and J. C. Behrendt, "Basin Structure of the U.S. Atlantic Margin", in J.

S. Watkins, Lucien Montadert, and P. W. Dickerson, eds., "Geological and Geophysical Investigations of Continental Margins", American Association of Petroleum Geologists, Memoir 29, p.85-112, (1979).

2.5-101 L. T. Long, "Speculations Concerning Southeastern Earthquakes, Mafic Intrusions, Gravity Anomalies, and Stress Amplification", Earthquake Notes, Seismological Society of America, 47 (3), pp. 29-36, (1976).

2.5-102 L. T. Long and J. W. Champion, Jr., "Bouguer Gravity Map of the Summerville -

Charleston, South Carolina, Epicentral Zone and Tectonic Implications", in D. W.

Rankin, ed., "Studies Related to the Charleston, South Carolina Earthquake of 1886

- A Preliminary Report", USGS Professional Paper 1028, pp. 151-166, (1977).

CHAPTER 02 2.5-104 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-103 F. A. McKeown, "Hypothesis: Mafic Intrusions and Their Contact Zones Are Source Zones of Many Earthquakes in Central and Southeastern United States", (Abs.)

Earthquake Notes, 46 (4), p. 53, (1975).

2.5-104 F. A. McKeown, "Hypothesis: Many Earthquakes in the Central and Southeastern United States Are Causally Related to Mafic Intrusive Bodies", USGS, Journal Research, 6 (1), pp. 41-50, (1978).

2.5-105 R. B. Mixon and W. L. Newell, "Preliminary Investigation of Faults and Folds along the Inner Edge of the Coastal Plain in Northeastern Virginia", USGS, Open File Report 76-330, (1976).

2.5-106 R. B. Mixon and W. L. Newell, "Stafford Fault System: Structures Documenting Cretaceous and Tertiary Deformation along the Fall Line in Northeastern Virginia",

Geology, 5, pp. 437-440, (1977).

2.5-107 R. B. Mixon and W. L. Newell, "The Faulted Coastal Plain Margin at Fredericksburg, Virginia", Guidebook for Tenth Annual Virginia Geology Field Conference, October 13-14, 1978, USGS, Reston, Virginia, p. 50, (1978).

2.5-108 R. H. Moench, et al, "Comment and Reply, 'Thin-Skinned Tectonics in the Crystalline Southern Appalachians; COCORP Seismic-Reflection Profiling of the Blue Ridge and Piedmont'", Geology, 8, pp. 402-404, (1980).

2.5-109 H. J. Neugebauer and T. Spohn, "Metastable Phase Transitions and Progressive Decline of Gravitational Energy: Aspects of Atlantic-Type Margin Dynamics", in R.

A. Scrutton, ed., "Dynamics of Passive Margins, Geodynamics Series 6", American Geophysical Union, (1982).

2.5-110 W. L. Newell, D. C. Prowell and R. B. Mixon, "Detailed Investigation of a Coastal Plain - Piedmont Fault Contact in Northeastern Virginia", USGS Open File Report 76-329, (1976).

2.5-111 B. J. O'Connor and D. C. Prowell, "The Geology of the Belair Fault Zone and Basement Rocks of the Augusta, Georgia Area", Georgia Geological Society, Guidebook 16, (1976).

2.5-112 P. E. Olsen, "Triassic and Jurassic Formations of the Newark Basin", in Warren Manspeizer, ed., Field Studies of New Jersey Geology and Guide to Field Trips, 52nd Annual Meeting of the New York State Geological Association, Geology Department, Newark College of Arts and Sciences, Rutgers University, Newark, New Jersey, pp. 2-39, (1980).

2.5-113 Potomac Electric Power Company, "Geologic Investigation of the Stafford Fault Zone", Report by Dames and Moore dated June 14, 1976; Docket No. 50-448/449, (July 28, 1976).

CHAPTER 02 2.5-105 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-114 D. C. Prowell, B. J. O'Connor, and Meyer Rubin, "Preliminary Evidence for Holocene Movement along the Belair Fault Zone near Augusta, Georgia", USGS, Open File Report 75-680, (1975).

2.5-115 D. C. Prowell and B. J. O'Connor, "Belair Fault Zone: Evidence of Tertiary Fault Displacement in Eastern Georgia", Geology, 6, pp. 681-684, (1978).

2.5-116 D. W. Rankin, "The Continental Margin of Eastern North America in the Southern Appalachians: The Opening and Closing of the Proto-Atlantic Ocean", American Journal of Science, 275-A, pp. 298-236, (1975) 2.5-117 D. W. Rankin, "Appalachian Salients and Recesses: Late Precambrian Continental Breakup and the Opening of the Iapetus Ocean", Journal of Geophysical Research, 81-(32), pp. 5605-5617, (1976).

2.5-118 D. W. Rankin, Peter Popenoe and K. D. Klitgord, "The Tectonic Setting of Charleston, South Carolina", Geological Society of America, Abstracts with Programs, 10-(4), p. 195, (1978).

2.5-119 N. M. Ratcliffe, "Brittle Faults (Ramapo Fault) and Phyllonitic Ductile Shear Zones in the Basement Rocks of the Ramapo Seismic Zone, New York and New Jersey, and Their Relationship to Current Seismicity", in Warren Manspeizer, ed., Field Studies of New Jersey Geology and Guide to Field Trips, 52nd Annual Meeting of the New York State Geological Association, Geology Department, Newark College of Arts and Sciences, Rutgers University, Newark, New Jersey, pp. 278-311, (1980).

2.5-120 N. M. Ratcliffe, et. al, "Emplacement History and Tectonic Significance of the Courtlandt Complex, Related Plutons, and Dike Swarms in the Taconide Zone of Southeastern New York Based on K-Ar and Rb-Sr Investigations", American Journal of Science, 282, pp. 358-390, (1982).

2.5-121 F. S. Schilt, et al, "The Heterogeneity of the Continental Crust: Results from Deep Seismic Reflection Profiling Using the VIBROSEIS Technique", Reviews of Geophysics and Space Physics, 17, pp. 354-368, (1979).

2.5-122 F. S. Schilt, et al, "Subsurface Structure Near Charleston, South Carolina - Results of COCORP Reflection Profiling in the Atlantic Coastal Plain", in G. S. Gohn, ed.,

"Studies Related to the Charleston, South Carolina, Earthquake of 1886 - Tectonics and Seismicity (Collected Abstracts)", USGS, Open File Report 82-134, pp. 15-16, (1982).

2.5-123 Leonardo Seebar and J. G. Armbruster, "The 1886 Charleston Earthquake and the Appalachian Detachment", Journal of Geophysical Research, 86, pp. 7874-7894, (1981).

2.5-124 Leonardo Seeber, J. G. Armbruster and G. A. Bollinger, "Large-Scale Patterns of Seismicity before and after the 1886 South Carolina Earthquake", Geology, 10, pp.

382- 386, (1982).

CHAPTER 02 2.5-106 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-125 N. H. Sleep, "Thermal Contractions Beneath Atlantic Margins", in R. A. Scrutton, ed., "Dynamics of Passive Margins", Geodynamics Series 6, American Geophysical Union, (1982).

2.5-126 M. S. Steckler and A. B. Watts, "Subsidence History and Tectonic Evolution of Atlantic-Type Continental Margins", in R. A. Scrutton, ed., "Dynamics of Passive Margins", Geodynamics Series 6, American Geophysical Union, (1982).

2.5-127 D. E. Stephenson and H. R. Pratt, "In Situ Stress-Field in the Southeastern United States and Its Implications", Southeastern Geology, 22 (3), pp. 115-121, (1981).

2.5-128 L. R. Sykes, "Intraplate Seismicity, Reactivation of Pre-existing Zones of Weakness, Alkaline Magmatism, and Other Tectonism Post-dating Continental Fragmentation",

Review Geophysics and Space Physics, 16, pp. 621-688, (1978).

2.5-129 Pradeep Talwani, "An Internally Consistent Pattern of Seismicity near Charleston, South Carolina", Geology, (In Press).

2.5-130 Pradeep Talwani, B. Rastogi and D. Stevenson, "Induced Seismicity and Earthquake Prediction Studies in South Carolina", Tenth Technical Report, sponsored by the USGS, p. 212, (1980).

2.5-131 A. C. Tarr, et al, "Results of Recent South Carolina Seismological Studies",

Seismological Society of America, Bulletin, 71, pp. 1883-1902, (1981).

2.5-132 A. C. Tarr and Susan Rhea, "Seismicity Near Charleston, South Carolina, March 1973 to December 1979", in G. S. Gohn, ed., "Studies Related to the Charleston, South Carolina, Earthquake of 1886 - Tectonics and Seismicity (Collected Abstracts)", USGS, Open File Report 82-134, pp. 32-33, (1982).

2.5-133 S. R. Taylor and M. N. Toksoz, "Crust and Upper-Mantle Velocity Structure in the Appalachian Orogenic Belt: Implications for Tectonic Structure", Geological Society of America Bulletin, 93 (4), pp. 315-329, (1982).

2.5-134 C. M. Wentworth and Marcia Mergner-Keefer, "Reverse Faulting Along the Eastern Seaboard and the Potential for Large Earthquakes", in J. E. Beavers, ed.,

"Earthquakes and Earthquake Engineering: the Eastern United States", Volume 1, Ann Arbor Science Publishers, Inc., Ann Arbor, Michigan, pp. 109-128, (1981).

2.5-135 C. M. Wentworth of Marcia Mergner-Keefer, "Regenerate Faults of Small Cenozoic Offset as Probable Earthquake Sources in the Southeastern United States", USGS, Open File Report,81-356, (1981).

2.5-136 Harold Williams, et al, "Comments and Replies on 'Thin- Skinned Tectonics in the Crystalline Southern Appalachians; COCORP Seismic-Reflection Profiling of the Blue Ridge and Piedmont' and 'Sequential Development of the Appalachian Orogen Above a Master Decollement - A Hypothesis'", Geology, 8, pp. 211-216, (1980).

CHAPTER 02 2.5-107 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-137 Jih-Ping Yang and Y. P. Aggarwal, "Seismotectonics of Northeastern United States and Adjacent Canada", Journal of Geophysical Research, 86 (B6), pp. 4981-4998, (1981).

2.5-138 E-An Zen, "An Alternative Model for the Development of the Southern Appalachian Piedmont", American Journal of Science, 281, pp. 1153-1163, (1981).

2.5-139 M. D. Zoback, et al, "Normal Faulting and In Situ Stress in the South Carolina Coastal Plain near Charleston", Geology, 6, pp. 147-152, (1978).

2.5-140 M. D. Zoback and M. L. Zoback, "State of Stress and Intraplate Earthquakes in the United States", Science, 213 (3), pp.96-104, (1981).

2.5-141 Dames & Moore, Report, "Laboratory Test Specifications for Geotechnical Studies at Limerick Generating Station", (December 29, 1982).

2.5-142 A. W. Bishop and D. J. Henkel, "The Measurement of Soil Properties in the Triaxial Test", Edward Arnold LTD Publishers, London, U.K., pp. 122-131, (1957).

2.5-143 P.B. Schnable, J. Lysmer, and H.B. Seed, "SHAKE, A Computer Program for Earthquake Response Analysis of Horizontally Layered Sites", Earthquake Engineering Research Center Report EERC 72-12, University of California, Berkeley, California, (December 1972).

2.5-144 H.B. Seed, and I.M. Idriss, "Soil Moduli and Damping Factors for Dynamic Response Analyses", Earthquake Engineering Research Center Report EERC 70-10, University of California, Berkeley, California, (December 1970).

2.5-145 Bechtel Power Corporation, Report, "Spray Pond Seepage Test Report for the Limerick Generating Station Units 1 and 2", (October 1983).

2.5-146 R.E. Bischke, "The Abington-Cheltenham, PA. Earthquake Sequence of March-May 1980", Pennsylvania Geological Survey, V. II, pp. 10-13, (1980).

2.5-147 R.E. Bischke, "The Cornwells Heights, Pa.-Burlington, N.J. Earthquake of April 12, 1982", submitted for publication, Pennsylvania Geological Survey, (1982).

2.5-148 M.L. Sbar, et al. "The Delaware-New Jersey Earthquake of February 28, 1973",

Bulletin, Seismological Society of America, V. 65, N. 1, pp. 85-92, (1975).

2.5-149 N. Spoljaric, "Normal Faults in Basement Rocks of the Northern Coastal Plain, Delaware", Bulletin, Geological Society of America, V. 84, pp. 2781-2784, (1973).

2.5-150 R.B. Mixon, "The Thornburg scarp: A late Tertiary marine shoreline across the Stafford fault system", in R.B. Mixon, and W.L. Newell, op. cit., pp. 41-43, (1978).

2.5-151 V.N. Seiders, and R.B. Mixon, "Geologic map of the Occoquan quadrangle and part of the Fort Belvoir quadrangle, Prince William and Fairfax Counties, Virginia";

USGS Miscellaneous Investigation Map I-1175, (1981).

CHAPTER 02 2.5-108 REV. 13, SEPTEMBER 2006

LGS UFSAR 2.5-152 S.T. Algermissen, and D.M. Perkins, "A Probabilistic Estimate of Maximum Acceleration in Rock in the Contiguous United States", USGS Open File Report 76-416, (1976).

2.5-153 N.M. Ambraseys, and A.J. Hendron, Jr., "Dynamic Behavior of Rock Masses", in K.G. Stagg and O.C. Zienkiewicz, eds., Rock Mechanics in Engineering Practice, Wiley, pp. 203-236, (1968).

2.5-154 Dames and Moore, "Report, Site Environmental Studies, Limerick Generating Station", Limerick Township, Pennsylvania, submitted to PECo, (July 31, 1970).

2.5-155 W.J. Hall, N.M. Newmark and A.J. Hendron, Jr., "Classification, Engineering Properties and Exploration of Soils, Intact Rock and In Situ Rock Masses", WASH-1301, prepared for AEC, (May 1974).

2.5-156 N.M Newmark, "Development of Criteria for Seismic Review of Selected Nuclear Power Plants", NUREG/CR-0098, (1978).

2.5-157 J.D. Raphael, and R.E. Goodman, "Strength and Deformability of Highly Fractured Rock", Journal of Geotechnical Engineering Division, ASCE, pp. 1285-1300, (November 1979).

2.5-158 P. Schnable, H.B. Seed, and J. Lysmer, "Modification of Seismograph Records for Effects of Local Soil Condition", EERC 71-8, (December 1971).

2.5-159 M.L. Silver, and H.B. Seed, "Deformation Characteristics of Sands under Cyclic Loading", Journal of Soil Mechanics and Foundations Division, ASCE, p. 1081, (August 1971).

2.5-160 Seed, H.B., and Goodman, R.E., "Earthquake Stability of Slopes of Cohesionless Soils", Journal of Soil Mechanics and Foundation Division, ASCE, Vol. 90, No.

SM6, pp. 43-72, (November 1964).

CHAPTER 02 2.5-109 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-1 MODIFIED MERCALLI INTENSITY SCALE OF 1931

[Abridged]

I. Not felt, except by a very few under especially favorable circumstances. (I Rossi-Forel scale.)

II. Felt only by a few persons at rest, especially on upper floors of buildings. Delicately suspended objects may swing. (I to II Rossi-Forel scale.)

III. Felt quite noticeably indoors, especially on upper floors of buildings, but many people do not recognize it as an earthquake. Standing motor cars may rock slightly. Vibration like passing of truck. Duration estimated. (III Rossi-Forel scale.)

IV. During the day felt indoors by many, outdoors by few. At night some awakened.

Dishes, windows, doors disturbed; walls make cracking sound. Sensation like heavy truck striking building. Standing motor cars rock noticeably. (IV to V Rossi-Forel scale.)

V. Felt by nearly everyone; many awakened. Some dishes, windows, etc., broken; a few instances of cracked plaster; unstable objects overturned. Disturbance of trees, poles, and other tall objects sometimes noticed. Pendulum clocks may stop. (V to VI Rossi-Forel scale.)

VI. Felt by all; many frightened and run outdoors. Some heavy furniture moved; a few instances of fallen plaster or damaged chimneys. Damage slight. (VI to VII Rossi-Forel scale.)

VII. Everybody runs outdoors. Damage negligible in buildings of good design and construction; slight to moderate in well-built ordinary structures; considerable in poorly built or badly designed structures; some chimneys broken. Noticed by persons driving motor cars. (VIII Rossi-Forel scale.)

VIII. Damage slight in specially designed structures; considerable in ordinary substantial buildings, with partial collapse; great in poorly built structures. Panel walls thrown out of frame structures. Fall of chimneys, factory stacks, columns, monuments, walls. Heavy furniture overturned. Sand and mud ejected in small amounts.

Changes in well water. Disturbs persons driving motor cars. (VIII+ to IX Rossi-Forel scale.)

CHAPTER 02 2.5-110 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-1 (Cont'd)

IX. Damage considerable in specially designed structures; well-designed frame structures thrown out of plumb; great in substantial buildings, with partial collapse.

Buildings shifted off foundations. Ground cracked conspicuously. Underground pipes broken. (IX+ Rossi-Forel scale.)

X. Some well-built wooden structures destroyed; most masonry and frame structures destroyed with foundations; ground badly cracked. Rails bent. Landslides considerable from river banks and steep slopes. Shifted sand and mud. Water splashed (slopped) over banks. (X Rossi-Forel scale.)

XI. Few, if any, (masonry) structures remain standing. Bridges destroyed. Broad fissures in ground. Underground pipe lines completely out of service. Earth slumps and land slips in soft ground. Rails bent greatly.

XII. Damage total. Waves seen on ground surfaces. Lines of sight and level distorted.

Objects thrown upward into the air.

CHAPTER 02 2.5-111 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-2 (1)

EARTHQUAKES WITH EPICENTERS WITHIN APPROXIMATELY 200 MILES OF THE SITE DISTANCE FROM SITE(3)

YEAR DAY LOCAL TIME INTENSITY(2) LOCATION N. LAT. W. LONG. (mi.)

1737 Dec. 18 2230-2300 VII Near New York City 40.5 74 85 1758 Apr. 24 2130 - Annapolis, Md. 39 76.5 98 1763 Oct. 30 1625 IV-V New Jersey - - -

1773 Oct. 27 0000-0100 IV-V New York City - - -

1774 Feb. 21 1400 VI Eastern Va. 37.5 77.5 214 1774 Feb. 22 PM V-VI Eastern Va. 37.5 77.5 214

-(4) 1775 July 6 1555 V Jessopborough, N.Y. - -

1775 July 12 1555 VI-VII S. of Lake George, N.Y. 43.25 73.75 229 1783 Nov. 29 2215 IV-V Philadelphia, Pa. - - -

1783 Nov. 29 2250 VI W. of New York City 41 74.5 78 1791 May 16 0800 VII Moodus, Conn. 41.5 72.5 183 1792 Aug. 28 2200 IV-V Near E. Haddam, Conn. 41.5 72.5 183 1793 Jan. 11 0800 IV-V Near E. Haddam, Conn. 41.5 72.5 183 1794 Mar. 6 1400 IV-V Near E. Haddam, Conn. 41.5 72.5 183 1796 Dec. 24 1300 V-VI New London, Conn. 41.5 72 207 1800 Mar. 17 - - Philadelphia, Penn. 39.75 75.25 37 1800 Nov. 29 - - Philadelphia, Penn. 39.75 75.25 37 1802 Aug. 23 0500 V Charlotte, Va. 37.6 77.4 206 1805 Aug. 11 1900 IV-V Near E. Haddam, Conn. 41.5 72.5 183 1827 Aug. 23 - IV-V Near New London, Conn. 41.5 72.25 195 CHAPTER 02 2.5-112 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-2 (Cont'd)

DISTANCE FROM (3)

SITE (2)

YEAR DAY LOCAL TIME INTENSITY LOCATION N. LAT. W. LONG. (mi.)

1833 Aug. 27 0600 V-VI Eastern Va. 37.75 78 214 1837 Apr. 12 - V Hartford, Conn. 41.75 72.75 181 1840 Jan. 16 2000 V-VI Near Utica, N.Y. 43 75 194 1840 Aug. 9 1530 V Conn. River Valley 41.5 72.75 172 1840 Nov. 11 - - Philadelphia, Penn. 39.75 75.25 37 1845 Oct. 26 - V-VI W. Hudson River Valley, N.Y. 42.5 73.75 183 1847 Sep. 2 - V Probably near N.Y. City 40.5 74 85 1847 Sep. 29 - V New York City - - -

1848 Sep. 9 2200 V Probably near N.Y. City 40.5 74 85 1852 Nov. 2 1835 VI Eastern Va. 37.75 78 214 1855 Feb. 6 2330 V-VI Hudson River Valley, N.Y. 42 74 148 1858 June 30 2245 IV-V Near New Haven, Conn. 41.25 73 152 1861 Mar. 5 1200 IV-V Bloomfield, N.J. 40.8 74.2 82 1871 Oct. 9 0940 VII Wilmington, Del. 39.75 75.5 33 1872 July 11 0525 IV-V Westchester Co., N.Y. 41 73.75 110 1874 Dec. 10 2225 V-VI Westchester Co., N.Y. 41 73.75 110 1875 July 28 0410 V NW Conn. 41.75 73.25 161 1875 Dec. 22 2345 VII Near Richmond, Va. 37.5 77.5 214 1877 Jan. 3 2300 IV-V Emmittsburg, MD. 39.5 77.3 104 1877 Aug. 31 0959 IV-V Laurel, MD. 39.2 77.2 112 1877 Sep. 10 0959 IV-V Delaware Valley 40.1 74.9 37 CHAPTER 02 2.5-113 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-2 (Cont'd)

DISTANCE FROM (3)

SITE (2)

YEAR DAY LOCAL TIME INTENSITY LOCATION N. LAT. W. LONG. (mi.)

1878 Feb. 5 1120 V New York City 40.5 74 85 1878 Oct. 4 0230 V Hudson River Valley, N.Y. 41.5 74 85 1879 Mar. 25 1930 IV-V Delaware Valley 39.75 75.5 33 1881 Feb. 4 0430 IV-V Marlborough, N.Y. 41.6 74.0 125 1881 Apr. 21 1130 V-VI Port Jefferson, N.Y. 40.9 73.1 138 1882 Apr. 2 evening V New Market, Va. 38.7 78.7 197 1883 Mar. 11 1857 IV-V Harford County, Md. 39.5 76.4 66 1883 Mar. 12 0000-0100 IV-V Harford County, Md. 39.5 76.4 66 1884 May 31 - V Allentown, Penn. 40.6 75.5 20 1884 Aug. 10 1407 VII Near New York City 40.5 74 85 1885 Jan. 2 2116 V Md. and Va. 39.2 77.5 124 1885 Mar. 8 - IV-V Lancaster, Pa. 40.0 76.3 41 1886 Jun. 12 0005 IV-V Asbury Park, N.J. 40.5 74 85 1887 Jan. 2 2330 V Baltimore, MD. 39.8 77.0 81 1887 Mar. 2 1613 IV-V Seaford, De. 40.6 73.3 123 1889 Mar. 8 1840 VI SE Penn. 40 76.75 63 1893 Mar. 9 0030 V Near New York City 40.5 74 85 1893 Mar. 14 1505 V Brooklyn, N.Y. 40.7 73.9 94 1894 Jan. 25 2040 IV-V Annapolis, MD. 39.0 76.5 98 1895 Sep. 1 0609 VI Near High Bridge, N.J. 40.7 74.8 53 1897 Nov. 27 1556 V Ashland, Va. 37.7 77.5 202 CHAPTER 02 2.5-114 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-2 (Cont'd)

DISTANCE FROM (3)

SITE (2)

YEAR DAY LOCAL TIME INTENSITY LOCATION N. LAT. W. LONG. (mi.)

1897 Dec. 18 1845 V Near Ashland, Va. 37.7 77.5 202 1899 May 16 2015 V Conn. River Valley 41.5 72.5 183 1900 Apr. 28 1907 IV-V Camden, N.J. 39.9 75.1 34 1906 May 8 1241 V Seaford, Del. 38.7 75.7 105 1906 May 11 - IV-V Rockaway, N.Y. 40.6 73.8 97 1906 May 29 evening IV-V Reading, Pa. 40.3 75.9 18 1907 Jan. 24 0630 IV-V Near Schenectady, N.Y. 42.75 74 192 1907 Feb. 11 0822 VI Near Arvonia, Va. 37.7 78.3 227 1908 Feb. 5 0320 IV-V Housatonic Valley, Conn. 41.5 73.25 150 1908 May 31 1242 VI Allentown, Penn. 40.6 75.5 26 1908 Aug. 23 0430 V Near Powhatan, Va. 37.5 77.9 225 1909 Apr. 2 0225 V-VI Near border of Va. 39.4 78.0 140 1910 Apr. 23 0408 IV-V Atlantic City, N.J. 39.3 74.4 90 1912 Nov. 6 1530 V Atlantic City, N.J. 39.2 74.4 95 1914 Mar. 6 evening IV-V Newark, N.J. 40.8 74.1 87 1914 Mar. 25 0200 IV-V Five Mile Beach, N.J. 39.2 74.2 102 1916 Feb. 2 2326 V Mohawk Valley, N.Y. 43 74 208 1916 June 8 1615 IV-V Westchester Co., N.Y. 41 73.75 110 1918 Apr. 9 2109 V Near Luray, Va. 38.75 78.5 186 1919 Sep. 5 2146 VI Near Front Royal, Va. 38.75 78.25 175 1920 Jul. 25 early morn IV-V Lauray, Va. 38.7 78.4 184 CHAPTER 02 2.5-115 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-2 (Cont'd)

DISTANCE FROM (3)

SITE (2)

YEAR DAY LOCAL TIME INTENSITY LOCATION N. LAT. W. LONG. (mi.)

1921 Jan. 26 1840 V Moorestown, N.J. 40.0 75.0 34 1925 Jul. 14 0400 IV-V Richmond, Va. 37.6 77.4 206 1925 Nov. 14 0804 VI Near Moodus, Conn. 41.5 72.5 183 1926 May 11 2230 V New Rochelle, N.Y. 40.9 73.9 100 1927 June 1 0720 VII New Jersey Coast 40.3 74.0 83 1928 Jan. 13 1450 IV-V Block Island, R.I. 41 71.5 220 1929 Aug. 12 0625 VIII Near Attica, N.Y. 42.9 78.3 232 1929 Dec. 26 2156 VI Near Charlottesville, Va. 38 78.5 219 1933 Jan. 24 2100 V Central N.J. 40.1 74.5 58 1938 July 15 1745 VI S. Blair Co., Penn. 40.4 78.2 138 1938 Aug. 22 2236 V Central N.J. 40.1 74.5 58 1939 Nov. 14 2154 V Salem Co., N.J. 39.6 75.2 48 1940 Mar. 25 2228 IV-V Near Woodstock, Va. 38.9 78.6 185 1947 Jan. 4 1351 V Near Greenwich, Conn. 41 73.5 121 1949 May 8 0601 IV-V Near Powhatan, Va. 37.5 78 228 1951 Sep. 3 2026 V Rockland Co., N.Y. 41.2 74.1 103 1952 Aug. 24 1907 V Mohawk Valley, N.Y. 43.0 74.5 199 1952 Oct. 8 1640 V Poughkeepsie, N.Y. 41.7 74.0 131 1953 Mar. 27 0350 V Stamford, Conn. 41.1 73.5 125 1954 Jan. 7 0225 VI Sinking Spring, Penn. 40.3 76.0 23 1954 Feb. 21 1500 VII Wilkes-Barre, Penn. 41.2 75.9 69 CHAPTER 02 2.5-116 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-2 (Cont'd)

DISTANCE FROM (3)

SITE (2)

YEAR DAY LOCAL TIME INTENSITY LOCATION N. LAT. W. LONG. (mi.)

1954 Feb. 23 2255 VI Wilkes-Barre, Penn. 41.2 75.9 69 1955 Jan. 21 0304 V Malta, N.Y. 43.0 73.8 212 1957 Mar. 23 1403 VI West-Central N.J. 40.6 74.8 49 1961 Sep. 14 2117 V Lehigh Valley, Penn. 40.6 75.4 28 1961 Dec. 27 1206 V Penn.-N.J. border 40.1 74.8 42 1964 May 12 0445 VI Near Cornwall, Penn. 40.2 76.5 48 1964 Nov. 17 1208 V Westchester Co., N.Y. 41.2 73.7 119 1966 Jan. 1 0824 VI Near Attica, N.Y. 42.8 78.2 223 1966 May 31 0119 IV-V Eastern Va. 37.6 78.0 223 1967 June 13 1409 VI Near Attica, N.Y. 42.9 78.2 229 1967 Nov. 22 1710 V Westchester Co., N.Y. 41.1 73.8 111 1968 Nov. 3 0334 V S. Conn. 41.3 72.5 177 1968 Dec. 10 0413 V N.J. 39.7 74.6 63 1971 Sep. 11 1907 V Va. 38.1 77.4 176 1972 Sep. 5 1100 V Richmond, Va. 37.6 77.7 214 1973 Feb. 28 0322 V N.J. 39.7 75.4 34 1974 June 7 1446 VI Wappingers Falls, N.Y. 41.6 73.9 129 1976 Mar. 11 1607 VI Northeastern N.J. 40.84 74.05 91 1977 Feb. 27 1506 V Va. 37.9 78.6 229 1978 Jun. 30 1513 IV-V between Mahwah & 41.08 74.20 93 Oakland, N.J.

1978 Jul. 16 0139 V Lancaster, Pa. 39.90 76.22 41 CHAPTER 02 2.5-117 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-2 (Cont'd)

DISTANCE FROM (3)

SITE (2)

YEAR DAY LOCAL TIME INTENSITY LOCATION N. LAT. W. LONG. (mi.)

1978 Oct. 6 1425 VI Lancaster, Pa. 40.05 76.09 30 1979 Jan. 30 1130 V Cheesequake, N.J. 40.321 74.264 70 1979 Mar. 9 2349 V Bernardsville, N.J. 40.721 74.504 66 1979 Dec. 30 0915 V Mt. Kisco, N.J. 41.140 73.690 117 1980 Mar. 5 1206 IV-V Abington, Pa. 40.17 75.07 27 1980 Mar. 11 0100 IV-V Abington, Pa. 40.15 75.09 26 (1)

Only earthquakes with intensities greater than IV on the Modified Mercalli Intensity Scale are included in this table.

(2)

Intensity based on Modified Mercalli Intensity Scale of 1931.

(3)

Calculated using site coordinates of 40.23 N. latitude, 75.58 W. longitude.

(4)

This event may not have occurred within 200 miles of the site. Jessopborough may correspond to Jessup Falls in Warren Co., formerly Washington Co. in the 1700s.

CHAPTER 02 2.5-118 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-3 UNCONFINED COMPRESSION TESTS ON ROCK SAMPLES UNCONFINED BORING ELEVATION COMPRESSIVE DENSITY NO. (feet) STRENGTH (psi) (lbs/ft3) 2 139.0 17,490 151 2 91.5 15,000 149 2 88.0 17,850 144 9 148.0 6,370 140 9 106.5 14,520 153 20 175.0 19,630 159 22 190.5 20,110 162 22 179.0 14,570 155 25A 204.5 16,560 155 25A 201.5 13,670 150 120 124.0 19,290 153 120 122.5 22,140 161 120 118.0 24,540 162 120 76.0 10,950 144 121 97.0 12,100 145 122 109.5 8,340 150 CHAPTER 02 2.5-119 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-4

SUMMARY

OF ENGINEERING PROPERTIES OF IN SITU SOIL AT SPRAY POND PROPERTIES RANGE AVERAGE In situ moisture content (%) 11.9 - 38.7 21.7 3

In situ total unit weight (lb/ft ) 98.6 - 137.2 122.0 Grain-size distribution:

Medium grain-size, D50 (mm) 0.006 - 4.4 0.32 Percent by weight passing 15 - 100 72 No. 200 sieve Atterberg limits:

Liquid limit 27 - 51 37 Plasticity index 2 - 27 15 Effective consolidated-undrained shear strength:

c (psf) - 0 (deg) - 33.5 Undrained shear strength, s (psf) - 1.2 P o(1)

(2)

Dynamic shear strength - 0.61o Cyclic stress ratio - 0.61 Standard penetration resistance 7 - 86 36 (blows/ft)

Specific gravity 2.70 - 2.80 2.76 CHAPTER 02 2.5-120 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-4 (Cont'd)

OTHER THAN AT SPRAY POND SITE PROPERTIES RANGE AVERAGE In situ Moisture Content (%) 8.3 - 21.3 13.4 In situ Total Unit Weight (pcf) 126.0-140.6 132.8 Grain-Size Distribution Gravel (retain on No. 4 sieve) 4.0 - 25.0 14.0 Sand (passing No. 4 and retain on 12.0 - 54.0 27.0 No. 200)

Silt/Clay (passing No. 200 sieve) 26.0 - 82.0 59.0 Atterberg Limits Liquid Limit 18.0 - 37.0 25.0 Plasticity Index 3.0 - 17.0 8.0 Standard Penetration Resistance 2 -.98 .42 (blows/ft) o Total Undrained Shear Strength c = 3.0 ksf, = 18 Effective Shear Strength c = 0, = 26.5o (1)

P o = mean effective principal stress, ( 1 + 3)/2 (2) o = effective overburden pressure CHAPTER 02 2.5-121 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-5

SUMMARY

OF DESIGN PARAMETERS OF SOILS AND SOIL-BENTONITE MIXTURES IN SITU SOIL-BENTONITE PARAMETER SOIL SOIL COVER(1) LINER(1)

Unit weight (lb/ft3)

Total 122.0 120.5 113.0 Saturated 126.4 123.8 119.0 Effective shear strength parameters c (psf) 0 0 0 o (deg.) 33.5 33.5 29.0 Undrained shear strength (psf)(2) 1.2 P 1.2 P 0.95 P Dynamic shear strength(3) 0.61 o 0.61 o = 20o c = 225 lb/ft2 Permeability (ft/yr) - - 1.0 (1)

Both soil cover and soil-bentonite liner are compacted to 95% of maximum dry density determined in accordance with ASTM D698.

P = mean effective principal stress, ( 1 + 3)/2 (2)

(3) o= effective overburden pressure CHAPTER 02 2.5-122 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-6 SOIL TEST RESULTS

SUMMARY

(SPRAY POND TEST PITS)

MECHANICAL ATTERBERG NATURAL ANALYSIS LIMITS MOISTURE TEST PIT DEPTH LABORATORY GRAVEL SANDS FINES CONTENT NO. (FEET) CLASS. (%) (%) (%) LL PL PI (%)

TP-1 8.0 CL 11 13 76 32 20 12 20.1 TP-2 5.0 CL 12 24 64 34 21 13 20.9 TP-3 5.0 CL 20 19 61 37 21 16 21.1 TP-4 5.0 CL 5 7 88 39 21 18 21.9 TP-5 5.0 SC 31 22 47 48 25 23 29.5 TP-6 4.0 CL 10 25 65 44 25 19 25.1 TP-7 5.0 CL 6 13 81 40 22 18 21.8 TP-8 4.5 GC - - - - - - 24.8 TP-9 6.0 CL 5 12 83 41 23 18 26.7 TP-10 4.0 CL 3 9 88 40 21 19 20.7 TP-11 7.0 ML 13 14 73 45 28 17 30.7 TP-12 9.0 SC 45 14 41 35 20 15 15.0 TP-13 7.0 ML 5 11 84 41 27 14 28.5 CHAPTER 02 2.5-123 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-7 RESULTS OF STABILITY ANALYSIS (2H:1V ROCK SLOPE - SPRAY POND)(2)

Minimum Water level el of Water Factor of Safety(1) Factor of Safety(1) Minimum in Pond Table in Slope Loading without shotcrete with shotcrete Empty - rapid 251 static 1.5 5.7 drawdown Full - el 251' 251 (saturated slope) static >10 >10 Full - el 251' 251 (saturated slope) SSE 1.5 4.5 Full - el 251' 241 (drained slope) static >10 >10 Full - el 251' 241 (drained slope) SSE 6.0 >10 (1)

These minimum factors of safety against sliding include the effect of the rock bolts. Also, the factors apply only to the least stable bedding-planes of the 2:1 rock slope, where failure of such planes would occur at the toe of the slope. The other portions of the 2:1 slope as well as all other rock slopes have factors of safety exceeding the above values.

(2)

The applicable slope is shown in profile as section F, Figure 2.5-49.

DESIGN PROPERTIES USED IN ANALYSIS Unit weight of soi 122 pcf (Table 2.5-5)

Unit weight of roc 152 pcf (Table 2.5-11)

DIP of failure plane 11o Friction angle (along failure plane) 20o Cohesion C (along failure plane) 0 psi SSE loading (parallel to failure plane) 15% of total static weight above failure plane CHAPTER 02 2.5-124 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-8 EARTHQUAKES WITH EPICENTERS WITHIN APPROXIMATELY 50 MILES OF THE SITE DISTANCE FROM (2)

SITE (1)

YEAR DAY LOCAL TIME INTENSITY LOCATION N. LAT. W. LONG. (mi.)

1755 Nov 26 2000 - Philadelphia, Pa. - - -

1763 Mar 22 - - Delaware Co., Pa. - - -

1763 Oct 13 - - Philadelphia, Pa. - - -

(3) 1763 Oct 30 1625 IV-V New Jersey - - -

1772 Apr 25 0800-0900 IV Cohantle, N.J. - - -(3)(4) 1777 Nov 22 - - Philadelphia, Pa. - - -

1777 Nov 23 0700 - Delaware Co., Pa. - - -

1783 Nov 29 2215 IV-V Philadelphia, Pa. - - -

1783 Nov 30 0100-0200? IV Pennsylvania 40.0 75.1 30 1799 Mar 17 - - Philadelphia, Pa. - - -

1800 Mar 17 - - Philadelphia, Pa. 39.75 75.25 37 1800 Nov 29 - IV Philadelphia, Pa. 39.75 75.25 37 1801 Nov 12 - - Philadelphia, Pa. - - -

1811 Dec 08 2000 III Pennsylvania 39.8 75.5 30 1840 Nov 11 - IV Philadelphia, Pa. 39.75 75.25 37 1840 Nov 14 - IV Philadelphia, Pa. 39.7 75.2 36 1870 Oct 12 - III Wilmington, De. 39.8 75.5 30 1871 Oct 09 0940 VII Wilmington, De. 39.75 75.5 33 1877 Aug 10 - III near Trenton, N.J. 40.1 74.8 42 1877 Sep 10 0959 IV-V Delaware Valley 40.1 74.9 37 1879 Mar 25 1930 IV-V Delaware Valley 39.75 75.5 33 1884 May 31 - V Allentown, Pa. 40.6 75.5 26 1885 Jan 15 0410 III Pennsylvania 40.3 76.3 38 1885 Mar 08 - IV-V Lancaster, Pa. 40.0 76.3 41 1889 Mar 08 1840 V Pennsylvania 40 76 27 1889 Mar 09 0500 - Pennsylvania 40 76 27 1892 Oct 10 - III Wilmington, De. 39.8 75.5 30 1893 Apr 26 morning III Lancaster, Pa. 40.1 76.2 34 1895 Sep 01 0609 VI near High Bridge, N.J. 40.7 74.8 53 1895 Nov 20 0300 IV Claymont, De. 39.8 75.6 30 1900 Mar 17 - III Philadelphia, Pa. 39.9 75.2 30 1900 Apr 28 1907 IV-V Camden, N.J. 39.9 75.1 34 1900 Nov 29 - III Philadelphia, Pa. 39.9 75.2 30 1906 May 28 1730 III Pennsylvania 40.2 75.8 12 1906 May 29 evening IV-V Reading, Pa. 40.3 75.9 18 1908 May 31 1242 VI Allentown, Pa. 40.6 75.5 26 1909 Feb 06 early morning III Trenton, N.J. 40.2 74.7 46 1921 Jan 26 1840 V Moorestown, N.J. 40.0 75.0 34 1933 Jan 24 2100 V central New Jersey 40.1 74.5 58 CHAPTER 02 2.5-125 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-8 (Cont'd)

DISTANCE FROM (2)

SITE (1)

YEAR DAY LOCAL TIME INTENSITY LOCATION N. LAT. W. LONG. (mi.)

1937 Jun 08 1904 - Reading, Pa. 40.3 75.9 18 1938 Aug 22 2236 V central New Jersey 40.1 74.5 58 1939 Apr 01 2200 - Lancaster, Pa. 40.0 76.3 41 1939 Nov 14 2154 V Salem Co., N.J. 39 6 75.2 48 1944 Jan 08 - - Wilmington, De. 39.8 75.5 30 1949 Oct 16 1833 - Massena, N.Y. 40.4 74.8 43 1951 Nov 23 0145 - Allentown, Pa. 40.7 75.5 33 1954 Jan 07 0225 VI Sinking Spring, Pa. 40.3 76.0 23 1954 Jan 23 2230 - Sinking Spring, Pa. 40.3 76.0 23 1954 Aug 10 2240 - Sinking Spring, Pa. 40.3 76.0 23 1955 Jan 19 2200 IV Berks Co., Pa. 40.9 75.9 49 1957 Mar 23 1403 VI west-central N.J. 40.6 74.8 49 1961 Sep 14 2117 V Lehigh Valley, Pa. 40.6 75.4 28 1961 Dec 27 1206 V Pa. - N.J. border 40.1 74.8 42 1964 May 12 0145 VI(4.5) near Cornwall, Pa. 40.2 76.5 48 1972 Dec 07 2200 IV-V Pennsylvania 40.1 76.2 34 1973 Feb 28 0322 V-VI New Jersey 39.7 75.4 34 1974 Apr 27 0945 (3.0) Pennsylvania 41.00 75.96 57 1977 Jun 10 0748 (1.1) near High Bridge, N.J. 40.70 74.89 49 1977 Jul 02 0613 (2.3) Hampton, N.J. 40.70 74.93 47 1978 Jul 16 0139 V(3.3) Lancaster, Pa. 39.90 76.22 41 1978 Oct 06 1425 VI(3.0) Lancaster, Pa. 40.05 76.09 30 1979 Feb 23 0523 IV New Jersey 40.80 74.81 56 1980 Mar 02 0654 (2.8) Abington, Pa. 40.21 75.08 26 1980 Mar 05 1206 IV-V(3.5) Abington, Pa. 40.17 75.07 27 1980 Mar 05 1220 (3.1) Abington, Pa. 40.18 75.07 27 1980 Mar 11 0100 IV-V(3.7) Abington, Pa. 40.15 75.09 26 1980 Mar 11 1116 (2.8) Abington, Pa. 40.25 74.99 31 Middleton (Abington),

1980 Apr 10 1210 (2.8) Pa. 40.2 75.0 31 Newtown (Abington),

1980 Apr 16 1317 (3.2) Pa. 40.3 75.0 31 Langhorne (Abington),

1980 May 02 1023 (2.8) Pa. 40.16 74.99 32 1980 May 02 1402 (3.0) Jamison (Abington), Pa. 40.24 75.03 29 1980 Aug 30 0419 (3.0) Medford, N.J. 39.84 74.86 47 (1)

Intensity based on Modified Mercalli Intensity Scale of 1931. Values in parentheses are body wave or code length type magnitudes given for earthquake size comparison.

(2)

Calculated using site coordinates of 40.23 N. latitude, 75.58 W. longitude.

(3)

This event may not have occurred within 50 miles of the site.

(4)

Cohantle may be a misspelling of Cohansey, formerly either Bridgeton or Greenwich, N.J., approximately 60 miles from the site.

CHAPTER 02 2.5-126 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-9 DESIGN PARAMETERS OF SAFETY-RELATED STRUCTURES(1)(2)

Static Maximum Approx. Design Hydro-Safety-Related Basement Bearing Bearing static Discussion in Structure Dimension Elev. Pressure Pressure UFSAR Section Containment 100'-4" dia. 174' 8.3 ksf N/A 2.5.4.10.1.2 Reactor Enclosure 326'x137' (Units 1 & 2) 174' 31.4 ksf 1123 psf 2.5.4.10.1.2 Control Structure 132'x62' 178' 31.4 ksf 936 psf 2.5.4.10.1.2 Spray Pond 83'x22'-9" 257'-6" 14.1 ksf 1810 psf 2.5.4.10.1.3 Pumphouse 151'x22'-9" 235' Diesel Generator 273'x86' 190' 22.1 ksf Above 2.5.4.10.1.1 Enclosure (Units 1 or 2) ground-water level Valve Pit 7'x8' 198' 2.7 ksf Above 2.5.4.10.2.3 (on soil) (on fill) ground-water level Electrical +/- 4' width 6' 2.7 ksf Above 2.5.4.10.2.4 Duct Bank x variable Below ground-(portions on length Grade water soil level Diesel Oil 59'x123' 189'-9" 5.4 ksf Not req'd 2.5.4.10.1.4 Storage Tank (For all for a slab Foundation tanks) w/no walls (1)

Bearing capacities of rock material = 60 ksf (Table 2.5-11)

(2)

Bearing capacities of soil material = 6 ksf (Section 2.5.4.10.2.4)

CHAPTER 02 2.5-127 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-10 SOIL-RESPONSE STUDY

SUMMARY

OF PARAMETERS AND RESULTS DESCRIPTPION OF PARAMETERS CASE A CASE B CASE C STRATIGRAPHY Type 1 Fill Depth, ft 40 40 40 K2 140 100 180 Strain level, in/in 10-6 10-6 10-6 Total unit wt, pcf 140 140 140 At rest earth pressure coefficient 0.7 0.7 0.7 Depth, Centerline Pipe, ft 13 13 13 Rock Shear modulus, psf 1700x105 1700x105 1700x105

RESPONSE

Response at Centerline Pipe Peak acceleration, g 0.36 0.32 0.36 Peak particle velocity, ft/sec 0.87 0.76 0.83 Frequency of soil column, Hz 5.9 4.8 7.1 CHAPTER 02 2.5-128 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-11 REPRESENTATIVE ENGINEERING PROPERTIES (1)

OF SOUND FOUNDATION ROCK APPROXIMATE AVERAGE PROPERTY VALUE RANGE REFERENCE Unconfined compressive 6,370 -

strength 15,820 psi 24,540 psi Table 2.5-3 Allowable bearing pressure (2)

Normal load 420 psi -

Normal plus dynamic (2) load 625 psi -

140 - 162 3 3 Density 152 lbs/ft lbs/ft Table 2.5-3 Compressional wave 7,000 - Section velocity 12,500 fps 20,000 fps 2.5.4.4.1 (refraction)

Shear-wave velocity 5,800 - Section (refraction) 5,950 fps 6,100 fps 2.5.4.4.2 Poisson's ratio (2)

(measured) 0.30 -

Modulus of (3) elasticity 6

(dynamic) 3.0x10 psi - -

Modulus of (2) elasticity 5

(static) 6.9x10 psi - -

(3) 6 Shear modulus 1.2x10 psi - -

(1)

From measurements on unweathered rock in power block area.

(2)

Dames and Moore, Foundation Report dated October 5, 1970, and letter from Dames and Moore to PECo dated August 19, 1971.

(3)

Calculated from average shear-wave velocity and Poission's ratio given above.

CHAPTER 02 2.5-129 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-12 STATIC MODULI OF FOUNDATION ROCK(1)

PEAK COMPRESSIVE YOUNG'S BORING ELEVATION STRESS MODULUS (E)

NUMBER (ft) (psi) (psix106) 232 193.0 24,810 5.17 271 157.0 8,000 6.67 272 191.0 18,800 4.0 274 180.0 10,950 3.33 275 190.5 16,670 3.22 276 198.5 18,630 3.2 279 209.0 14,950 2.7 285 148.6 10,900 1.8 285 145.1 9,260 1.2 286 131.8 16,970 3.4 286 121.8 19,440 5.6 287 137.0 28,180 8.3 287 117.5 32,870 7.8 288 150.8 15,450 3.3 288 142.3 9,100 1.8 (1)

Source: From appendix, table II of Dames and Moore Foundation Report dated October 5, 1970.

CHAPTER 02 2.5-130 REV. 13, SEPTEMBER 2006

LGS UFSAR Table 2.5-13 SONIC TESTS ON INTACT CORES(1)

Compressional Depth Wave Velocity Boring No. (ft) (fps) 120 31.9 12,200 120 32.7 16,100 120 37.9 13,900 120 79.5 12,300 121 27.8 10,400 121 31.6 11,000 121 22.8 10,400 232 41 11,110 272 43 11,160 (1)

Source: From table 2.5-2 of Dames and Moore Site Environmental Studies dated July 31, 1970, and appendix, table 1 of Dames and Moore Foundation Report dated October 5, 1970.

CHAPTER 02 2.5-131 REV. 13, SEPTEMBER 2006