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Page 1 of 87 DSAR-APPENDIX C Foundation Studies Rev 0 Safety Classification: Usage Level:
Safety Information Change No.: EC 69283 Reason for Change: LIC-16-0074 certifies permanent removal of fuel from the FCS reactor core. This information Supports USAR 5.7 Piling. Since this information will no longer change, this appendix is being archived.
Preparer: J. Carlson Fort Calhoun Station ARCHIVED TEXT*
- USAR pages labeled as "ARCHIVED TEXT" are pages with text which is not revised or updated. Information on "ARCHIVED TEXT" pages is A) of historical nature significant to the original licensing basis of the plant OR B) not meaningful to update.
DSAR-Appendix C Information Use Page 2 of 87 Foundation Studies Rev.0 ARCHIVED TEXT*
REPORT FOUNDATION STUDIES FORT CALHOUN STATION - UNIT NO. 1 NEAR FORT CALHOUN, NEBRASKA OMAHA PUBLIC POWER DISTRICT DAMES AND MOORE APPLIED EARTH SCIENCES NY - 3 EO - 1 CH - 2 SF - 1 JOB NO. 4862-002-10
DSAR-Appendix C Information Use Page 3 of 87 Foundation Studies Rev.0 ARCHIVED TEXT*
DAMES & MOORE ATLANTA NEW YORK CONSULTING ENGINEERS IN THE APPLIED EARTH SCIENCE CHICAGO PORTLAND DENVER SALT LAKE CITY HONOLULU SAN FRANCISCO HOUSTON SEATTLE LOS ANGELES WASHINGTON, DC LONDON, ENGLAND MADRID, SPAIN TEHRAN. IRAN CANADA HALIFAX TORONTO SAINT JOHN. N.B.
100 CHURCH STREET NEW YORK.NEW YORK 10007 (212) 267-1810 PARTNERS: GARDNER M. REYNOLDS ROBERT M. PERRY JOSEPH A. FlSCHER ASSOCIATES: FRANCIS E. RANFT LOUIS I. STERN R. BRIAN ELLWOOO January 30, 1968 Omaha Public Power District 1623 Harney Street Omaha, Nebraska 68102 Attention: Mr. J. L. Wilkins, Assistant General Manager Gentlemen:
We submit herewith 125 Copies of our "Report, Foundation Studies, Fort Calhoun Station - Unit No. 1, Near Fort Calhoun, Nebraska, Omaha Public Power District."
The scope of this study was planned in cooperation with representatives of the Omaha Public Power District, Gibbs & Hill, Incorporated and Gibbs, Hill, Durham and Richardson. During the course of the work, numerous meetings were held with representatives of both the owner and consulting engineers to keep the interested parties aware of the progress of the work and of our preliminary conclusions and recommendations.
The site is underlain by loose to compact generally granular fluvial deposits overlying limestone bedrock, underlain, in turn, by other sedimentary rock strata. The limestone contains some small solution cavities and occasional open fissures.
A number of foundation alternates were discussed with Gibbs & Hill, Inc. and Omaha Public Power District. On the basis of operational requirements for planned floor grades and significant safety features, a pile foundation scheme is planned for support of the proposed facilities. The use of a pile foundation will necessitate exploration of the bedrock at each pile location, since the exact definition of every cavity and fissure by means of a normal test boring program is a virtual impossibility.
In addition, analyses indicate that the relatively loose upper sands may be subject to liquefaction under the maximum credible earthquake. Should production pile installation not induce sufficient compaction of these sands, a remedial compactive technique, such as vibroflotation, will be required.
CABLE ADDRESS- DAMEMORE
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We would like to express our appreciation to the Omaha Public Power District for the confidence they have expressed in Dames & Moore and to representatives of the power company for their assistance during the various phases of our studies.
Yours very truly, DAMES & MOORE Joseph A. Fischer JAF/deb (copies submitted)
DSAR-Appendix C Information Use Page 5 of 87 Foundation Studies Rev.0 ARCHIVED TEXT*
Table of Contents DESIGN CONSIDERATIONS.................................................................................................. 6 PURPOSES............................................................................................................................. 7 SCOPE OF WORK .................................................................................................................. 7 SITE CONDITIONS ............................................................................................................... 10 SURFACE FEATURES ......................................................................................................... 10 SUBSURFACE CONDITIONS............................................................................................... 10 DISCUSSION AND RECOMMENDATIONS.......................................................................... 13 FOUNDATION REQUIREMENTS ......................................................................................... 13 CORROSIVITY ...................................................................................................................... 22 FOUNDATION INSTALLATION ............................................................................................ 22 Class I Structures: ................................................................................................................. 22 Class II Structure: .................................................................................................................. 24 PILE LOAD TEST PROGRAM .............................................................................................. 24 APPENDIX ............................................................................................................................ 29 FIELD EXPLORATIONS AND LABORATORY TESTS ......................................................... 29 LABORATORY TESTS ......................................................................................................... 32
DSAR-Appendix C Information Use Page 6 of 87 Foundation Studies Rev.0 ARCHIVED TEXT*
REPORT FOUNDATION STUDIES FORT CALHOUN STATION - UNIT NO. 1 NEAR FORT CALHOUN, NEBRASKA OMAHA PUBLIC POWER DISTRICT INTRODUCTION GENERAL This report presents the result of foundation studies conducted for the proposed Fort Calhoun Station - Unit No. 1. The proposed nuclear power plant is to be constructed for the Omaha Public Power District adjacent to and southwest of the Missouri River, approximately four miles southeast of Blair, Nebraska. The location of the site is shown relative to the surrounding cultural and topographic features on the DESIGN CONSIDERATIONS The proposed nuclear generating station will be composed of both Class I and appurtenant facilities. The Class I Units are:
- 1) a reactor containment building;
- 2) an auxiliary building; and
- 3) a pumping station.
The major appurtenant facilities are a turbine-generator building and an administration building.
Design data pertaining to the major structures were supplied by Gibbs and Hill, Inc. and are tabulated on the following page.
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APPROXIMATE BASE DESIGN PRESSURE STRUCTURE PLAN DIMENSIONS ELEVATION* AT BASE (feet) (feet) (psf)
Containment Bldg. 120 (diameter) +983 6,000 Auxiliary Bldg. 220 x 240 +981 3,500 (irregularly shaped)
Pumping Station 70 x 100 +963 4,000 Turbine-Generator 112 x 224 +987 2,000 Bldg.
Turbine-Generator 75 x 170 +982 5,000 Mat PURPOSES The purposes of this investigation were to:
- 1) explore the subsurface conditions at the specific locations of the proposed major structures;
- 2) evaluate foundation requirements based upon definitive operating floor levels and anticipated design loads; and
- 3) recommend foundation design criteria for each of the principle and appurtenant structures.
SCOPE OF WORK Our investigations related to construction of the proposed power station were divided into two parts: site environmental studies and foundation studies. The results of our site environmental studies are contained in a prior report** dated March 30, 1967. The present report is concerned exclusively with our foundation studies.
- All elevations is this report refer to Mean Sea Level Datum.
- Report, Site Environmental Studies, Fort Calhoun Station - Unit No. 1, Near Fort Calhoun, Nebraska, Omaha Public Power District.
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A total of 95 borings have been drilled at the site. Six borings were performed by others prior to our work on the project, to obtain preliminary subsurface information. Sixteen test borings were drilled during our site environmental studies to obtain specific geologic data and to provide general foundation design criteria for planning and cost estimating purposes.
Subsurface conditions in the proposed plant area were explored in detail during this investigation by drilling 73 borings under the technical direction and supervision of Dames & Moore soils engineers and engineering geologists in order to provide specific foundation design criteria.
Initially, the present study consisted of 12 test borings in the proposed plant area. During the course of the field investigation, however, a cavity was encountered in the underlying limestone. The presence of cavitation was particularly significant in view of the fact that a foundation system utilizing piles bearing on rock was desired by the structural engineers.
Eight additional borings were drilled in an attempt to define the nature and extent of the void.
Certain of these borings also encountered limestone solution cavities. Since all previous explorations at the site revealed sound limestone, it was believed that the observed cavitation was confined to a limited area. It was, therefore, decided to relocate the plant 90 feet upstream, in an attempt to avoid the zone of observed cavitation.
The rock in the revised plant area was then explored. A program of 118 borings, located on a random grid pattern within the revised plant area, was formulated. During the course of this investigation, it was discovered that the general plant area is underlain by limestone containing solution cavities. As the investigation continued, it became apparent that a conventional pile foundation was not feasible for this project because of the uncertainty of rock conditions at any given location. It was then decided that open-end pipe piles would be utilized for foundation support and a test boring would be drilled through and beneath each pile to evaluate rock conditions under all piles installed to support Class I structures. In view of this decision, the drilling program was terminated before completion of the proposed additional borings.
Of the 73 borings drilled during this phase of our investigation, 12 were drilled in the original plant area to evaluate both soil and rock conditions; seven were drilled in the relocated plant area to evaluate soil and rock conditions; and 54 were drilled to investigate bedrock characteristics only. The locations of the borings are shown on the Plot Plan, Plate 3.
Undisturbed soil samples, suitable for laboratory testing, were extracted from selected borings utilizing the Dames & Moore soil sampler illustrated in the Appendix. Standard split-spoon samples were also obtained in selected borings for correlation and identification purposes. Rock cores were recovered from 71 borings. All soil samples and selected rock cores were forwarded to our New York office for further examination and laboratory testing.
The remaining rock cores are stored near the site at a facility provided by the Omaha Public Power District.
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The manner in which the field investigation was conducted is more fully described in the Appendix to this report. The results of the field explorations and laboratory tests, which provide the basis for our engineering analyses and recommendations, are also presented in the Appendix.
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SITE CONDITIONS SURFACE FEATURES The site is located adjacent to the Missouri River, just upstream of the DeSoto Bend Cutoff, in Washington County, Nebraska. Surface elevations at the site vary from below +1000 feet to more than +1100 feet, at the southern limit of the property. The proposed plant will be located in the northeastern portion of the site where surface elevations vary from about
+1004 to +997 feet.
The most prominent surface feature on the site is a northerly flowing drainage ditch which empties into a river swale located about 400 feet south of the Missouri River. The swale is more than 100 feet wide and contains about three feet of water during normal river conditions. The swale is separated from the river by a strip of accretion land which is subjected to periodic inundation.
Some minor clearing and grading was performed during the field investigation to provide access for drilling equipment. Dikes have been constructed across the swale and the drainage ditch, and fill was placed in certain of the lower marshy site areas.
Much of the site is being farmed. However, the accretion land area is dormant and lightly wooded. Surficial site geology is shown on the Site Plan, Plate 2. The prominent surface features in the plant area are shown on the Plot Plan, Plate 3.
With the exception of a recently constructed meteorological monitoring station, the site was free of structures during this investigation.
SUBSURFACE CONDITIONS Soil: The borings drilled during this and prior investigations indicate that the site is underlain by granular soils of fluvial and glacial outwash origin overlying limestone bedrock.
The surficial soils consist of loose fine sands with varying amounts of silt to a depth of about ten feet. Soft dark brown clayey silt is encountered at the ground surface near the existing swale. These near surface materials are underlain by loose to moderately compact fine sand to a depth of about 30 to 35 feet below existing grade. A layer of compact fine sand, approximately five to ten feet thick, was encountered in most borings at this level. Below this dense layer, the sand is somewhat less compact and ranges from poorly-graded to well-graded, and contains thin seams of silty clay and some gravel.
More detailed descriptions of the soil conditions encountered in the borings are presented on the Logs of Borings contained in the Appendix to this report.
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Bedrock: Bedrock was encountered at depths ranging from 58 to 67 feet below the ground surface in the plant area. The elevation of the bedrock surface varies from approximately
+931 feet to +935 feet.
The bedrock correlates with the Winterset limestone member of the Dennis Formation. The upper four to eight feet of the formation is a massive, gray, thick-bedded medium to fine-grained oolitic limestone. Below the oolitic is a light gray, thin to moderately thick bedded, very fine grained (aphanitic) limestone. The aphanitic limestone ranges in thickness from 19 to 21 feet. Thin shale seams, one-sixteenth to two inches in thickness, often were observed in the aphanitic limestone. Two such seams consistently appeared: one at the interface of the two limestones, and another several inches below the interface. Below the Dennis Formation are alternating strata of shale and limestone to the maximum depths explored (40 feet beneath rock surface).
Cavities were observed in the aphanitic limestone during this investigation. The cavities were encountered at the base of the oolitic limestone and extended from a few inches to approximately 14 feet into the aphanitic limestone.
The cavities are believed to develop by the enlargement of major vertical joints by solution.
The Dennis Formation is traversed by two major joint systems which are essentially vertical and continuous from the top to bottom. Strong solution erosion and/or redeposition of calcite along the joints were observed in many of the recovered cores. Examination of two quarries near the site also revealed calcite deposition in joints. Although information obtained at the quarries should not be considered as specifically representative of subsurface conditions at the site, we believe it is indicative of the general rock characteristics beneath the proposed facilities.
We believe the cavities may have formed by the enlargement of major vertical joints because:
- 1) the rock cores indicate presence of vertical joints below the cavities showing joint characteristics (such as soil erosion, secondary calcite deposition, etc.) similar to those observed in the quarries;
- 2) the trend of the cavities is similar to the directions of joints observed in the quarries;
- 3) the apparent lineal shape of the cavities as determined by a number of closely spaced borings drilled to investigate the extent of encountered cavities is characteristic of an enlarged joint system; and
- 4) evidence of stronger weathering on one side of a rock core than the other indicates the proximity of the specimen of a cavity of vertical nature.
The cavities are believed to have the cross-sectional shape shown on Plate 4, Hypothetical Cross Section of Typical Cavity.
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Cavities tend to develop in the aphanitic limestone since the oolitic is more pervious than the aphanitic and ground water concentrates on the surface of the aphanitic. The flow concentration is channeled downward and laterally along vertical joints in the less pervious aphanitic. The enlargement probably initiates at the interface of the two limestones, where the flow is first concentrated, and progresses along the joints within the aphanitic limestone.
The presence of shale seams near the interface probably facilitates erosion and may account for the top of the cavities being wider than the lower portion (see Plate 4). Erosion progresses along the joints and results in long, linear shaped cavities. A cavity may expand where a softer or more easily eroded material is encountered within the limestone.
With time, weathering and spalling of the oolitic cap rock also takes place, causing enlargement of the cavity into the oolitic limestone, and resulting in a variation in the thickness of cap rock. Simultaneously, the downward flow of water also enlarges the vertical joints within the oolitic limestone. Two borings drilled during this investigation encountered vertical joints in the oolitic large enough to permit the drill bit to penetrate to the cavity floor.
In both instances, we believe the fissure is narrow, probably on the order of from two to six inches wide.
It is significant to note that granitic gravel and silica sand were found in cavities. The presence of materials alien to the Dennis Formation indicates that the cavities are or were connected to the overburden soils. The maximum gravel size encountered in a cavity was about two inches. Most cavities were partially filled with soft decomposed limestone (almost a fluid). However, in some cavities the decomposed limestone was covered by granular overburden soils to the rock surface.
The horizontal extent of the cavities was not fully evaluated during this investigation, nor was continuity between the cavities established. Two fluoroscene dye tests were performed in an attempt to establish cavity interconnection, but the results were inconclusive.
No cavities or weathered rock were encountered below the Dennis Formation. It is likely that the Galesburg Formation shale provides an impermeable barrier which protects the lower limestone of the Swope Formation from erosion. More detailed descriptions of bedrock conditions are shown on the Log of Borings presented in the Appendix.
Groundwater: The elevation of the ground water table during this investigation was approximately +993 feet, essentially the same as the water level in the Missouri River. The water table fluctuates to some extent in response to variations in the river level. At the time of this investigation, the static water table sloped gently toward the Missouri River. Data concerning the depth to the ground water table are presented on the boring logs.
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DISCUSSION AND RECOMMENDATIONS FOUNDATION REQUIREMENTS The heavy structures of the station which are sensitive to settlement, may be supported either by spread or mat foundations established in the lower compact sands (or on bedrock) or piles installed to sound bedrock.
Several alternate foundation schemes were investigated. On the basis of data originally presented during the progress of this work and in our previous report, the designers, Gibbs & Hill, Inc., selected a pile foundation. Therefore, subsequent discussion in this report is based only upon a pile foundation for major structures.
As a result of the high pile design loads it is necessary to reach the limestone bedrock at the site to achieve the required pile capacities. Since the limestone contains areas of solution cavities, an extensive program of rock exploration and subsequent remedial pile installation operations are anticipated. Also, the reduction of inter-granular soil pressures from a conventional mat foundation to those below a pile cap result in a situation where the underlying soils are more susceptible to liquefaction.
Under the "maximum credible earthquake condition" postulated by the Atomic Energy Commission seismology consultants, our analyses indicate that the upper sands may be subject to liquefaction.
Accordingly, it will be necessary to achieve an adequate relative density in the upper sands below the critical structures. We believe some compaction will occur as a result of pile installation. The density of the upper sands should be measured after completion of pile driving operation by means of a thorough exploration program. Should insufficient compaction be noted, remedial treatments, such as vibroflotation or a sand pile installation, will be necessary.
Rock Exploration: As previously indicated, there is a possibility that a cavity may exist at any given spot within the construction area. To provide complete assurance that a cavity does not exist below a pile, it will be necessary to drill a boring at the location of each pile installed beneath the Class I structures.
Rock exploration at the location of an individual pile may be accomplished in one of the following ways:
- 1) a boring may be drilled at the predetermined pile location and the pile driven at that location;
- 2) open-end piles may be driven, and an exploratory boring drilled through and beneath the pile.
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Of the two alternatives, the open end pipe pile scheme would provide the greater degree of certainty regarding rock conditions beneath the pile since the boring will be drilled through an in-place pile. However, good inspection and control of drilling and pile driving operations may afford a comparable degree of confidence and eliminate potential installation problems associated with driving and cleaning open-end, concrete filled pipe piles. The possible installation problems include efficient removal of materials within the pile, loss of concrete through rock seams beneath the pile and inflow of peripheral soils into the pile from beneath the pile tip during jetting.
The exploratory borings should extend a minimum of 15 feet beneath the pile tip. It is our opinion that the boring may be either drilled or cored. It should not be necessary to obtain rock cores from each boring since a good definition of bedrock conditions will be afforded by the number of explorations made. In addition, the apparent nature of the cavities (sharply sloped walls) may negate any benefits which could be derived from the examination of every core since rock conditions may vary considerably over small horizontal distances. However, it is recommended that in any individual boring, rock cores be obtained whenever erratic or unexpected drill behavior is observed. It is recommended that the drilling program be supervised by an experienced geologist or engineer familiar with bedrock conditions and drilling techniques.
Should the boring drilled at a given pile location reveal sound rock to a depth of 15 feet beneath the pile tip, the pile may be considered satisfactory and thus be founded on the initially encountered bedrock surface. In the event that a cavity is encountered in the boring, it will be necessary to advance the pile to the sound rock which will be encountered at the base of the cavity. A more detailed discussion of possible installation procedures for this operation is contained in the Foundation Installation section of this report.
Pile Design Criteria - Class I Structure: Initially, open-end pipe, closed-end pipe, and H-piles were given primary consideration for the Class I structures. For scheduling and installation reasons, open-end pipe piles were chosen by Gibbs & Hill.
Piles founded on sound rock, either at the initially encountered bedrock surface or at the base of a penetrated cavity, may be designed for compressional capacities up to the structural limits of the piles. We understand that maximum compressional design loads of 325 tons are anticipated in areas of Class I construction. Either the sound oolitic or sound aphanitic limestones are capable of safely supporting these loads with negligible elastic deformation.
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We have analyzed a number of pile types and sizes to determine their uplift and lateral capacities. Estimated uplift capacities are presented in Table I. Estimated lateral capacities are presented in Table 2. An appropriate factor of safety should be applied to these values.
We have performed analyses to determine pile group capacities based on a pile layout predicated upon a total of about 1,000 piles. These analyses indicate that individual pile failure will occur before group failure. Consequently, the capacities indicated in Tables 1 and 2 should be utilized for design purposes with an appropriate safety factor.
Table 1 - Estimated Uplift Pile Capacities PILE TYPE CUTOFF EL. = +983 FEET CUTOFF EL. = +967 FEET 16" OD Pipe Pile 50 tons 34 tons 18" OD Pipe Pile 56 tons 38 tons 20" OD Pipe Pile 62 tons 42 tons 22" OD Pipe Pile 68 tons 46 tons Table 2 - Estimated Lateral Pile Capacities NORMAL DESIGN MAXIMUM DESIGN ESTIMATED ULTIMATE ESTIMATED LATERAL LATERAL LATERAL LATERAL PILE TYPE CAPACITY DISPLACEMENT CAPACITY DISPLACEMENT (tons) (inches) (tons) (inches) 16" OD, concrete filled pipe pile (2" wall thickness) 18 1/4 50 3/4 to 13 18" OD unconcreted pipe pile (2" wall thickness) 9 1/4 52 3/4 to 13 18" OD unconcreted pipe pile (3/4" wall thickness) 22 1/4 60 3/4 to 13 20" OD unconcreted pipe pile (2" wall thickness 23 1/4 70 3/4 to 13 20" OD unconcreted pipe pile (.782" wall thickness) 27 1/4 100 1 to 12 22" unconcreted OD pipe pile (2" wall thickness) 35 1/4 100 3/4 to 13 DAMES & MOORE DSAR-Appendix C Information Use Page 16 of 87 Foundation Studies Rev.0 ARCHIVED TEXT*
The indicated uplift and lateral capacities were computed utilizing soil parameters consistent with a minimum in-situ relative density of 70 percent. These criteria were chosen since, as a result of discussions with the Atomic Energy Commission, it was decided that improvement of the upper sands would be necessary to avoid the possibility of liquefaction under the maximum credible earthquake. As subsequently discussed, it has been resolved that the upper sands will be compacted to an average relative density of 85 percent. Since the minimum relative density will be 70 percent or so, the indicated pile capacities are probably somewhat conservative.
Pile Design Criteria Class II Structures: The proposed turbine, administration, and other support buildings of the proposed plant will be founded between Elevation +983 and +987 feet. We understand that required pile capacities are on the order of 90 tons per pile.
We understand that both concrete filled pipe and Raymond step-taper piles are being considered for support of Class II structures. Either pile type should be driven to essential refusal on the underlying limestone bedrock encountered at depths of about 60 to 63 feet beneath existing grade in these areas. Detailed rock exploration under each pile is not warranted beneath Class II structures because of the relatively light loads upon each pile.
Accordingly, should concrete filled pipe piles be selected, it is recommended they be driven closed-ended in order to avoid potential installation difficulties. A more detailed discussion of foundation installation is presented in a subsequent section of this report. We believe that piles installed in accord with these recommendations will experience negligible elastic deflection under the design loads.
The use of displacement piles (such as a step taper or closed-end pipe) will significantly improve the subsurface conditions by compacting the in-situ soils. The resulting compaction will reduce the tendency toward liquefaction and increase lateral and uplift pile capacities beneath areas of Class II construction.
Pile Design Criteria: We have analyzed the planned pile types for uplift and lateral pile capacities. The results of these analyses are presented in Tables 3 and 4. An appropriate factor of safety should be included in design.
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Table 3 - Uplift Pile Capacities PILE TYPE CUTOFF ELEVATION - +983 Feet (Tons) 12-3/4" OD Pipe 35 Raymond Step Taper 65 (with appropriate reinforcing)
Table 4 - Lateral Pile Capacities NORMAL DESIGN MAXIMUM DESIGN ESTIMATED ESTIMATED LATERAL LATERAL LATERAL LATERAL PILE TYPE CAPACITY DISPLACEMENT CAPACITY DISPLACEMENT (tons) (inches) (tons) (inches) 12-3/4" OD concrete-filled pipe pile (3" wall thickness) 11 1/4 30 2 to 3/4 Raymond Step Taper *
(with appropriate reinforcing)
- The lateral capacity of a Raymond step taper pile is greatly dependent on the amount of reinforcing steel placed within the pile. The final pile selection may be analyzed in accordance with "Lateral Resistance of Piles in Cohesionless Soils", by Bengt B. Broms, Journal of the Soil Mechanics and Foundation Division, ASCE, May 1964, Part 1. We would be pleased to provide estimates of lateral pile capacities after a reinforcing system is selected.
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Liquefaction: When subjected to a shaking motion, loose to medium dense sands tend to compact with a consequent reduction in the volume of the pore space. If the sand has saturated pore spaces, this tendency to compact will increase the pressure in the pore water.
If this excess pore water pressure cannot be dissipated by drainage and increases until it is equal to the overburden pressure, the confining pressure, (effective stress) on the sand will be reduced to zero. When this occurs, the sand undergoes a complete loss of strength and liquefaction develops.
The liquefaction phenomenon is affected by many quantities and properties, all of which must be evaluated before an estimate of the liquefaction potential of any specific soil deposit can be made. Of these the principle variables are the density characteristics of the sand, the magnitude of the confining pressure, the magnitude, frequency and duration of the shaking forces, the boundary conditions affecting drainage and the particle size distribution. The general effects of each of these variables will be discussed before considering their application to the conditions at the project site.
Sand may exist over a wide range of densities. In soil mechanics this density condition is usually defined either by means of the void ratio (the ratio of the volume of the void space to the volume of the solids) or the porosity (the ratio of the volume of the voids to the total volume).
All variables must be considered together in each case. If all other variables can be held constant, however, a sand with a high void ratio will liquefy before an identical sand with a low void ratio.
An increase in confining pressure will make liquefaction of the sample more difficult. Physical properties of the sand, such as the elastic moduli and damping factors, are also dependent on confining pressures and can change the response of the system which may be subject to liquefaction.
Increasing either or both the magnitude of the dynamic stress and the number of applications of the cyclic loading will increase the probability that any given deposit will liquefy. The loading condition must be such that the shear stresses be reduced to zero or be reversed in direction during the loading to produce liquefaction. This condition is satisfied under normal earthquake conditions in uniform level deposits when the cyclic shearing stresses result principally from vertically progressing shear waves producing cyclic strains which reverse direction many times. Under structures, existing shear stresses may prevent reversal during shaking.
If the increase in pore pressures can be dissipated as the shaking motion occurs, the pore pressure may never rise to equal the confining pressure. This would prevent liquefaction from occurring. This situation might exist in a dike or similar structures with short drainage paths. For large areas, the effective drainage during cycling is expected to be small and any beneficial effects which might occur due to drainage should be ignored in a conservative approach.
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Where sands have non-uniform grain sizes or appreciable amounts of silt or clay particles, the tendency toward liquefaction under cyclic loading may be reduced. The effect of varying particle size and varying amounts of fines can be estimated in direct laboratory testing under dynamic conditions.
In our liquefaction studies for the subject project, a maximum credible acceleration of 17 percent of gravity was selected for use based on recommendations of the U. S. Coast and Geodetic Survey. We believe an earthquake in this region would have a duration of strong shaking of not more than about seven seconds. During this period, several cycles of large shearing stresses, together with many smaller cycles, may be expected. The strongest soil vibrations will be primarily in the first mode with a fundamental period of slightly more than one-half second. This would give a total number of 13-14 oscillations during the seven second duration. Because of the random nature of the forcing vibration of the bedrock and the effect of soil damping, only half of these oscillations will have appreciable magnitudes.
Thus, it might be expected that there may be six to seven cycles of shearing stress of a sufficient magnitude to cause concern about liquefaction during the hypothetical earthquake.
The results of our dynamic laboratory tests which simulated the vibratory motion that may be expected during the design earthquake, are presented in the Appendix to this report.
The presently proposed reactor foundation level ranges from about Elevation +983 feet to
+967 feet. The founding level of the turbine generator building ranges from about Elevation
+987 to +983 feet. The base of the auxiliary building is at approximately Elevation +986 feet.
The results of our testing and analyses indicate that the possibility of liquefaction at the site under the maximum credible earthquake is marginal to about Elevation +965 or above, under existing subsurface conditions. Beneath this elevation the soils either become well graded or compact, resulting in a negligible tendency toward liquefaction.
Our analyses, performed under very conservative assumptions, indicate that the existing soils beneath the main construction area would not liquefy if at a relative density of 67 percent.
Outside this area, the in-situ sands would not liquefy if at a relative density of 58 percent, due to the increase in effective stress induced by fill placement. Existing relative densities within the upper strata vary from approximately 40 to 87 percent, averaging about 62 percent.
Thus, under existing soil conditions and the loading which would be expected from the maximum credible earthquake, a conservative analysis indicates that liquefaction of the upper sands is likely.
The analysis used to determine the liquefaction criteria indicated above reflects several very conservative assumptions. The assumptions were:
- 1) The shearing stresses induced by the maximum credible earthquake are a maximum at all levels in the soil stratum at the same time.
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- 2) There is no transfer of load from the piles and pile cap to the upper soils.
- 3) There is no distribution of load from the adjacent overburden soils to the area below the main structures.
An analysis incorporating more realistic conditions than the above assumption indicates that the upper soils would not liquefy if above an average relative density of approximately 35 percent.
In addition, we believe the upper soils at the completion of pile installation may be more compact than at present for the reasons described in subsequent paragraphs.
Soil displacement will occur from the driving of each pile, thus increasing the relative density of the peripheral soils. The use of closed-end pipe would be more beneficial than open-end piles.
A core at the center of the reactor location will be founded at that about Elevation +967 feet.
Installation of the core will require a cofferdam and dewatering system inducing a drawdown in the existing water level of approximately 36 feet. The results of our pumping test as presented in the site environmental studies for this project indicate that the required depth of dewatering necessary to keep the water beneath the base of the excavation will be approximately Elevation +957 feet. The pumping operations will have two effects which will increase the relative density of sands beneath the proposed plant:
- 1) The required depth of dewatering will produce a drawdown in the static water level of approximately 20 feet for a distance of 200 feet from the core center.
This is equivalent to applying a surcharge over the entire construction area. This operation will cause some consolidation of the fine sands and a consequent increase in relative density.
- 2) The downward movement of water through the sand strata will force the particles to readjust themselves to a more compact condition. The cessation of dewatering operations and the subsequent water rise to the normal static water level will not produce a considerable decrease in the improved relative density since the soils will be loaded by the planned construction.
Exterior plant subgrade will be Elevation +1004 feet. Thus, fills of from 4 to 15 feet will be required in the general area. The increase in overburden pressure will increase the effective stress, even under the building areas, and result in a decreased tendency toward liquefaction.
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It is recommended that, at the completion of pile driving and dewatering operations, a number of exploratory borings be drilled within the proposed construction area to evaluate the actual in-situ relative density. The borings should penetrate to at least below Elevation +965 feet.
Most borings should be drilled at least four inches in diameter. Standard penetration tests should be performed at close intervals for the length of the boring. In addition, undisturbed samples suitable for laboratory density testing should be obtained in several borings. The results of these borings will enable us to determine actual field conditions at the time of construction.
In the event that present relative densities have not substantially changed, compactive operations will be necessary to densify the fine sand stratum. Possible methods of increasing of the relative density include:
- 1) Installation of sand piles; and
- 2) vibroflotation.
In accord with a conservative approach, as recommended by the Atomic Energy Commission, the in-situ soils will be improved to an average relative density of 85 percent. It was agreed that this relative density would be determined in accord with "Research on Determining the Density of Sand by Spoon Penetration Testing" by Gibbs and Holtz.
It is strongly recommended that all operations relative to pile driving (and rock exploration),
soil exploration after pile driving, and execution of remedial compactive efforts (if required) be continuously supervised and controlled by qualified soils engineers thoroughly familiar with subsurface conditions and operational objectives.
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CORROSIVITY The results of chemical analyses performed on the soils and ground water at the site for our Site Environmental Studies and this investigation indicate the subsurface materials are only slightly basic and should not adversely affect buried steel members. Similarly, soluble sulphate determinations indicate a negligible potential toward concrete corrosion.
FOUNDATION INSTALLATION In order to avoid potentially serious installation problems, it is recommended that unconcreted, open-end pipe or closed-end concrete filled pipe piles be installed beneath Class I Structures and closed pipe, or Raymond Shell (or similar) piles be installed beneath Class II Structures. All piles should be installed with a heavy hammer delivering at least 24,000 foot-pounds of energy. Piles driven to support the planned design loads of 325 tons should be driven with a hammer delivering at least 40,000 foot-pounds of energy. The piles should be driven to essential refusal (ten blows to the final 1/4 inch) on the limestone bedrock encountered at Elevation +931 to +935 feet.
Class I Structures: A number of difficulties may be encountered in installing open-end concrete filled pipe piles. It may prove difficult to adequately clean a driven pile for two reasons:
- 1) The fine sand which overlies the bedrock in the area may flow into the pile from beneath the tip as the pile is being jetted, despite careful seating attempts. (This was a recurring and difficult problem in drilling operations during performance of the boring program.) Should the sand inflow become severe, the lateral and uplift capacities of the piles could be affected due to significant disturbance of the peripheral soils.
- 2) Fine to coarse gravel was encountered in numerous borings in the lower sand strata. It may prove difficult to remove the larger gravel from within the pile in jetting operations, thus resulting in incomplete and unsatisfactory concreting of the pile.
To eliminate these potential problems, it is recommended that unconcreted open-end pipe or closed-end pipe piles be driven beneath Class I structures.
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Should closed-end pipe piles be utilized for Class I structures, the following installation scheme should be considered:
A boring should be drilled at the predetermined pile location. If the boring encounters sound bedrock at the pile location, the pile may be driven to bedrock and concreted.
Should the boring reveal the presence of a cavity, a large diameter casing may be installed to bedrock surface. The cap rock above the cavity may then be reamed using standard drilling equipment, and the pile seated upon a flat base established by the drill at the cavity floor. The casing may then be pulled, and the peripheral soils disturbed by the operation restored to the required density by vibroflotation or by installing sand piles. Penetration tests to establish the density of the peripheral soils can be used. In the event that the soils are not of adequate density, vibroflotation or additional sand piles should be employed.
Should open-end, unconcreted pipe piles be selected, the boring may be drilled either before or after pile driving. If the boring is drilled before pile driving and a cavity is encountered, the previously outlined method may be used to advance the pile to suitable bearing. If the boring is drilled through a previously driven pile and a cavity is encountered, an under-reaming technique will be necessary. (The boring can be a probing operation if properly inspected.
The ability to core rock upon meeting any condition not easily interpreted by drilling operations is imperative. The decision as to where sound rock is found, or where to core should be performed by an experienced engineer.) The following scheme has been suggested by Gibbs & Hill, Inc.:
The cap rock may be reamed out using an expandable drill bit. We understand this tool may be lowered within the pipe pile subsequent to cleaning operations and be made to expand by the use of hydraulic pressure, subsequently drilling a hole in the bedrock larger than the outside diameter of the pipe pile. The suitability of this tool for a production pile installation should be proven in a pile load test program, since it is usually used for oil well drilling operations.
Should a boring reveal no cavity, the pile may be founded on the initially encountered bedrock.
The following criteria should apply to all piles installed in areas in Class I Construction:
- 1) The piles must be founded on a flat base (either at the bedrock surface or the cavity floor).
- 2) The piles should be essentially plumb and not allowed to follow the slope of the cavity wall.
- 3) Close control of the welding operations should be maintained because of the high structural stresses which will be induced by the anticipated loads.
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Due to the complexity of subsurface conditions and unusual foundation types, it is strongly recommended that all operations related to foundation installation be under the inspection of a qualified soil mechanics engineer. This inspection should cover pile driving operations, rock explorations, underreaming operations, soil explorations after pile driving, and any required remedial compaction operations.
Class II Structure: While the piles below Class II structures will not carry as high a structural load as those under Class I facilities, they should also be installed with care. All piles should be fully seated on the limestone. A refusal criteria of about ten blows per 1/4 inch should be suitable for piles driven with a hammer delivering greater than 24,000 foot-pounds of energy.
The pile installation operation for Class II structures should essentially follow conventional procedures used with an important structure PILE LOAD TEST PROGRAM A comprehensive pile load test program is in progress. The program will evaluate actual pile capacities, appropriate soil and rock parameters related to pile behavior, and the feasibility of planned foundation installation techniques. The results of this program will be submitted in a subsequent report.
The following Plates and Appendix are attached and complete this report:
Plate 1 - Map of Area Plate 2 - Site Plan Showing Surficial Geology Plate 3 - Plot Plan Relative to Surface Features Plate 4 - Hypothetical Cross Section of Typical Cavity Appendix - Field Explorations and Laboratory Tests Respectively submitted, DAMES & MOORE Joseph A. Fischer Thomas E. Tully JAF/TET/deb DAMES & MOORE DSAR-Appendix C Information Use Page 25 of 87 Foundation Studies Rev.0 ARCHIVED TEXT*
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APPENDIX FIELD EXPLORATIONS AND LABORATORY TESTS FIELD EXPLORATIONS Subsurface conditions at the proposed site were explored by drilling a total of 73 borings to depths ranging from 60-1/2 to 109-1/2 feet. The locations of the borings are shown relative to the proposed construction on the Plot Plan, Plate 3.
Nineteen borings were drilled to investigate soil and rock conditions beneath the originally proposed and subsequently revised plant locations. Fifty-four borings were drilled to investigate bedrock conditions in the general plant area. A number of borings were concentrated in areas of previously encountered cavitation in an attempt to define the nature and extent of typical cavities.
All field work was performed under the technical direction of Dames & Moore engineers and engineering geologists. The borings were drilled approximately four inches in diameter, utilizing truck mounted rotary drilling equipment. Casing was used to prevent the walls of the borings from caving.
Continuous observations of the soil and rock encountered in the borings were recorded in the field by our engineers and geologists. Undisturbed soil samples, suitable for laboratory testing, were extracted from 13 of the borings utilizing the Dames & Moore soil sampler illustrated on Page 29 of this Appendix. Standard penetration tests utilizing the standard split spoon sampler were performed on the overburden soils in 11 borings. Both samplers were also used to sample materials found within certain cavities. Rock cores were extracted from most borings utilizing either an NX or BX size core barrel. NX size core is approximately two and one-eighth inches in diameter. BX size core is approximately one and three quarters inches in diameter.
The soil samples extracted from the borings were transported to our New York office where they were further examined and subjected to appropriate laboratory tests. The rock cores obtained from the borings were either sent to our New York office or stored at the site.
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Detailed descriptions of the soil and rock encountered are presented on Plates A-1A through A-1A0, Log of Borings. The soils were classified in accordance with the Unified Soil Classification System described on Plate A-2. The number of blows required to drive the Dames & Moore sampler, or the standard split spoon sampler, a distance of one foot into the undisturbed material is presented in the column at the left of each log of boring. The percent of rock recovery obtained during coring operations is also presented in this column. Data relative to the energy used to advance either sampler is presented at the bottom of the boring logs.
The elevations shown at the top of the boring logs refer to U.S.C. & G.S. Mean Sea Level Datum. The depth to which casing was used, and the date on which the boring was completed appear beneath the log of each boring.
Resistivity tests were conducted at the site in order to determine the resistance to electrical current of the in-situ soils. The tests were performed at the location of borings 28 and 31.
The results of the tests are presented in Table 1.
Table 1 - Resistivity Test Results Boring 28 Surface Elevation = +996.2' To Depth Apparent Resistivity (feet) (Ohm-cm) 3 870 6 1245 9 1470 12 1610 15 1755 18 1920 21 2070 24 2230 27 2370 30 2470 33 2650 36 2780 39 2910 42 3020 45 3140 48 3260 51 3400 54 3540 57 3660 DAMES & MOORE DSAR-Appendix C Information Use Page 32 of 87 Foundation Studies Rev.0 ARCHIVED TEXT*
Boring 31 Surface Elevation = +1001.9' To Depth Apparent Resistivity (feet) (Ohm-cm) 5 1590 10 1835 15 1790 20 1870 25 2060 30 2290 35 2475 40 2685 45 2915 50 3140 55 3320 A survey of a quarry located near the site was made for the purpose of determining the joint directions and spacing of the limestone bedrock. The results of the survey are not presented in this report. Copies of the traverse were retained by both Omaha Public Power District and Dames & Moore, and are available in our files if needed.
LABORATORY TESTS Strength-Tests: Selected representative soil samples recovered from the borings were tested to determine their strength characteristics. These tests were performed to evaluate the cohesion and angle of internal friction of the in-situ soils.
Direct shear tests and friction tests were performed on selected soil samples as described on Page 32. Static triaxial compression tests were performed as described on Page 33.
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Method of Performing Direct Shear and Friction Tests Direct Shear Tests are performed to determine the shearing strengths of soils. Friction tests are performed to determine the frictional resistances between soils and various other materials such as wood, steel or concrete. The tests are performed in the laboratory to simulate anticipated field conditions.
Each sample is tested within three brass rings, two and one-half inches in diameter and one inch in length. Undisturbed samples of in-place soils are tested in rings taken from the sampling device in which the samples were obtained. Loose samples of soils to be used in constructing earth fills are compacted in rings to predetermined conditions and tested.
Direct Shear Tests A three-inch length of the sample is tested in direct double shear. A constant pressure, appropriate to the conditions of the problem for which the test is being performed, is applied normal to the ends of the sample through porous stones. A shearing failure of the sample is caused by moving the center ring in a direction perpendicular to the axis of the sample, transverse movement of the outer rings is prevented.
The shearing failure may be accomplished by applying to the center ring either a constant rate of load, a constant rate of deflection, or increments of load or deflection. In each case, the shearing load and the deflections in both the axial and transverse directions are recorded and plotted. The shearing strength of the soil is determined from the resulting load-deflection curves.
Friction Tests In order to determine the frictional resistance between soil and the surfaces of various materials, the center ring of soil in the direct shear test is replaced by a disk of the material to be tested. The test is then performed in the same manner as the direct shear test by forcing the disk of material from the soil surfaces.
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Methods of Performing Unconfined Compression and Triaxial Compression Tests The shearing strengths of soils are determined from the results of unconfined compression and triaxial compression tests. In triaxial compression tests the test method and the magnitude of the confining pressure are chosen to simulate anticipated field conditions.
Unconfined compression and triaxial compression tests are performed on undisturbed or remolded samples of soil approximately six inches in length and two and one-half inches in diameter. The tests are run either strain-controlled or stress-controlled. In a strain-controlled test the sample is subjected to a constant rate of deflection and the resulting stresses are recorded. In a stress-controlled test the sample is subjected to equal increments of load with each increment being maintained until an equilibrium condition with respect to strain is achieved.
Yield, peak, or ultimate stresses are determined from the stress-strain plot for each sample and the principle stresses are evaluated. The principle stresses are plotted on a Mohr's circle diagram to determine the shearing strength of the soil type being tested.
Unconfined compression tests can be performed only on samples with sufficient cohesion so that the soil will stand as an unsupported cylinder. These tests may be run at natural moisture content or on artificially saturated soils.
In a triaxial compression test the sample is encased in a rubber membrane, placed in a test chamber, and subjected to a confining pressure throughout the duration of the test.
Normally, this confining pressure is maintained at a constant level, although for special tests it may be varied in relation to the measured stresses. Triaxial compression tests may be run on soils at field moisture content or an artificially saturated samples. The tests are performed in one of the following ways:
- Unconsolidated-Undrained: The confining pressure is imposed on the sample at the start of the test. No drainage is permitted and the stresses which are measured represent the sum of the intergranular stresses and pore water pressures.
- Consolidated-Undrained: The sample is allowed to consolidate fully under the applied confining pressure prior to the start of the test. The volume change is determined by measuring the water and/or air expelled during consolidation. No drainage is permitted during the test and the stresses which are measured are the same as for the unconsolidated-undrained test.
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- Drained: The intergranular stresses in a sample may be measured by performing a drained, or slow, test. In this test the sample is fully saturated and consolidated prior to the start of the test. During the test, drainage is permitted and the test is performed at a slow enough rate to prevent the buildup of pore water pressures. The resulting stresses which are measured represent only the intergranular stresses. These tests are usually performed on samples of generally non-cohesive soils, although the test procedure is applicable to cohesive soils if a sufficiently slow test rate is used.
All alternate means of obtaining the data resulting from the drained test is to perform an undrained test in which special equipment is used to measure the pore water pressures. The differences between the total stresses and the pore water pressures measured are the intergranular stresses.
A load deflection curve was plotted for each strength test and the strength of the soil was determined from this curve. Determinations of the field moisture content and dry density of the soil were made in conjunction with each strength test. The results of the strength tests and the corresponding moisture content and dry density determinations are presented to the left of the boring logs in a manner described by the Key to Test Data shown on Plate A-2.
Several samples were subjected to dynamic triaxial compression tests in order to evaluate the liquefaction potential of the in-situ soils. Selected soil samples were reconstructed to a range of relative densities representative of the in-situ characteristics of the upper fine sands.
Each sample was then subjected to a predetermined chamber pressure. A series of increasing cyclic loads was then applied until total liquefaction occurred. The ratio of shearing stress to effective stress at liquefaction for each sample was then computed and plotted against relative density. The minimum relative density required to avoid liquefaction under the design earthquake loading was then determined from this graph.
The results of the dynamic triaxial compression tests are presented in Table 2 on pages 35 through 37.
Confined Compression (Consolidation) Tests: Representative samples of the soil underlying the site were subjected to confined compression tests. These tests were performed to evaluate the compressibility characteristics of the natural materials.
The method of performing consolidation tests is described on Page 38. Confined compression tests are similar to consolidation tests except they are performed on predominantly granular materials, rather than cohesive soils, and the loads are applied relatively quickly. The results of these tests, along with the associated moisture content and dry density determinations are presented on Plate A-4, Confined Compression Test Data.
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Particle-Size Analysis: A number of selected soil samples were analyzed in order to determine their grain-size distribution. The results of these tests were used for classification purposes and in liquefaction studies. Grain-size curves illustrating the results of the particle-size analyses are presented on Plates A-3A through A-3C.
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Table 2 - "Dynamic Triaxial Compression Test Results" Boring 22 at 26 feet Relative Density = 40 percent Confining Pressure = 1,500 psf Cyclic Loading Sequence Maximum Deviator Minimum Deviator Stress Stress Number of Cycles (psf) (psf) 1765 1575 8 1650 1430 15 1710 1280 15 1770 1280 6 to total liquefaction Boring 22 at 26 feet Relative Density = 64 percent Confining Pressure = 2,000 psf Cyclic Loading Sequence Maximum Deviator Minimum Deviator Stress Stress Number of Cycles (psf) (psf) 2136 2047 2136 1945 15 2139 1858 7 2230 1740 15 2335 1660 15 2450 1540 15 2590 1390 15 2710 1350 15 2800 1230 15 2940 1110 4 Equivalent to 19 to liquefaction*
- Significant pore pressure increase occurred during the combined 19 cycles prior to liquefaction.
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Boring 27 at 18-1/2 feet Relative Density = 40 percent Confining Pressure = 1,000 psf Cyclic Loading Sequence Maximum Deviator Minimum Deviator Stress Stress Number of Cycles (psf) (psf) 1136 1008 15 1140 858 15 1170 792 15 1235 742 3 to total liquefaction Boring 27 at 18-1/2 feet Relative Density = 90 percent Confining Pressure = 1,000 psf Maximum Deviator Minimum Deviator Stress Stress Number of Cycles (psf) (psf) 1180 1100 15 1180 910 15 1180 830 15 1240 780 15 1290 720 15 1350 670 15 1390 620 15 1440 530 15 1500 470 15 1560 440 15 1630 350 15 1690 350 15 Equivalent to 30 to 710 290 15 liquefaction*
- Significant pore pressure increase occurred during the combined 30 cycles prior to liquefaction.
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Boring 27 at 56 feet Relative Density = 68 percent Confining Pressure = 2,500 psf Maximum Deviator Minimum Deviator Stress Stress Number of Cycles (psf) (psf) 2850 2370 15 2950 2120 15 3070 2000 15 3170 1860 15 3250 1750 15 4350 1640 15 3460 1480 15 3660 1280 15 Equivalent to 29 to 3840 1190 14 Liquefaction*
- Significant pore pressure increase occurred during the combined 29 cycles prior to liquefaction.
Composite sample from Boring 22 at 26 feet and Boring 27 at 18-1/2 feet Relative Density = 40 percent Confining Pressure = 2,000 psf Maximum Deviator Minimum Deviator Stress Stress Number of Cycles (psf) (psf) 2142 1840 15 2196 1765 15 2300 1685 15 2370 1630 15 2485 1515 16 to total liquefaction DAMES & MOORE DSAR-Appendix C Information Use Page 40 of 87 Foundation Studies Rev.0 ARCHIVED TEXT*
Method of Performing Consolidation Tests Consolidation tests are performed to evaluate the volume changes of soils subjected to increased loads, time-consolidation and pressure-consolidation curves may be plotted from the data obtained in the tests. Engineering analysis based on these curves permit estimates to be made of the probable magnitude and rate of settlement of the tested soils under applied loads.
Each sample is tested within brass rings two and one-half inches in diameter and one inch in length. Undisturbed samples of in-place soils are tested in rings taken from the sampling device in which the samples were obtained.
Loose samples of soils to be used in constructing earth fills are compacted in rings to predetermined conditions and tested.
In testing, the sample is rigidly confined laterally by the brass ring, axial loads are transmitted to the ends of the sample by porous disks. The disks allow drainage of the loaded sample. The axial compression or expansion of the sample is measured by a micrometer dial indicator at appropriate time intervals after each load increment is applied. Each load is ordinarily twice the preceding load. The increments are selected to obtain consolidation data representing the field loading conditions for which the test is being performed. Each load increment is allowed to act over an interval of time dependent on the type and extent of the soil in the field.
Relative Density Determination: A number of selected soil samples were tested to determine their relative density. Relative density is a comparison of in-situ density to the maximum and minimum densities of a soil. The results of these tests were used primarily in conjunction with the results of the dynamic triaxial compression tests in evaluating the tendency toward liquefaction of the on-site granular soils. The results of the relative density determinations are presented in Table 3.
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Table 3 - "Results of Relative Density Determinations" Boring Depth Field Dry Density Relative Density (feet) (lbs/ft3) (percent) 21 202 97 61 22 21 99 71 23 352 100 64 23 452 91 47 26 41 102 78 27 26 101 64 29 182 101 68 29 21 101 60 30 21 100 74 30 31 99 69 30 51 114 82 Moisture and Density Determination: In addition to the moisture content and dry density determinations made in conjunction with each strength test, independent moisture and density tests were performed on other undisturbed soil samples for correlation purposes.
The results of all moisture and density determinations are presented on the boring logs.
Chemical Tests: Selected soil samples were subjected to chemical tests. The results obtained were used to determine and compare the chemical composition (and derivative histories) of the materials encountered within and above one of the limestone cavities at the site, as well as to evaluate the possible corrosive effects of the soils on steel and concrete.
The results of these tests are presented in Table 4.
Table 4 - "Results of Chemical Tests" Boring 30 @ 61' Boring 30A @ 69' Boring 30A @ 72' pH 7.5 7.9 8.0
% Sulphates (as SO4) 0.0010 0.0015 0.0205
% Chlorides (as Cl) 0.0014 0.0010 0.0017
% Iron 0.0002 0.00001 0.00006
% Silicates 73 25.0 7.0
% Carbonates 21 70.0 91.0
% Organic Material negligible negligible negligible DAMES & MOORE DSAR-Appendix C Information Use Page 42 of 87 Foundation Studies Rev. 0 ARCHIVED TEXT*
The following plates are attached and complete this Appendix:
Plate A-1A - Log of Boring (Boring 20)
Plate A-1B - Log of Boring (Boring 21)
Plate A-1C - Log of Boring (Boring 22)
Plate A-1D - Log of Boring (Boring 23)
Plate A-1E - Log of Borings (Borings 24 and 25)
Plate A-1F - Log of Boring (Boring 26)
Plate A-1G - Log of Boring (Boring 27)
Plate A-1H - Log of Boring (Boring 28)
Plate A-1I - Log of Boring (Boring 29)
Plate A-1J - Log of Boring (Boring 30)
Plate A-1K - Log of Borings (Borings 30A and 30B)
Plate A-1L - Log of Borings (Borings 30C and 30D)
Plate A-1M - Log of Borings (Borings 30E and 30F)
Plate A-1N - Log of Borings (Borings 30G and 30H)
Plate A-1O - Log of Borings (Borings 30J and 30K)
Plate A-1P - Log of Borings (Borings 30L and 30M)
Plate A-1Q - Log of Borings (Borings 30N and 30P)
Plate A-1R - Log of Borings (Borings 30Q and 31)
Plate A-1S - Log of Borings (Borings 32 and 33)
Plate A-1T - Log of Borings (Borings 38 and 39)
Plate A-1U - Log of Borings (Borings 60 and 62)
Plate A-1V - Log of Borings (Borings 63 and 67)
Plate A-1W - Log of Borings (Borings 70 and 72)
Plate A-1X - Log of Borings (Borings 72A and 72B)
Plate A-1Y - Log of Borings (Borings 72C and 72D)
Plate A-1Z - Log of Borings (Borings 72E and 72F)
Plate A-1AA - Log of Borings (Borings 72G and 72H)
Plate A-1AB - Log of Borings (Borings 76 and 80)
Plate A-1AC - Log of Borings (Borings 85 and 88)
Plate A-1AD - Log of Borings (Borings 91 and 98)
Plate A-1AE - Log of Borings (Borings 100 and 103)
Plate A-1AF - Log of Borings (Borings 103A and 104)
Plate A-1AG - Log of Borings (Borings 104A and 104B)
Plate A-1AH - Log of Borings (Borings 106 and 108)
Plate A-1AI - Log of Borings (Borings 110 and 113)
Plate A-1AJ - Log of Borings (Borings 116 and 118)
Plate A-1AK - Log of Borings (Borings 123 and 126)
Plate A-1AL - Log of Borings (Borings 127 and 128)
Plate A-1AM - Log of Borings (Borings 131 and 133)
Plate A-1AN - Log of Borings (Borings 139 and 141)
Plate A-1AO - Log of Borings (Boring 144 and 150)
Plate A Unified Soil Classification System Plate A-3A - Particle Size Analysis DAMES & MOORE DSAR-Appendix C Information Use Page 43 of 87 Foundation Studies Rev. 0 ARCHIVED TEXT*
Plate A-3B - Particle Size Analysis Plate A-3C - Particle Size Analysis Plate A Confined Compression Test Data DAMES & MOORE DSAR-Appendix C Information Use Page 44 of 87 Foundation Studies Rev. 0 ARCHIVED TEXT*
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