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Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - i Table of Contents 2.0 Site Characteristics 2.1 Geography and Demography 2.1.1 Site Location and Description 2.1.1.1 Specification of Location 2.1.1.2 Site Area 2.1.1.3 Boundaries for Establishing Effluent Release Limits 2.1.2 Exclusion Area Authority and Control 2.1.2.1 Authority 2.1.2.2 Control of Activities Unrelated to Plant Operation 2.1.2.3 Arrangements for Traffic Control 2.1.2.4 Abandonment or Relocation of Roads 2.1.3 Population Distribution 2.1.3.1 Population Within 10 Miles 2.1.3.2 Population Between 10 and 50 Miles 2.1.3.3 Transient Population 2.1.3.4 Low Population Zone 2.1.3.5 Population Center 2.1.3.6 Population Density 2.1.4 References 2.2 Nearby Industrial, Transportation, and Military Facilities 2.2.1 Locations and Routes 2.2.2 Descriptions 2.2.2.1 Description of Facilities 2.2.2.2 Description of Products and Materials 2.2.2.3 Pipelines 2.2.2.3.1 Natural Gas Pipelines 2.2.2.3.2 Propane Facilities 2.2.2.4 Waterways 2.2.2.5 Aircraft Activities 2.2.2.5.1 Rock Hill Airport 2.2.2.5.2 Charlotte/Douglas International Airport 2.2.2.5.3 Gastonia Municipal Airport 2.2.2.5.4 Flight Paths and Pilot Training Areas 2.2.2.6 Projections of Industrial Growth 2.2.3 Evaluation of Potential Accidents 2.2.3.1 Determination of Design Basis Events 2.2.3.1.1 Liquified Petroleum Gas - Storage Site Explosion 2.2.3.1.2 Delayed Vapor Explosion 2.2.3.1.3 Aircraft Hazards 2.2.3.1.4 Toxic Gases 2.2.3.1.5 Fires 2.2.4 References 2.3 Meteorology 2.3.1 Regional Climatology 2.3.1.1 General Climate 2.3.1.2 Regional Meteorological Conditions for Design and Operating Bases 2.3.2 Local Meteorology 2.3.2.1 Normal and Extreme Values of Meteorological Parameters 2.3.2.2 Potential Influence of the Plant and Its Facilities on Local Meteorology
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - ii 2.3.2.3 Local Meteorological Conditions for Design and Operating Bases 2.3.3 Onsite Meteorological Measurements Program 2.3.3.1 December 17, 1975 through December 16, 1977 2.3.3.2 January 6, 1985 through June 13, 1996 2.3.3.3 June 13, 1996 through Present 2.3.4 Short-Term (Accident) Diffusion Estimates 2.3.4.1 Objectives 2.3.4.2 Calculations 2.3.5 Long Term (Routine) Diffusion Estimates 2.3.5.1 Objectives 2.3.5.2 Calculations 2.3.6 Other Diffusion Considerations 2.3.7 Summary of Offsite Diffusion Estimates 2.3.8 References 2.4 Hydrologic Engineering 2.4.1 Hydrologic Description 2.4.1.1 Site and Facilities 2.4.1.2 Hydrosphere 2.4.2 Floods 2.4.2.1 Flood History 2.4.2.2 Flood Design Considerations 2.4.2.3 Effects of Local Intense Precipitation 2.4.2.3.1 General 2.4.2.3.2 Probable Maximum Precipitation and Runoff Models 2.4.2.3.3 Powerhouse Yard 2.4.2.3.4 Switchyard 2.4.2.3.5 Cooling Tower Yard 2.4.2.3.6 Site Evaluation Using HMR 51 and 52 2.4.3 Probable Maximum Flood (PMF) on Streams and Rivers 2.4.3.1 Probable Maximum Precipitation 2.4.3.2 Precipitation Losses 2.4.3.3 Runoff and Stream Course Models 2.4.3.4 Probable Maximum Flood Flow 2.4.3.5 Water Level Determination 2.4.3.6 Coincident Wind Wave Activity 2.4.3.7 Regulatory Guide 1.59 2.4.4 Potential Dam Failures 2.4.4.1 Dam Failure Permutations 2.4.4.2 Unsteady Flow Analysis of Potential Dam Failures 2.4.4.3 Water Level at Plant Site 2.4.5 Probable Maximum Surge and Seiche Flooding 2.4.5.1 Probable Maximum Winds and Associated Meteorological Parameters 2.4.5.2 Surge and Seiche Water Levels 2.4.5.3 Wave Action 2.4.5.4 Resonance 2.4.5.5 Protective Structures 2.4.6 Probable Maximum Tsunami Flooding 2.4.7 Ice Effects 2.4.8 Cooling Water Canals and Reservoirs 2.4.9 Channel Diversions 2.4.10 Flooding Protection Requirements 2.4.11 Low Water Considerations 2.4.11.1 Low Flow in Streams 2.4.11.2 Low Water Resulting from Surges, Seiches, or Tsunami 2.4.11.3 Historical Low Water
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - iii 2.4.11.4 Future Control 2.4.11.5 Plant Requirements 2.4.11.6 Heat Sink Dependability Requirements 2.4.12 Dispersion, Dilution and Travel Times of Accidental Releases of Liquid Effluents in Surface Waters 2.4.13 Groundwater 2.4.13.1 Description And On Site Use 2.4.13.1.1 Regional Groundwater Conditions 2.4.13.1.2 Station Groundwater Use 2.4.13.2 Sources 2.4.13.2.1 Groundwater Users 2.4.13.2.2 Program of Investigation 2.4.13.2.3 General Geologic Conditions of the Site 2.4.13.2.4 Groundwater Levels 2.4.13.2.5 Permeability Tests 2.4.13.2.6 Constant-Discharge Pumping Tests 2.4.13.2.7 Groundwater Quality 2.4.13.3 Accident Effects 2.4.13.3.1 Contaminant Transport Model 2.4.13.3.2 Ion Exchange Capacity 2.4.13.3.3 Transport of Postulated Contaminants 2.4.13.4 Monitoring or Safeguard Requirements 2.4.13.5 Design Basis for Subsurface Hydrostatic Loading 2.4.14 Selected Licensee Commitments and Emergency Operation Requirements 2.4.15 References 2.5 Geology and Seismology 2.5.1 Basic Geologic and Seismic Information 2.5.1.1 Regional Geology 2.5.1.1.1 Geologic History 2.5.1.1.2 Physiographic, Lithologic, Stratigraphic and Structural Settings 2.5.1.2 Site Geology 2.5.1.2.1 Geology of the Subregion 2.5.1.2.2 Geology of the Site 2.5.2 Vibratory Ground Motion 2.5.2.1 Seismicity 2.5.2.2 Geologic Structure and Tectonic Activity 2.5.2.3 Correlation of Earthquake Activity With Geologic Structure 2.5.2.3.1 Piedmont and Upper Coastal Plain 2.5.2.3.2 Blue Ridge 2.5.2.3.3 Charleston Epicentral Area 2.5.2.3.4 Deformed Appalachian Highlands 2.5.2.3.5 New Madrid Faulted Zone 2.5.2.3.6 Central Stable Region 2.5.2.3.7 Florida Platform and Lower Coastal Plain 2.5.2.4 Maximum Earthquake 2.5.2.5 Seismic Wave Transmission Characteristics of the Site 2.5.2.6 Safe Shutdown Earthquake 2.5.2.7 Operating Basis Earthquake 2.5.2.8 Design Response Spectra 2.5.3 Surface Faulting 2.5.3.1 Geologic Conditions of the Site 2.5.3.2 Evidence of Fault Offset 2.5.3.3 Earthquakes Associated With Capable Faults 2.5.3.4 Investigation of Capable Faults 2.5.3.5 Correlation of Epicenters With Capable Faults
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - iv 2.5.3.6 Description of Capable Faults 2.5.3.7 Zone Requiring Detailed Faulting Investigation 2.5.3.8 Results of Faulting Investigation 2.5.4 Stability of Subsurface Materials and Foundations 2.5.4.1 Geologic Features 2.5.4.2 Properties of Subsurface Materials 2.5.4.2.1 Field Investigation 2.5.4.2.2 Laboratory Testing 2.5.4.2.3 Generalized Subsurface Profile 2.5.4.2.4 Classification and Engineering Properties of Soil and Rock 2.5.4.3 Exploration 2.5.4.4 Geophysical Surveys 2.5.4.5 Excavations and Backfill 2.5.4.5.1 Excavations 2.5.4.5.2 Dewatering 2.5.4.5.3 Foundation Preparation 2.5.4.5.4 Backfill 2.5.4.6 Groundwater Conditions 2.5.4.7 Response of Soil and Rock to Dynamic Loading 2.5.4.8 Liquefaction Potential 2.5.4.8.1 Earth Backfill 2.5.4.8.2 Granular Backfill 2.5.4.8.3 Residual Soil 2.5.4.8.4 Alluvial Soil 2.5.4.9 Earthquake Design Basis 2.5.4.10 Static Stability 2.5.4.10.1 Category I Mat Foundations on Rock 2.5.4.10.2 Category I Mat Foundations On Partially Weathered Rock 2.5.4.10.3 Category I Structures on Soil 2.5.4.10.4 Subsurface Static Lateral Loading 2.5.4.11 Design Criteria 2.5.4.12 Techniques to Improve Subsurface Conditions 2.5.4.13 Subsurface Instrumentation 2.5.4.14 Construction Notes 2.5.5 Stability of Slopes 2.5.6 Embankments and Dams 2.5.6.1 General 2.5.6.1.1 Standby Nuclear Service Water Pond Dam 2.5.6.2 Exploration 2.5.6.2.1 SNSW Dam Foundation Materials 2.5.6.2.2 Embankment Material 2.5.6.2.3 Geologic Features of the Foundation Materials 2.5.6.3 Foundation and Abutments 2.5.6.3.1 General Treatment 2.5.6.3.2 Construction Specification 2.5.6.4 Embankment 2.5.6.4.1 Embankment Features 2.5.6.4.2 Embankment Material Properties 2.5.6.4.3 Slope Protection 2.5.6.4.4 Settlement and Overbuild 2.5.6.4.5 Fill Placement 2.5.6.4.6 Protection Required of Fill Surfaces and Stockpiles During Construction 2.5.6.4.7 Blanket Drain 2.5.6.4.8 Special Fill Placement Activities 2.5.6.4.9 Construction Specifications 2.5.6.4.10 Significant or Unusual Construction Activities
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - v 2.5.6.5 Slope Stability 2.5.6.5.1 Stability Criteria 2.5.6.5.2 Design Methods 2.5.6.5.3 Static Stability Evaluation 2.5.6.5.4 Dynamic Stability Evaluation 2.5.6.5.5 Conclusions 2.5.6.6 Seepage Control 2.5.6.6.1 Permeability of Foundation and Embankment Soils 2.5.6.6.2 Seepage Analysis 2.5.6.6.3 Embankment Drainage 2.5.6.7 Diversion and Closure 2.5.6.8 Performance Monitoring 2.5.6.8.1 Instrumentation 2.5.6.8.2 Inspection 2.5.6.9 Construction Notes 2.5.6.10 Operational Notes 2.5.7 References 2.6 Review and Evaluation of Recent Geologic Information With Specific Reference to the Charleston Epicentral Area 2.6.1 Glossary 2.6.2 Review and Evaluation of Recent Geologic Information With Specific Reference to the Charleston Epicentral Area 2.6.3 Tectonic Evolution of the Southern Appalachians 2.6.4 Tectonic Models of Charleston Seismicity 2.6.4.1 Decollement Reactivation 2.6.4.2 Stress Amplification at Margins of Mafic Plutons 2.6.4.3 Reactivation of Steep Basement Faults 2.6.5 Summary and Conclusions 2.6.6 References
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - vi List of Tables Table 2-1. 1977 Population 0-5 Miles (0-8 km)
Table 2-2. 1970 Population Distribution 0-10 Miles (0-16.1 km)
Table 2-3. 1980 Projected Population Distribution 0-10 Miles (0-16.1 km)
Table 2-4. 1981 Projected Population Distribution (Initial Expected Year of Plant Start-up) 0-10 Miles (0-16.1 km)
Table 2-5. 1990 Projected Population Distribution 0-10 Miles (0-16.1 km)
Table 2-6. 2000 Projected Population Distribution 0-10 Miles (0-16.1 km)
Table 2-7. 2010 Projected Population Distribution 0-10 Miles (0-16.1 km)
Table 2-8. 2020 Projected Population Distribution 0-10 Miles (0-16.1 km)
Table 2-9. 1970 Population Distribution 0-50 Miles (0-80.4 km)
Table 2-10. 1980 Projected Population Distribution 0-50 Miles (0-80.4 km)
Table 2-11. 1981 Projected Population Distribution (Initial Expected Year of Plant Start-up) 0-50 Miles (0-80.4 km)
Table 2-12. 1990 Projected Population Distribution 0-50 Miles (0-80.4 km)
Table 2-13. 2000 Projected Population Distribution 0-50 Miles (0-80.4 km)
Table 2-14. 2010 Projected Population Distribution 0-50 Miles (0-80.4 km)
Table 2-15. 2020 Projected Population Distribution 0-50 Miles (0-80.4 km)
Table 2-16. 1977 Seasonal Recreational Transient Population Table 2-17. 1977 Average Daily Recreational Transient Population Table 2-18. 1977 Daily Industrial Transient Population Table 2-19. Peak Day Transient Population 0-5 Miles (0-8 km)
Table 2-20. 1977 Transient Population Recreation Areas and Facilities Table 2-21. 1977 Transient Population Industrial and Institutional Table 2-22. Description of Nearby and Significant Industries Table 2-23. Annual Aircraft Operations Douglas Municipal Airport. Extracted from Final Addendum to Environmental Impact Statement, Construction of Runway 18R/36L, Douglas Municipal Airport, Charlotte, N. C. Feb. 1977.
Table 2-24. Aviation Routes - Annual Usage. (FAA Atlanta ARTC Center, letter dated January 3, 1978)
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - vii Table 2-25. Hazardous Chemicals Received at Nearby Facilities Table 2-26. Hazardous Chemicals Shipped and Received by Nearby Industries Table 2-27. Tropical Cyclones Affecting the Site Area, 1964-1998 Table 2-28. Tornado Occurrences 1956-1980 (2° Square Centered on Site)
Table 2-29. Catawba Nuclear Station Vicinity Climatology Table 2-30. Catawba Nuclear Station Onsite Data January 1, 1976 - December 31, 1977 Table 2-31. Structural Design Meteorological Data Table 2-32. Ultimate Heat Sink Meteorological Data Worst Cooling Periods Table 2-33. Ultimate Heat Sink Meteorological Data Highest Evaporation Periods Table 2-34. Catawba Semiannual Calibration Listing Table 2-35. 1975 - 1977 Wind Occurrences (10 m)
Table 2-36. Hourly Dilution Factors at the Exclusion Area Boundary for Each Sector Table 2-37. Hourly Dilution Factors at the Low Population Zone Boundary for Selected Sectors Table 2-38. Cumulative Frequency Distribution at Integer Percentile Levels - 8 Hour Average Table 2-39. Cumulative Frequency Distribution at Integer Percentile Levels - 16 Hour Average Table 2-40. Cumulative Frequency Distribution at Integer Percentile Levels - 72 Hour Average Table 2-41. Cumulative Frequency Distribution at Integer Percentile Levels - 624 Hour Average Table 2-42. Annual Average X/Q Values to 50 Miles - Sector Average Table 2-43. Annual Average X/Q Values with Depletion to 50 Miles - Sector Average Table 2-44. 1975 - 1977 Wind Occurrences (40 m)
Table 2-45. Annual Average D/Q Values to 50 Miles - Sector Average Table 2-46. Catawba Nuclear Station Annual Average X/Q Values at Intake Vents Control Room X/Q Values for Accident Conditions Table 2-47. Dilution Factors for Accident Releases Table 2-48. Catawba Nuclear Station Dilution Factors for Routine Releases - Offsite Table 2-49. Relative Frequency Distribution Table 2-50. Charlotte, N. C. - Climatic Comparison Table 2-51. Dams on the Catawba River Table 2-52. Surface Water Intakes
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - viii Table 2-53. Flood Peak Return Period. U.S.G.S. Gage No. 1460 Near Rock Hill, South Carolina; D.A.
3050 Square Miles; Period of Record 1942-1969 Table 2-54. Deleted Per 1992 Update Table 2-55. Maximum Storms in the Southeast. Maximum Depth of Rainfall Over 3.020 Sq. Mi.
Table 2-56. Deleted Per 1990 Update Table 2-57. Deleted Per 1990 Update Table 2-58. Rainfall Subarea Coefficients Table 2-59. Parameters Used in Flood Routing Table 2-60. Storage And Discharge Values For Reservoirs For Rising Reservoir Condition Table 2-61. Maximum Reservoir Elevations Resulting From PMF Table 2-62. Reservoir Inflow, Outflow and Elevation for PMF Table 2-63. Coincident Wind/Wave Activity Table 2-64. Maximum Reservoir Elevations Resulting From Seismic Dam Failures With SPF Table 2-65. Maximum Reservoir Elevations Resulting From Seismic Failure at Cowans Ford Dam With SPS Table 2-66. Standby Nuclear Service Water Pond Table 2-67. Low Flow into Lake Wylie Table 2-68. Lake Wylie Minimum Surface Water Elevations1 Table 2-69. Summary of Residential Well Survey Data Immediate Vicinity of Site Table 2-70. Groundwater Intakes Table 2-71. River Bank Groundwater Intakes Table 2-72. Rock Permeability Test Results Table 2-73. Soil Permeability Test Results Table 2-74. Summary of Pumping Test No. 1 - Test Well A-85TW Table 2-75. Summary of Pumping Test No. 2 - Test Well A-48TW Table 2-76. Results of Physical and Chemical Tests on Groundwater Table 2-77. Ion Exchange Capacity of Soils Table 2-78. Groundwater Monitoring Program Table 2-79. Observation Well Monitoring
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - ix Table 2-80. Geologic Time Scale Table 2-81. Summary of Regional Tectonic Structures Table 2-82. Summary Of Stress Measurements Table 2-83. Site Geologic History As Releated to Regional Geology Table 2-84. Historic Earthquakes--Southeastern United States Table 2-85. Earthquakes With Epicenters Located Within 50 Miles of the Catawba Site Table 2-86. Historic Earthquakes Probably Felt At The Catawba Site Table 2-87. Summary of Rock And Water Levels In The Test Borings Table 2-88. Summary of Test Pits Table 2-89. Laboratory Permeability Test Results Compacted Plant Area Residual Soils Table 2-90. Summary of Significant Foundation Engineering Characteristics. (Table 2-90, Table 2-91 and Table 2-92 were originally one table.)
Table 2-91. Summary of Significant Foundation Engineering Characteristics. (Table 2-90, Table 2-91 and Table 2-92 were originally one table.)
Table 2-92. Summary of Significant Foundation Engineering Characteristics - Seismic Design Criteria.
(Table 2-90, Table 2-91 and Table 2-92 were originally one table.)
Table 2-93. Soil and Partially Weathered Rock Permeability Constant Head Field Tests Standby Nuclear Service Water Pond Dam Table 2-94. Summary of Design Static Shear Strength Parameters For Standby Nuclear Service Water Pond Dam Table 2-95. Rock Permeability Constant Head Field Tests Standby Nuclear Service Water Pond Dam Foundation Table 2-96. SNSW Pond Dam - Equipment Used During Dam Construction Table 2-97. Safety Factors of SNSW Pond Dam Under Static Loading Table 2-98. Laboratory Cyclic Triaxial Tests. Remolded Samples Kc=1.0 Table 2-99. Laboratory Cyclic Triaxial Tests. Remolded Samples Kc=1.5 Table 2-100. Laboratory Cyclic Triaxial Tests. Remolded Samples Table 2-101. Safety Factors of SNSW Pond Dam Under Earthquake Loading, Pseudo-Dynamic and Newmark Analyses Section Through Borings A117 & A118 Table 2-102. Soil Permeability Laboratory Test Results. Foundation (Undisturbed Samples)
Table 2-103. Soil Permeability Laboratory Test Results. Embankment (Remolded to 96% of Maximum Dry Density)
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - x Table 2-104. SNSW Pond Dam Filter Permeability Laboratory Test Results Table 2-105. Design Basis Values of Control Room X/Qs
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - xi List of Figures Figure 2-1. Regional Location Figure 2-2. Site Location Figure 2-3. Site Area Figure 2-4. Release Points Figure 2-5. Significant Population Groupings, Ten Miles Figure 2-6. Significant Population Groupings, Fifty Miles Figure 2-7. Vicinity Topography Figure 2-8. Five Mile (Eight Kilometer) Area Figure 2-9. Population Centers Figure 2-10. Cumulative Population, Year of Plant Start-Up Figure 2-11. Cumulative Population, End of Plant Life Figure 2-12. Locations and Routes Figure 2-13. Tornado Frequency 1916 - 1955 Figure 2-14. Visible Plume Length Frequency (June - November)
Figure 2-15. Visible Plume Length Frequency (December - May)
Figure 2-16. Vicinity Topography - Profile Figure 2-17. Relative Elevations of Meteorological Instruments Figure 2-18. Cumulative Frequency Distribution of Hourly Dilution Factors at the Exclusion Area Boundary (Northeast Sector) for Short Term Accident (X/Q)s Figure 2-19. Cumulative Frequency Distribution of Hourly Dilution Factors at the Low Population Zone Boundary (Northeast Sector) for Short Term Accident (X/Q)s Figure 2-20. Site Earthwork Figure 2-21. Plan-Profile of Catawba River Figure 2-22. Surface Water Users Figure 2-23. Yard Drainage Figure 2-24. Power Block Area Features Figure 2-25. Catch Basin Inlet Details
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - xii Figure 2-26. Administration and Security Fence Base Details Figure 2-27. Local Intense Precipitation PMP Flood Routing Using Puls Graphical Method Figure 2-28. PMP Flood Routing; Type I Inlets - 100% Efficient; Type II Inlets - 100% Efficient Figure 2-29. PMP Flood Routing; Type I Inlets - 100% Efficient; Type II Inlets - 50% Efficient Figure 2-30. PMP Flood Routing; Type I Inlets - 100% Efficient; Type II Inlets - 0% Efficient Figure 2-31. PMP Flood Routing; Switchyard Figure 2-32. PMP Flood Routing; Cooling Tower Yard Figure 2-33. Deleted Per 2009 Update Figure 2-34. Probable Maximum Precipitation Figure 2-35. Basin Subareas and Thiessen Polygons and Isohyetal Map for Probable Maximum Precipitation Figure 2-36. Lag Time Figure 2-37. Basin Subarea Unit Hydrographs Figure 2-38. Reservoir Inflows, Outflows, and Elevations for Wylie Reservoir Figure 2-39. PMF Hydrograph, Lake Wylie Figure 2-40. Fetch Map Figure 2-41. Wylie Reservoir Inflows, Outflows and Elevations Figure 2-42. SNSW PMF Hydrographs Figure 2-43. NSW Pump House Sections Figure 2-44. LPSW Intake Structure Figure 2-45. NSW and SNSW Discharge Structures Figure 2-46. SNSW Intake Structure Figure 2-47. Location of Wells Surveyed Figure 2-48. Groundwater Users Figure 2-49. River Bank Wells Figure 2-50. Reconstruction Groundwater Contour Map Figure 2-51. Site Cross-Section Figure 2-52. Monthly Groundwater Hydrograph (October, 1973 to September, 1977)
Figure 2-53. Groundwater Level, Daily Rainfall and Lake Level
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - xiii Figure 2-54. Schematic Equipment Arrangement for Rock Permeability Testing Figure 2-55. Schematic Equipment Arrangement for Soil Permeability Testing Figure 2-56. Pumping Test No. 1 - A85-TW Location, Plan Figure 2-57. Pumping Test No. 2 - A48TW Location, Plan Figure 2-58. Core of Depression - Pumping Test 1 Figure 2-59. Core of Depression - Pumping Test 2 Figure 2-60. Auxiliary and Reactor Building Groundwater Drainage System Figure 2-61. Auxiliary and Reactor Building Groundwater Drainage System Figure 2-62. Plot of Water Level at Structure Versus Rebound Time Figure 2-63. Auxiliary and Reactor Building Groundwater Drainage System Figure 2-64. Auxiliary and Reactor Building Groundwater Drainage System Figure 2-65. Powerhouse Foundation Figure 2-66. Location of Compacted Backfill (Powerhouse Area)
Figure 2-67. Regional Physiographic Map Figure 2-68. Regional Geologic Map Figure 2-69. Regional Tectonic Map Figure 2-70. Regional Geologic Cross-Section Figure 2-71. Regional Aeromagnetic Map Figure 2-72. Regional Bouguer Gravity Anomaly Map Figure 2-73. Subregional Geologic Index Figure 2-74. Regional Linear Features Map Figure 2-75. Subregional Geologic Map Figure 2-76. Subregional Drainage Map with Linear Stream Segments Figure 2-77. Regional In Situ Stresses Figure 2-78. Boring Location Plan Figure 2-79. Boring Location Plan Figure 2-80. Subsurface Profiles Figure 2-81. Subsurface Profiles
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - xiv Figure 2-82. Subsurface Profiles Figure 2-83. Subsurface Profiles Figure 2-84. Subsurface Profiles Figure 2-85. Subsurface Profiles Figure 2-86. Subsurface Profiles Figure 2-87. Subsurface Profiles Figure 2-88. Subsurface Profiles Figure 2-89. Subsurface Profiles Figure 2-90. Subsurface Profiles Figure 2-91. Subsurface Profiles Figure 2-92. Subsurface Profiles Figure 2-93. Subsurface Profiles Figure 2-94. Seismic Refraction Traverses Figure 2-95. Seismic Refraction Traverses Figure 2-96. Seismic Refraction Traverses Figure 2-97. Seismic Refraction Traverses Figure 2-98. Uphole Seismic Data, Boring A-61 Figure 2-99. Uphole Seismic Data, Boring A-63 Figure 2-100. Contours on Top of Continous Rock Figure 2-101. Equal Area Projections of Poles to Planes - Powerhouse Excavation Figure 2-102. Equal Area Projections of Poles to Planes - SNSW Pond Dam Excavations Figure 2-103. Geologic Map of Powerhouse Excavation - Reactor and Auxiliary Building Area Figure 2-104. Geologic Map of Powerhouse Excavation - Turbine and Service Building Area Figure 2-105. Geologic Map of SNSW Pond Dam Excavation Figure 2-106. Regional Epicenter Map Figure 2-107. Epicenters of Earthquakes Located Within 50-Mile Radius of the Site Figure 2-108. Epicenters of Earthquakes Probably Felt at the Catawba Site Figure 2-109. Acceleration - Intensity Relationships
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - xv Figure 2-110. Design Response Spectra,.08g Figure 2-111. Design Response Spectra,.08g Figure 2-112. Design Response Spectra,.15g Figure 2-113. Design Response Spectra, 25 feet of Overburden,.08g at Rock Figure 2-114. Actual Foundation Conditions at Structures Designed Using Response Spectrum Computed for 25 FT. of Overburden (Fill)
Figure 2-115. Soil Columns for Unit 2 Fuel Oil Storage Tanks and Unit 2 Above Ground Storage Tank Figure 2-116. Comparison of Computed Response Spectra and Design Response Spectrum Figure 2-117. Location Plan for Seismic Refraction Lines on Rock In Powehouse Excavation Figure 2-118. Results of Seismic Refraction Lines - NSW Intake Structure and NSW Pump Structure Figure 2-119. Seismic Refraction Profiles - NSW Pipe Lines Figure 2-120. Results of Seismic Refraction Lines in Powerhouse Excavation Figure 2-121. Summary of Grain Size Distribution of Residual Soils from Adamellite Figure 2-122. Summary of Optimum Moisture and Maximum Dry Density of Residual Soils from Plant Area Figure 2-123. Typical Consolidation Test Results of Compacted Residual Soils from Plant Area Figure 2-124. Summary of Consolidation Characteristics of Partially Weathered Rock Materials Figure 2-125. Summary of Mohr Diagrams of Compacted Residual Soils from Plant Area Figure 2-126. Summary of Mohr Diagrams for Partially Weathered Rock Materials Figure 2-127. Summary of Stress-Strain Data on Rock Core Figure 2-128. Summary of Rock Test Data From Plant Area Substructure Excavation Figure 2-129. Dynamic Moduli of Foundation Materials - Plant Area Figure 2-130. Contours of Plant Excavation Bottom at Start of Blasting Figure 2-131. Soil Column Analysis - NSW Pipeline Figure 2-132. Sections Through Powerhouse Figure 2-133. Cross Sections Through NSW and SNSW Pump Structure and Diesel Fuel Oil Storage Tanks Figure 2-134. Selected Subsurface Profiles Through NSW Pipe Lines Figure 2-135. Earth Pressure Distribution on Subsurface Walls Figure 2-136. Number of Moisture Tests Versus Moisture Content for Compacted Group I Backfill
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - xvi Figure 2-137. Number of Compaction Tests Versus Percent Compaction for Group I Backfill Figure 2-138. Number of Compaction Tests Versus Relative Density for Granular Backfill Figure 2-139. Nuclear Service Water System Yard Layout - Longitudinal Profiles Figure 2-140. Typical Section, Standby Nuclear Service Water Pond Dam Figure 2-141. Standby Nuclear Service Water Pond Dam, Details Figure 2-142. Plan, Standby Nuclear Service Water Pond Dam Figure 2-143. Profile of Standby Nuclear Service Water Pond Outlet Works Figure 2-144. Boring Location Plan, Standby Nuclear Service Water Pond Dam Figure 2-145. Section Along Axis of Plan Standby Nuclear Service Water Pond Dam Figure 2-146. Section A-87 Through A-96, South of Standby Nuclear Service Water Pond Dam Figure 2-147. Section A-103 S1 Through A-103 S2, Standby Nuclear Service Water Pond Dam Figure 2-148. Standby Nuclear Service Water Pond Dam, Section A-116 Through A-119 Figure 2-149. Standby Nuclear Service Water Pond Dam, Section A-117 Through A-118 Figure 2-150. Summary of Grain Size Distribution, Residual Soils at Standby Nuclear Service Water Pond Dam - Foundation Saprolites Figure 2-151. Summary of Mohr Diagrams for Foundation Soils, Standby Nuclear Service Water Pond Dam Figure 2-152. Cyclic Triaxial Strength Results Figure 2-153. Cyclic Strength of Soils Figure 2-154. Summary of Consolidation Characteristics for Residual Soils, Standby Nuclear Service Water Pond Dam Figure 2-155. Cross Hole Seismic Data, Boring A-103 Figure 2-156. Cross Hole Seismic Data, Boring A-105 Figure 2-157. Dynamic Tests on Foundation Soils Figure 2-158. Summary of Mohr Diagrams of Partially Weathered Rock Figure 2-159. Standby Nuclear Service Water Pond Dam Foundation Plan Figure 2-160. Standby Nuclear Service Water Pond Dam Foundation Bedrock-South Abutment Figure 2-161. Standby Nuclear Service Water Pond Dam Fill Sources and Quantities Placed Versus Date Figure 2-162. Summary Grain Size Distribution, Residual Sands - General Borrow Areas Investigated for PSAR
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - xvii Figure 2-163. Summary Grain Size Distribution Residual Soils-Silty Sands, SNSW Pond Dam Embankment Borrow Area Utilized Figure 2-164. Earth Embankment Material, Grain Size Distribution, Standby Nuclear Service Water Pond Dam Figure 2-165. Optimum Moisture, Maximum Dry Density for Soil During PSAR Investigation and Borrow Area Studies, Standby Nuclear Service Water Pond Dam Figure 2-166. Dry Density versus Optimum Moisture Content, From Proctor Tests on Soils Placed on SNSW Pond Dam Figure 2-167. Summary of Mohr Diagrams of Structural Fill Obtained During PSAR Investigations Figure 2-168. Summary of Mohr Diagrams of Structural Fill from Borrow Areas Used for Standby Nuclear Service Water Pond Dam Embankment Figure 2-169. Summary of Mohr Diagrams for Structural Fill Sampled from Standby Nuclear Service Water Pond Embankment Figure 2-170. Distribution of Field Density Tests with Percent Compaction on Standby Nuclear Service Water Pond Dam Embankment Figure 2-171. Typical Consolidation Tests of Compacted Soils Obtained During PSAR Investigation and from Borrow Areas Used for Standby Nuclear Service Water Pond Dam Embankment Figure 2-172. Slope Protection, Filter layer, Grain Size Distribution Figure 2-173. Number of Moisture Tests Versus Moisture Content for Standby Nuclear Service Water Pond Dam Figure 2-174. Number of Moisture Tests Versus Moisture Content for Standby Nuclear Service Water Pond Dam Figure 2-175. Number of Compaction Tests Versus Percent Compaction for Standby Nuclear Service Water Pond Dam Figure 2-176. Blanket Drain Course. Filter, Density Test Distribution Figure 2-177. Blanket Drain Top Layer Fine Filter Density Test Distribution Figure 2-178. Blanket Drain, Coarse Fitler, Grain Size Distribution Figure 2-179. Blanket Drain, Fine Filter-Top Layer, Grain Size Distribution Figure 2-180. Blanket Drain, Fine Filter-Bottom Layer, Grain Size Distribution Figure 2-181. Cross-Section for Stability Analysis Figure 2-182. Dynamic Modulus for Embankment and Foundation Materials Figure 2-183. Dynamic Modulus and Combining Pressure for Foundation Soils - Saprolites Figure 2-184. Failure Surfaces of Lowest Safety Factor - Static Analysis
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2 - xviii Figure 2-185. Finite Element Representation Figure 2-186. Static Finite Element Analysis, Vertical Effective Normal Stress in Embankment Figure 2-187. Static Finite Element Analysis, Horizontal Shear Stress in Embankment Figure 2-188. Static Finite Element Analysis, Principal Stress Ratios in Embankment Figure 2-189. Cyclic Triaxial Tests - Typical Strain and Pore Pressure Data (Remolded Samples)
Figure 2-190. Variation of Damping for Embankment and Foundation Materials Figure 2-191. Variation of Modulus for Embankment and Foundation Materials Figure 2-192. Dynamic Tests on Embankment Materials Figure 2-193. Time History of Synthetic EQ-1 Figure 2-194. Time History of Synthetic EQ-2 Figure 2-195. Time History of Synthetic EQ-3 Figure 2-196. Time History of Synthetic EQ-4 Figure 2-197. Failure Potential Under Cyclic Loading Figure 2-198. Seismic Coefficient SNSW Pond Dam Figure 2-199. Standby Nuclear Service Water Pond Dam Instrumentation Details Figure 2-200. Local Safety Factors Against 5% Strain Figure 2-201. SNSW Pond Dam Piezometric Readings Figure 2-202. SNSW Pond Dam Settlement Versus Time Figure 2-203. SNSW Pond Dam Crest Elevation Figure 2-204. Design Response Spectrum for Direct Generation
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.0 - 1 2.0 Site Characteristics Information provided in this chapter on the demographic, geologic, seismologic, hydrologic, and meteorologic characteristics of the Catawba Nuclear Station site and vicinity indicates how these characteristics influence station design and operating criteria and show the adequacy of the site for the station. Information pertaining to transportation routes, industrial and manufacturing facilities, site activities, and site control indicates the suitability of the site from a safety viewpoint.
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UFSAR Chapter 2 Catawba Nuclear Station 2.0 - 2 (09 OCT 2016)
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Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.1 - 1 2.1 Geography and Demography This section presents information on the Catawba Nuclear Station site location and description, boundaries for establishing effluent release limits, exclusion area control, and population distribution within 50 miles of the site.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.1.1 Site Location and Description 2.1.1.1 Specification of Location Catawba Nuclear Station is located in the north central portion of South Carolina approximately six miles north of Rock Hill and adjacent to Lake Wylie, as shown on Figure 2-2. The station center is located at latitude 35 degrees -3 minutes-5 seconds north and longitude 81 degrees-4 minutes-10 seconds west. The corresponding Universal Transverse Mercator Coordinates are E 493, 660 and N 3, 878, 558, zone 17.
The two Catawba reactors are located as follows:
Unit 1:
35°-03'-04" Lat., 81°-04'-10" long. UTM N-3, 878, 509 E-493, 660 Unit 2:
35°-03'-07" Lat., 81°-04'-10" long. UTM N-3, 878, 607 E-493, 660 Figure 2-2, the site location map, shows the station location with respect to local features.
The site is located in the northeastern portion of York County on a peninsula bounded by Beaver Dam Creek to the north, Big Allison Creek to the south, the main body of Lake Wylie to the east, and private property to the west. The Duke Power Company Wylie Dam and Hydroelectric Station are located approximately 4.5 mi (7.2 km) southeast of the site. Rock Hill, South Carolina and Charlotte, North Carolina are the nearest large cities. The city limit of Rock Hill is located approximately 3.5 mi (5.6 km) south-southeast of the site and the Charlotte city limit is located approximately 7.7 mi (12.4 km) east northeast of the site.
2.1.1.2 Site Area Figure 2-3, the site area map, shows the 391 acre (158 hectare) site, the site boundary, location and orientation of principal station structures, the 2500 ft exclusion radius, roads adjacent to the site, and locations, within the exclusion area boundary, of the visitors overlook, provided by Duke, and the Concord Cemetery property.
The site boundary lines are the same as the permanent perimeter fence. The perimeter fence erected around the immediate station area is shown on Figure 2-3. There are no industrial, commercial, institutional, or residential structures within the site area.
2.1.1.3 Boundaries for Establishing Effluent Release Limits The exclusion area boundary is set as the boundary for gaseous effluent release limits. The exclusion area boundary is formed by a 2500 ft (762 m) radius centered on the Reactor Building's centerlines as shown on Figure 2-3. Access to the area is controlled for security and radiation protection purposes by regular and routine patrols by security guards. All authorized
UFSAR Chapter 2 Catawba Nuclear Station 2.1 - 2 (09 OCT 2016) visitors are under the control of station personnel and trespassers are removed. There are no permanent residences within the exclusion boundary.
The liquid effluent discharge location is set as the boundary for liquid effluent release limits.
Liquid effluents are discharged into Lake Wylie at the station service water discharge structure and the Conventional Waste Water (WC) discharge location.
Distances of the effulent release points from the exclusion boundary are shown on Figure 2-4.
2.1.2 Exclusion Area Authority and Control 2.1.2.1 Authority Duke controls by ownership all land, mineral rights and easements within the exclusion boundary except the 1 Ac (.4 ha) Concord Cemetery Association property. An agreement with the Concord Cemetery Association exists concerning the use and occupancy of their property within the exclusion area. Arrangements are made for the immediate evacuation of the area in case of a major accident. The agreement provides Duke with the authority to evacuate the property of all persons and exclude all persons from the property when necessary or desirable in the interests of public health and safety and limits the use of the property to the purpose of conducting memorial or burial services. The site boundary is located approximately 1350 feet (412 m) east of the cemetery.
Coves of Lake Wylie protrude into the exclusion area at two locations: north and south of the station. Duke discourages public access to areas of the lake within the exclusion boundary.
Arrangements are made for the evacuation of those portions of the lake in case of a major accident.
2.1.2.2 Control of Activities Unrelated to Plant Operation Activities on the Concord Cemetery Assocation property consist of visits by relatively small numbers of people. There is no active church associated with the cemetery. Any temporary use, by swimmers or fishermen, of those portions of the lake within the exclusion boundary is unlikely to exceed a few hours per year. The approximately two acre (.8 ha) visitor's overlook area, shown on Figure 2-3, is a limited use picnic area having an estimated attendance of 27 visits per day. Duke maintains the right to limit access to and control evacuation from the exclusion area. The time required to evacuate the area is estimated to be 15 minutes.
2.1.2.3 Arrangements for Traffic Control There are no highways, railways, or waterways traversing the exclusion area.
2.1.2.4 Abandonment or Relocation of Roads County road 1132 is now relocated outside the exclusion area boundary as shown on Figure 2-
- 3.
2.1.3 Population Distribution For current information on population distribution see Reference 4, especially Figure J-5.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED Population within 50 mi (80.4 km) of Catawba is based on the 1970 census. Population distributions for 1980, 1981 (initial expected year of plant startup), and by decade to 2020 are
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.1 - 3 based on projections made by the United States Department of Commerce, Bureau of Economic Analysis (Reference 1 and 2). Though 2026 is the year of expected end of plant life, the year 2020 is used for end of plant life population distribution.
The disaggregation of the 1970 census county subdivisions into each radial sector is based on road densities, population accumulations, land usage, and general area information. The population distribution within 5 mi (8 km) of the site is based on an actual house count performed in 1971. The distribution of the projected populations is based on the ratio of the distributed 1970 populations within each radial sector to the total county population. An additional house count within 5 mi (8 km) of the site made in December 1977, is used to establish an adjusted distribution within Mecklenburg County, North Carolina and York County, South Carolina. The December 1977 population within 5 miles (8 km) is shown on Table 2-1.
2.1.3.1 Population Within 10 Miles Figure 2-5 identifies places of significant population groupings within 10 mi (16.1 km) of the station. Table 2-2 gives 1970 population distribution within ten miles. Projected population distributions by census decade (1980 through 2020) and for 1981 are shown on Table 2-3 through Table 2-8.
2.1.3.2 Population Between 10 and 50 Miles Places of significant population groupings in the area from 10 mi (16.1 km) to 50 mi (80.5 km) of the station are shown on Figure 2-6. Table 2-9 through Table 2-15 detail the 1970 and projected population distributions.
2.1.3.3 Transient Population Transient population within 5 mi (8 km) of Catawba Nuclear Station is primarily recreational on and along the shores of Lake Wylie. Industrial facilities in the northeastern quadrant and in the southeastern quadrant are the major sources of transient population between 5 and 10 mi (8 to 16.1 km). Carowinds Theme Park, located approximately 8 mi (12.8 km) to the east-north east, is the largest recreational area within 50 mi (80.5 km) of the site. Carowinds expected attendance in 1978 is 1,150,000 with a daily average attendance of 11,058. The projected increase in attendance to the year 2020 is insignificant (Reference 3).
Table 2-16 and Table 2-17 show 1977 seasonal and average daily recreational transient population distribution within 10 mi (16.1 km) of the station. Table 2-18 shows the daily industrial transient population distribution within 10 mi (16.1 km) of the site.
No large industries or businesses to provide job opportunities are located within 3.5 mi (5.6 km) of the site. A reduction of daily population in the vicinity of the station due to workers commuting to population centers where job opportunities exist is expected.
2.1.3.4 Low Population Zone The nearest boundary of Rock Hill, the closest population center, is located 3.5 miles (5.6 km) south-southeast of the site.
Due to the relatively low population density in the vicinity of the site and the size of Rock Hill, the Low Population Zone is an area extending from the Reactor Building's centerlines to a radius of 20,000 feet (6096 m). The Low Population Zone distance is well within the limits of Section 100.11(a) of 10 CFR Part 100.
UFSAR Chapter 2 Catawba Nuclear Station 2.1 - 4 (09 OCT 2016)
Figure 2-7 and Figure 2-8 show the topographic features and the transportation routes and facilities, including all industries and institutions within the Low Population Zone and to a distance of 5 mi (8 km). There are no prisons or hospitals within 5 mi (8 km) of the station.
Current and projected permanent population distributions by sector and 1 mi (1.6 km) radii to 5 mi (8 km) are shown on Table 2-1 through Table 2-8. The seasonal recreational transient population and daily industrial transient population for each sector to 5 miles (8 km) are shown on Table 2-16 and Table 2-18 respectively. The peak daily transient population distribution within 5 mi (8 km) of the site is shown on Table 2-19.
The annual, seasonal, and peak day attendance, and location of recreational areas and facilities within 5 mi (8 km) are shown on Table 2-20. The daily attendance and location of the two schools and two industries within 5 mi (8 km) of the site are listed on Table 2-21.
2.1.3.5 Population Center The nearest population center, as defined in 10 CFR Part 100 is Rock Hill, South Carolina with a 1970 population of 33,846 and a population density of approximately 2380 people per square mile. The city of Rock Hill dominates the 5 to 10 mile south-southeast sector. The 1970 population density for the 5 to 10 mile south-southeast sector is 2087 people per square mile.
There are no population groupings, including transient population, nearer the site having population distributions or densities higher than those in the 5 to 10 mile south-southeast sector.
Of the adjacent 4 to 5 mile sectors, the south-south east sector has the highest population density with 265 people per square mile. Therefore, the political boundary of Rock Hill, 5.8 mi (9.3 km) south-southeast of the Reactor Building's centerlines, is the nearest population center.
Figure 2-9 shows population centers within 100 miles of the site.
2.1.3.6 Population Density Figure 2-10 compares the cumulative projected population distributions within 50 mi (80.5 km) for the year of initial expected plant startup (1981) with a constant population density of 500 people per square mile. Figure 2-11 comparison of the cumulative projected population distribution within 50 mi (80.5 km) for the year 2020 (end of plant life is 2026) with a constant population density of 1000 people per square mile.
2.1.4 References
- 1. U. S. Department of Commerce, Bureau of Economic Analysis, Projections, Economic Activity in North Carolina, Series E Projection, April 1976.
- 2. U. S. Department of Commerce, Bureau of Economic Analysis, Projection, Economic Activity in South Carolina, Series E. Projection, April 1976.
- 3. Letter, March 1, 1978 from Crescent Land and Timber Co.
- 4. Duke Energy Corporation, Catawba Nuclear Station, Emergency Plan
- 5. Duke Energy Corporation, Catawba Nuclear Station, Offsite Dose Calculation Manual THIS IS THE LAST PAGE OF THE TEXT SECTION 2.1.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.2 - 1 2.2 Nearby Industrial, Transportation, and Military Facilities This section provides information to establish whether the effects of potential accidents at facilities in the site vicinity should be used as design basis events. (Reference 15 documents no significant concerns from nearby facilities. Emergency planning monitors offsite hazardous chemicals on a monthly basis.) (Reference 16).
2.2.1 Locations and Routes HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED Figure 2-12 shows the location of the two small manufacturing plants and the one airport within 5 miles (8 km) of Catawba. Significant facilities, transportation routes, and pipelines within 10 mi (16.1 km) are also shown. The only significant facilities greater than 10 miles (16.1 km) from the station are the Gastonia and Douglas Municipal Airports located 11 mi (17.7 km) north-northwest and 13 mi (20.9 km) north-northeast, respectively. Figure 2-7 indicates the topographic features within 5 mi (8 km) of the site.
Except for local National Guard and military reserve units, the closest of which is in Rock Hill, there are no major military facilities in the vicinity of the site. The nearest major facilities are Fort Jackson, Shaw Air Force Base, and McEntire Air Base, all of which are located 75 mi (120 km),
or more, south of the station. The nearest military firing and bombing ranges are located at or near these facilities.
There are no mining or quarrying operations within 5 mi (8 km) of Catawba.
2.2.2 Descriptions Descriptions of the nearby facilities are presented in this section and are based on published information, letters from facility personnel, and telephone conversations where applicable.
2.2.2.1 Description of Facilities Table 2-22 describes the nearby and significant industries and indicates the number of employees of the two industries within 5 mi (8 km) of Catawba and the three significant industries nearby.
The closest rail approach is a mainline track of the Southern Railway 4.5 mi (7.2 km) south-southwest of the station.
The major highways in the vicinity are Interstate 77, located 6 mi (9.7 km) east of the site and US 321 9 mi (14.5 km) to the west. South Carolina Highway 161, located 4.5 mi (7.2 km) south-southwest carries local traffic between Rock Hill and York and gives Rock Hill access to US 321. The closest highway approach to the site is South Carolina Highway 274 located l.3 mi (2.1 km) west of the site.
S.C. 274 is not heavily traveled. Its location relative to the major highways in the area indicates that the primary usage is local traffic and access to recreation areas on the west side of Lake Wylie.
Traffic on S.C. 274 (based on 1971 statistics) prior to any traffic increase due to Catawba construction, averages approximately 1500 vehicles per day in comparison with 3500 per day on S.C. 161, 4300 per day on US 321 and 9000 per day on I-77.
UFSAR Chapter 2 Catawba Nuclear Station 2.2 - 2 (09 OCT 2016) 2.2.2.2 Description of Products and Materials Neither of the manufacturing firms within 5 mi (8 km) of Catawba, EFP Products Division of Kent Moore Corporation and C & B Machine Products, store or transport any product which affects the operation of the station.
Rock Hill Printing and Finishing, Traybor Inc., and Celenese Fibers and Celenese Chemical Co.
are located 9.5 mi (15.3 km) southeast, 7.8 mi (12.6 km) southeast, and 7.3 mi (11.7 km) southeast respectively of Catawba. Table 2-25 and Table 2-26 list hazardous chemicals shipped or received by these facilities.
None of these industries ship or receive hazardous materials which could affect the station.
Due to the location of chemical suppliers and receivers, it is unlikely that hazardous chemicals will be transported nearer to Catawba than South Carolina Highway 161 and the Southern Railway, which parallel each other 4.5 mi (7.2 km) south-southwest of the station.
2.2.2.3 Pipelines 2.2.2.3.1 Natural Gas Pipelines Carolina Pipeline Company, an intrastate carrier of natural gas, owns one 12 in. API 5L Grade B pipe 6.5 mi (10.5 km) SSW and one 6 in. API 5L Grade B pipe 4.3 mi (6.9 km) south in the station vicinity. The 12 in. (0.3 m) pipe, installed in 1958, has a design pressure of 1372 psig (9.46 mpa) and a maximum operating pressure of 686 psig (4.73 mpa). The 6 in. (0.15 m) pipe, also installed in 1958, has a design pressure of 1986 psig (13.69 mpa) and a maximum operating pressure of 717 psig (4.94 mpa). The minimum depth of burial of both pipes is 30 in.
(0.76 m). Isolation valves in the vicinity are shown on Figure 2-12. These valves are manually operated plug valves. Pressure in the pipeline is telemetered to Columbia, South Carolina and monitored on a 24-hour basis. A maximum of 1 hour1.157407e-5 days <br />2.777778e-4 hours <br />1.653439e-6 weeks <br />3.805e-7 months <br /> is required to isolate any section of pipe.
Normally, natural gas is transported in the pipeline, but during heavy demand periods a propane-air mixture is injected into the natural gas at valve 4A. There is no current use or future plans to use the pipeline for higher than normal pressure storage. There are no plans to use the pipeline for transport of other products.
2.2.2.3.2 Propane Facilities Carolina-Transco Propane Company operates a liquid propane storage facility 6.5 mi (10.5 Km) west-southwest of Catawba. The facility consists of a 16 million gallon (60,560 m3) cavern, a 50 million gallon (189,250 m3) cavern put into service in April, 1979, two 45,000 gallon (170 m3) and one 90,000 gallon (341 m3) above ground storage tanks and truck and railroad loading facilities. There are currently two employees at the storage site. This number may increase to four depending on future usage and demand.
The cavern roofs are approximately 410 ft. (125 m) underground. The caverns are a system of 25 ft. (7.6 m) wide by 40 ft. (12.2 m) high corridors separated by 40 ft. (12.2 m) square pillars.
The rail siding of the facility will hold seven 30,000 gallons (114 m3) tank cars. Though virtually no propane is received by rail, shipments of up to 50 rail cars per year are made.
The injection capacity of the existing cavern by truck is 500,000 gallons (1893 m3) per day.
Truck shipments for 1979 totaled approximately 500. This number could increase substantially, upwards to 50 shipments per day if the demand for propane were to increase or in the event of severe weather.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.2 - 3 Virtually all propane is now received through a new 6 in. (0.15 m) pipeline entering the site from Bethune, S.C. and paralleling the existing natural gas pipeline operated by Carolina Pipeline Co.
(See Figure 2-12). This line is capable of delivery or shipment of 630,000 gallons (2385 m3) of liquid propane per day. This line operates at a maximum pressure of 1440 psig (9.93 mpag).
The test pressure is 1750 psig (12.07 mpsg). Valve stations are also given in Figure 2-12.
There are no connections between the propane storage facility and the existing natural gas pipeline.
The possibility of expansion of the existing caverns, or the mining of additional caverns closer to the Catawba site is not considered feasible. This statement is based on siting studies done by consultants for Carolina-Transco to find the current cavern site. According to these studies, the present location is the only site in the state that has a suitable rock formation for cavern mining, and the full extent of this formation was utilized to develop the two caverns in operation.
(Reference 13).
2.2.2.4 Waterways The only waterborne traffic on Lake Wylie is recreational. There are no upstream or downstream waterway exits and no commercial traffic.
2.2.2.5 Aircraft Activities There is one airport, Rock Hill, within 5 mi (8 km) of Catawba. The other airports of significance in the area are Charlotte/Douglas International, 13 mi (20.9 km) and Gastonia Municipal, 11 mi (17.7 km) from the station, respectively. These facilities are located on Figure 2-12.
The nearest military pilot training and practice area is the Gamecock MOA. There are no unusual activities associated with this area. Unarmed aircraft are put through aerobatic training maneuvers. The nearest boundary of Gamecock MOA to the station is 30 mi (48.3 km) southeast. Military training flight path IR-082 lies approximately 1.2 status miles (1.9 km) south of Catawba. No military ordinance is carried along this flight path.
2.2.2.5.1 Rock Hill Airport Rock Hill Airport is the nearest airport to the station. It has one paved 5000 ft. (1524 m) runway oriented 10°/190° and is not tower controlled. Aircraft are controlled by FAA Regulations within 2 mi (3.2 km) of the airport. The end of the runway is 4.2 mi (6.8 km) to the south of the station.
There are currently 40 aircraft permanently based at the facility. The largest permanently based aircraft is a Cessna 402 having a gross weight of approximately 6300 lbs (2858 kg). Three larger aircraft, the largest a Lockheed Jet Star weighing approximately 58,000 lbs (26,309 kg,)
utilize the airport a total of 50 times a year. Total annual air operations are estimated to be 70,000. Approximately 85 percent of the traffic is local and within a 25 mi (40.2 km) radius of the facility. Much of this activity is touch-and-go landing. There is no Air Carrier or Military Traffic at this facility.
There are no plans for runway expansion at the Rock Hill Airport. There have been no accidents at the Rock Hill Airport in the last 5 years.
2.2.2.5.2 Charlotte/Douglas International Airport Charlotte/Douglas International Airport, the largest airport within 50 miles (80.5 m) is located approximately 13 miles (20.9 Km) from the site. There are three paved runways in operation.
UFSAR Chapter 2 Catawba Nuclear Station 2.2 - 4 (09 OCT 2016)
Runway 05/23 is 7500 ft. (2286 m) long, runway 18L/36R is 8845 ft. (2695 m) long and runway 18R/36L is 10,000 ft. (3048 m) long.
Air operations at Douglas of 191,000 in 1975 are expected to increase to 272,000 in 1980 and 340,000 by 1985. Table 2-23 shows 1975 and projected annual air operations by category and percent mix by aircraft type. The actual number of flights recorded in the 1993 Yearly Traffic Summary for the Charlotte Air Traffic Tower was 531,415 flights as shown on Table 2-23. This included military flights, overflights, and traffic into and out of secondary airports.
The North Carolina Air National Guard has nine C-130 transport aircraft based at Douglas which account for approximately 3100 of the annual operations. No significant increase in National Guard operations is anticipated.
There is record of one fatal Air Carrier accident at Charlotte/Douglas in the last 10 years.
2.2.2.5.3 Gastonia Municipal Airport The Gastonia Municipal Airport, located 11 mi (17.7 km) north-northwest of Catawba has two runways. Runway 03/21 is a 3500 ft (1067 m) paved runway and runway 32/14 is a 3000 ft (914 m) sod strip. There is no control tower at Gastonia Municipal and aircraft are controlled by FAA Regulations within 2 miles (3.2 km) of the runway. There are 75 aircraft permantly based at Gastonia Municpal, the largest of which are of the Cessna 402 and Piper Navaho class. King Airs and small jets use the facility occasionally.
Current air operations are conservatively estimated by the airport management at 18,000 operations per year. Annual air operations are anticipated to increase at about 5 percent per year. Approximately 50 percent of the operations involve student touch-and-go landings.
Approximately 98 percent of the Gastonia Municipal traffic is by private aircraft, the remaining 2 percent being charter. There are no Air Carrier operations at this facility.
A 500 ft (152 m) addition to the paved runway is approved, but there is no schedule for construction.
The runway extension, when completed, allows Gastonia Municipal to handle small jets and turbo props weighing 13,000 to 14,000 lbs (5897 to 6350 kg).
There is no record of fatal accidents at Gastona Municipal in the last 10 years.
2.2.2.5.4 Flight Paths and Pilot Training Areas Victor Airways in the vicinity of the site and the closest military training route, IR-082, are shown on Figure 2-12.
The distance from Catawba to the centerline of each airway and route, the corridor width, and the 1977 and 2020 projected annual usage are shown on Table 2-24.
2.2.2.6 Projections of Industrial Growth Little or no industrial growth is expected within 5 mi (8 km) of the station. Industrial growth in the vicinity is generally expected to be in the currently industrialized areas located in the vicinity of Charlotte and Rock Hill at greater than 7 mi (11.3 km) from Catawba.
2.2.3 Evaluation of Potential Accidents The facilities, located as described and detailed in Section 2.2 are evaluated to determine whether potential accidents at these facilities are considered as a design basis events.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.2 - 5 2.2.3.1 Determination of Design Basis Events The design basis of components, systems, and structures at Catawba are described in Chapter
- 3. The determination of these design bases includes an evaluation of events external to the station, with a probability of occurrence of approximately 10-7 per year. Those events are detailed below.
2.2.3.1.1 Liquified Petroleum Gas - Storage Site Explosion Carolina Transco Propane Company operates a mined underground cavern for the storage of liquid propane approximately 6.5 mi (10.5 km) west-southwest of the site (Figure 2-12). Present storage capacity is 66 millon gallons (249,810 m3). This capacity is present in two caverns, one having a volume of 16 million gallons (60,560 m3) and the other having a volume of 50 million gallons (189,250 m3).
The two caverns are located some 410 ft (125 m) below the ground surface. This depth causes a hydrostatic pressure at the cavern level that is greater than the vapor pressure of the propane.
Any leakage is into rather than out of the cavern. Full capacity of the caverns permit a 2-3 ft (0.6-0.9 m) vapor area at the top of the caverns to assure that the caverns do not exert any hydrostatic pressures on the surrounding rock.
Water pressure in rock surrounding the caverns is measured indirectly by monitoring the height of the groundwater table over the caverns through wells drilled into bedrock and equipped with automatic water level recorders. The pressure of the propane is also monitored to permit operation at a safe pressure differential (Reference 2).
Access to the smaller cavern is through three shafts, one 48 in. (1.2 m) in diameter and two 20 in. (0.5 m) in diameter. The larger cavern has five access shafts, two 60 in. (1.5 m) in diameter and three 20 in. (0.5 m) in diameter (Reference 3). There are three above ground storage tanks, one having a capacity of 90,000 gallons (341 m3) and two others, each having a capacity of 45,000 gallons (170 m3) at the facility (Reference 4).
Drainage at the facility is to Fishing Creek which eventually enters the Catawba River well downstream of Lake Wylie. However, due to the relative low boiling temperature of propane, spills quickly evaporate and do not enter nearby streams.
Failure of either cavern, resulting in a complete release of all propane, is extremely remote.
An event as severe as an earthquake is not considered to fracture the cavern and release the propane since: 1) seismic wavelengths are comparatively long and each cavern moves together as a unit, and 2) in the unlikely event of a fracture, the fracture still contains water under sufficient pressure to prevent escape of the propane (Reference 2).
Propane is transported to and from the cavern by rail, truck and pipeline. Quantities of propane contained by transport vehicles are less than the volume of onsite tanks or the possible release from the caverns. Therefore, cavern site transport vehicle accidents are considered less severe than other potential accidents. The closest approach by truck to the Catawba Station is on Highway 274, approximately 1.3 mi (2.1 km) west. Reference 5, Table 2 lists the minimum safe distance from inhabited unbarricaded structures for a truck load of TNT as 1500 ft (457 m).
Since the TNT equivalent of a truckload of propane is less than one (Reference 6), and the closest line of approach is greater than 1500 ft (457 m) truck transport is not a hazard to the station.
Rail traffic carrying propane does not have access to the rail spur to Catawba due to the location of a manual locking device at the Tirzah junction. The closest straight line distance to the plant that such traffic can approach is approximately 4.5 mi (7.2 km). Reference 5, Table 2
UFSAR Chapter 2 Catawba Nuclear Station 2.2 - 6 (09 OCT 2016) lists the minimum safe distance for a railcar of TNT as 2100 ft (640 m) or for a three railcar load as 3000 ft (914 m). Again, considering the distance and relative energies at detonation, rail traffic does not pose a hazard to the station.
From the above, the three sources with the greatest potential for a propane accident are the above ground tanks, the pipeline, and the caverns. A postulated rupture of the largest above ground tank results in the release of 90,000 gallons (341 m3) of propane that quickly vaporizes.
A postulated rupture of the 6 in. (.15 m) diameter pipeline releasing for a conservatively estimated 100 minutes results in a total release of 43,750 gallons (166 m3). This release is assumed to occur at the cavern site since it is the point of closest approach to Catawba (See Figure 2-12). A postulated rupture of one 6 in. (.15 m) diameter cavern pipe releasing for a conservatively estimated 100 minutes results in a total release of 100,000 gallons (379 m3). The actual closure time to isolate a rupture by valving is estimated by the owner to be ten minutes.
The use of 100 minutes in the calculations represents a factor of ten safety margin (Reference 7). The incident overpressure resulting from the detonation of 100,000 gallons (379 m3) of vapor phase propane assumed to remain at the cavern site and totally within the flammability limits is less than one psi, well within the design basis (References 5 and 8).
2.2.3.1.2 Delayed Vapor Explosion For the accidental propane release condition postulated in 2.2.3.1.1, it is assumed that cavern site detonation will not occur. Under certain postulated meteorological conditions, this vapor cloud is transported offsite toward the Catawba Station. A mean wind velocity was used as "worst case" meteorology because of its effect on the initial assumed gas configuration and the resulting downwind concentrations. Considering mean wind velocity, propane density, and other parameters, calculations are performed to conservatively estimate the point of closest approach that a vapor cloud within the flammability limits of propane makes to Catawba, and remains in an amount such that detonation is possible.
These calculations show that approximately 140,000 lb mass (63,504 kg) of propane reaches a point toward Catawba from the cavern site of 6550 ft (2000 m). This is approximately 5 mi (8.4 km) from Catawba. The peak incident over pressure resulting from the detonation of the above mass of vapor phase propane at the above distance from Catawba, assuming the entire mass is within flammability limits is less than one psi, well within the design basis (References 5 and 8).
2.2.3.1.3 Aircraft Hazards The airports whose flights may potentially effect the Catawba Nuclear Station are the Charlotte/Douglas Internatonal Airport and the Rock Hill Airport (known as Bryant Field). The Charlotte/ Douglas Airport and Bryant Field are located approximately 4.2 statute miles (6.8 km) and 13 statute miles (20.9 km), respectively, from Catawba. The nearest public airways within 10 statue miles (16.1 km) of Catawba are airways V-37, V-54, and V-454. The nearest military training route within 5 statute miles (8.1 km) of Catawba is flight path IR-082.
The aircraft operations into and out of the Charlotte/Douglas Air Traffic Control Tower control area, including overflights and landings at secondary airports for 1993, are 531,415 flights. The military traffic along training flight path IR-082, controlled by the Naval Air Station located at Pensacola, Florida, is less than 5 flights per month for 1993.
The probability of an aircraft accident at Catawba based on methods of Reference 11 is approximately 107 (see Reference 14). A probability of this magnitude constitutes an acceptable risk.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.2 - 7 2.2.3.1.4 Toxic Gases Industries having bulk storage of chemicals capable of producing toxic gases in the vicinity of Catawba are identified in Table 2-25 and Table 2-26 and are located on Figure 2-12. Due to the location and properties of these chemicals, at a distance greater than 5 mi (8.0 km) from Catawba, and the location of the routes where transport of the chemicals occurs (Reference 12), they are not a hazard to the station. (Note - There is no chlorine stored at the York County Water Tank located on Concord Road.)
Additional information about the potential hazards from onside toxic gases is presented in section 6.4.
2.2.3.1.5 Fires Because there are no industrial-chemical plants or storage facilities, oil and gas pipelines (Sections 2.2.2.3.1 and 2.2.2.3.2) or transportation routes adjacent to the site, consequences from fires are not considered justifiable for impact evaluation. Brush and forest fires would be handled by the station and are not considered to cause any impact; therefore, these fires were not evaluated.
2.2.4 References HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED
- 1. Kightlinger, Ray M., Vice President, Carolina Transco-Propane Company, Inc., Letter dated January 17, 1978.
- 2. Environmental Impact Statement - Underground Liquid Propane Storage Cavern and Subsurface Facilities, Carolina-Transco Propane Company, Columbia, S. C., September, 1973.
- 3. Clark, J. M., General Manager, Carolina-Transco Propane Company, Inc., Columbia, S. C.,
Telecon of April 14, 1978.
- 4. Kightlinger, Ray M. Vice President, Carolina Transco-Propane Company, Inc., Telecon dated March 23, 1978.
- 5. Regulatory Guide 1.91, Evaluation of Explosions Postulated to Occur on Transportation Routes Near Nuclear Power Plant Sites, January, 1975.
- 6. Handbook of Chemistry and Physics, 53rd Edition, 1972-1973, The Chemical Rubber Co.,
- p. D235.
- 7. Clark, J. M., General Manager, Carolina-Transco Propane Company, Inc., Letter of March 14, 1978.
- 8. Department of Army Technical Manual, TM 5-1300, Structures to Resist the Effects of and Protection Against Explosion of Munitions, Fuels, and other Hazardous Mixtures, Part 4, October 28, 1968.
- 9. Springer, Rebecca, Assistant to Airport Manager, Interview of May 25, 1978.
- 10. Morris, Phillip, Operations Control Supervisior, Douglas Municipal Airport, Interview of May 25, 1978.
- 11. Standard Review Plan, Aircraft Hazards, Section 3.5.1.6, November 24, 1975.
- 12. Handbook of Laboratory Safety, 2nd Edition, The Chemical Rubber Company, Cleveland, Ohio, 1971, pp. 816-817.
UFSAR Chapter 2 Catawba Nuclear Station 2.2 - 8 (09 OCT 2016)
- 13. Clark, J.M., General Manager, Carolina-Transco Propance Company, Inc., Telecon of August 15, 1980.
- 14. Duke Power Company Catawba Nuclear Station IPEEE Submittal Report, Dated June 21, 1994.
- 15. Duke Energy Company Catawba Nuclear Station, Hazardous Chemical Assessment, CNC-1211.00-00-0131
- 16. Duke Energy Company Catawba Nuclear Station, Monthly Review of Hazardous Materials THIS IS THE LAST PAGE OF THE TEXT SECTION 2.2.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 1 2.3 Meteorology Meteorology is evaluated for use in structural design, in sizing the ultimate heat sink, in assessing the effects of heat dissipation facilities on the atmosphere and in consideration of environmental safeguards for gaseous releases. The following paragraphs summarize the atmospheric characteristics pertinent to these design bases.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.3.1 Regional Climatology 2.3.1.1 General Climate Synoptic features during winter effect rather frequent alternation between mild and cool periods with occasional outbreaks of cold air. Such intrusions of cold air, however, are modified in the crossing and descent of the Appalachian Mountains. Summers, noted for their greater persistence in flow pattern, experience fairly constant trajectories from the south and southwest with advection of maritime tropical air. Wintertime precipitation occurs primarily in connection with migratory low pressure systems. Recurrence and areal distribution, therefore, are reasonably uniform. Summer rains, on the contrary, are associated more with showers and thundershowers of the air mass variety, occasioned by intense and uneven heating of the earth's surface. Local meteorological (site) conditions are in general dominated by synoptic scale processes.
2.3.1.2 Regional Meteorological Conditions for Design and Operating Bases Note:
This section of the FSAR contains information on the design bases and design criteria of this system/structure.
Winter conditions, as a rule, are not conducive to the development of major snow storms. Long-term records for the area show highest 24 hour2.777778e-4 days <br />0.00667 hours <br />3.968254e-5 weeks <br />9.132e-6 months <br /> snowfall near 18 in. (Winston Salem, N. C.,
December, 1930) (Reference 1). The ice storm, a much more frequent occurrence, does effect considerable damage over limited areas and is expected several times a year. Typical accumulations range from one-quarter to one-half inch radial thickness. The 100 year mean return period accumulation is taken as 2.5 inches (Reference 2).
Spring, summer and autumn storms, phenomena of widespread consequence, are the major bearers of severe weather. For the area of North Carolina, South Carolina and their coastal waters, an average of one tropical storm per year and one hurricane every other year is computed based on a period of record of 63 years (1901-1963) (Reference 3). Within this period, seven years are void of any activity while nine years produce a combined total of three storms per year. Table 2-27 identifies hurricanes, tropical storms, and tropical depressions that have affected the site area since 1963 (Reference 24). Highest winds over the area are 110 miles per hour (fastest mile, Cape Hatteras, N. C., September, 1944) along the coast and 87 miles per hour (fastest mile, Charleston, S.C., September 22, 1989) for inland maxima (Reference 1 and later information from the National Weather Service). Maximum 24 hour2.777778e-4 days <br />0.00667 hours <br />3.968254e-5 weeks <br />9.132e-6 months <br /> rainfalls, again higher for coastal stations, are recorded near 15 in. along the coast (Cape Hatteras, N. C., June, 1949) to 9.2 in. inland (Greenville - Spartanburg, S. C., August 26, 1995)
(Reference 1 and later information from the National Weather Service). Figure 2-13 relates tornado frequency to two degree squares for the period 1916-1955 (Reference 4). Table 2-28
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 2 (09 OCT 2016) identifies tornadoes that have occurred over a two degree square area in the vicinity of the site from 1955 to 1980 (References 25 and 26). The F scale is determined from Fujita (Reference 27). For the site area a total of 223 tornados occurred per two degree square (square area about 125 miles by 125 miles). Put in terms of probability for a point (nuclear station), such a translation predicts a recurrence interval of 1610 years (Reference 5). Thunderstorms, with greater frequencies during the summer, occur about 46 days per year (from Charlotte, N. C.,
period of record 73 years) (Reference 1). Thunderstorm occurrence by season averages: 11 for spring (March-May), 29 for summer (June-August), 5 for fall (September-November) and 1 for winter (December-February) (Reference 6). Lightning strike frequency at the plant site is estimated from Marshall (Reference 28). Assumptions include the most electrostatically attractive surface type for a circular area with radius ten times the reactor building height.
Estimates of mean frequency by season are: 1.0 for spring, 2.8 for summer, 0.5 for fall and 0.1 for winter. Annual mean frequency then is estimated at 4.4. Associated hail can be expected about one day per year over inland areas as indicated in Reference 7.
Mean mixing height estimates are (Reference 8):
Season Morning (m)
Afternoon (m)
Winter 400 1000 Spring 400 1700 Summer 400 1800 Fall 300 1400 Specific meteorological values used in structural and other design considerations are presented in Section 2.3.2.3.
2.3.2 Local Meteorology 2.3.2.1 Normal and Extreme Values of Meteorological Parameters Table 2-29 depicts normal and extreme values for the following parameters: temperature, rain, sleet and snow, fog, relative humidity, dew point and wind direction and speed (Reference 1).
2.3.2.2 Potential Influence of the Plant and Its Facilities on Local Meteorology The operation of closed-cycle mechanical draft cooling towers for waste heat dissipation in the condenser cooling water system at Catawba Nuclear Station is viewed from the standpoint of condensate plume effects and tower drift effects. Figure 2-14 and Figure 2-15 depict frequencies of condensate plumes by length and direction from the plant for summer and winter, respectively. Percentage occurrence is cumulative, is representative of mechanical draft towers and is without regard to height of the plume.
Frequencies are derived from empirical data on plume parameters for a mechanical draft cooling tower at the Duke Power Cliffside Plant (August, 1972-July, 1973), a 600 MW(e) station located 40 miles northwest of Catawba Nuclear Station.
Plume frequencies are derived from observations made at 0800 LST of plume rise, length and direction of drift to eight compass points. Rise characteristics are assessed by reference to a 500 foot stack adjacent to the cooling towers. Length and direction are estimated from an area map provided with range markers. Three helicopter flights were made at observation times during the period of record to ascertain the adequacy of ground based observations and assess
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 3 other factors relating to plume behavior; e.g., effects of elevated and ground based inversions on plume dissipation. Application of measured plume parameters at Cliffside Steam Station to represent plume behavior for mechanical draft cooling towers at Catawba involves: the extrapolation of observed lengths at Cliffside to account for a different heat load and redistribution directionwise of length by direction frequencies to coincide with observed 40 meter level wind directions at the Catawba site. This redistribution maintains the percentage breakdown of plume lengths within each sector as reported at Cliffside but changes the wind direction distribution to that at the Catawba site. This direction redistribution involves only minor changes in direction frequencies. Tower heat loads are approximately 820 MW(t) at Cliffside as versus 4630 MW(t) at Catawba. Heat load is adjusted assuming the same proportions of sensible and latent heat are released at each plant. A factor then of five is applied to the evaporation rate at Cliffside to approximate evaporation from the Catawba towers. In translating length frequencies by direction as observed at 0800 LST to frequencies representing occurrences based on total time or all hours of the day, persistence is assumed for 24 hours2.777778e-4 days <br />0.00667 hours <br />3.968254e-5 weeks <br />9.132e-6 months <br /> following each 0800 LST observation. This leads to an overstatement of the frequency of extended plume lengths in that early morning is a favored time for long plume occurrences.
Differences in evaporation rate are accounted for by extrapolation of observed plume lengths at Cliffside assuming a gaussian material distribution in the plume and a spread rate tending to maximize plume length (stability category D or F for plume heights of 500 and 1000 feet)
(Reference 10).
Fogging due to cooling tower operation is not expected to be a problem. For mechanical draft towers, based on one year of experience from the Cliffside towers, ground contact is limited to within 0.5 miles of the plant, occurring at a combined frequency of less than one percent for all temperatures and wind directions. The estimate for the extent of ground level fogging from mechanical draft cooling towers at Catawba (to 0.5 miles) is based on the observation of ground level fogging from the Cliffside towers. All cases of cooling tower plumes at ground level were reported to occur within 1000 feet of the towers. All plumes at ground level were observed to "take off" from the ground (buoyant rise) instead of dissipating from the action of atmospheric turbulence. Consideration of the differences in tower shape (circular at Catawba with crosssectional area approximately 13,500 ft2/ tower; rectangular at Cliffside with maximum cross-sectional area approximately 21,000 ft2/tower) and in heat load per tower (about 775 MW(t) at Catawba to 410 MW(t) at Cliffside) suggest some amelioration of ground level fogging at Catawba; low pressure wake effects would be lessened while plume buoyancy would be increased. A 1000 foot distance criterion has been used in design considerations with regard to the positioning of electrical equipment in the station yard. Since the nearest highway, S C 274, is approximately 1.5 miles from the cooling tower yard, no effect on ground transportation is expected.
A fog study was conducted to assess the effects of the station heat dissipation on the frequency and intensity of ground fog. Emphasis was on identifying measurable changes in fog characteristics which are potentially adverse economically or aesthetically to the communities on Lake Wylie. The study was designed to address the impact of increased surface water temperatures from plant discharges; increased atmospheric moisture from vapor due to cooling tower plume downwash, plume dispersion, or drift evaporation; and increased fog condensation nuclei from cooling tower drift.
A detailed description of the study and results can be found in the Duke Power Company Report (Reference 29) entitled "Cooling Tower Effects on Area Fog Events - Catawba Nuclear Station -
Clover, South Carolina" (April, 1988). The study did not find any discernible effects of cooling tower operation on the surrounding fog climatology. As a result, the monitoring program has been discontinued and mitigating actions were not recommended.
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 4 (09 OCT 2016)
Climatology of the Catawba Plant is influenced by the same regional weather regimes that affect the Cliffside Plant. With regard to the specifics for diffusion and background moisture considerations the following comparisons are drawn:
Mean Maximum Temperature (°F)
Mean Minimum Temperature (°F)
Mean Wind Speed (mph)
Charlotte Airport (Reference 1) 71.3 49.6 6.9 Mean Maximum Temperature (°F)
Mean Minimum Temperature (°F)
Mean Wind Speed (mph)
Greenville Airport (Reference 11) 70.5 51.5 8.2 Mean Maximum Mixing Height (m)
(Reference 8)
Mean Surface Dew Point (°F)
(Reference 12)
Cliffside Plant 1500 47 Catawba Nuclear Plant 1500 48 Tower drift effects, assessed in studies at the construction permit stage, are not significant, and therefore are not addressed in this safety report. Preoperational and postoperational terrestrial monitoring, however, is to be undertaken.
Figures 1-1 and 2-2 show the regional location in plan view to 50 mi; Figures 2-67 to 2-77 provide information on regional topography. Figure 2-7 is a detailed plan view, as modified by the plant to 5 mi. Figure 2-16 depicts maximum elevation versus distance to 5 mi for each of the sixteen 22.5° sectors.
2.3.2.3 Local Meteorological Conditions for Design and Operating Bases Table 2-30 shows all meteorological values used in structural design. Table 2-32 and 2-33 show meteorological data used in sizing of the ultimate heat sink. The period of record from which this data is selected is January, 1951-December, 1972. The source of this data is Douglas Municipal Airport in Charlotte, North Carolina. The worst cooling condition is defined as the period in which the equilibrium temperature is the highest. Procedures used to calculate daily average equilibrium temperatures are from Ryan and Harleman (Reference 9). The highest evaporation is also determined from Ryan and Harleman (Reference 9). Ultimate heat sink performance is discussed in Section 9.2. At the time of the original analysis, the period January 1951 through December 1972 was the extent of the available data base. Analysis of the most recent 30 year period of record (1951-1980) indicates that the worst cooling and worst evaporation periods identified in the original analysis remain the design basis periods. The method of analysis used with data obtained for the period 1973-1980 is the same method as that performed on data obtained for the period 1951-1972. This method of analysis is discussed in FSAR Section 9.2.5.3.2.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 5 2.3.3 Onsite Meteorological Measurements Program HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.3.3.1 December 17, 1975 through December 16, 1977 Pre-operational onsite meteorological measurements were made at the 10m and 40m levels from December 17, 1975 to December 16, 1977. Measurements included wind direction and speed, horizontal wind direction fluctuation, temperature and vertical temperature gradient, dew point and rainfall. Table 2-44 summarizes wind data at the 40m height for this period. A comparison of 10m wind measurements at plant grade and from the 10m tower in early studies did not reveal the occurrence of any systematic gravity flow onsite.
2.3.3.2 January 6, 1985 through June 13, 1996 The operational measurement program began on January 6, 1985. A new set of instrumentation was installed on a microwave tower and a nearby 10m tower. Measurements included wind speed, wind direction, and temperature at both the 10m and 40m levels. Vertical temperature gradient (delta-T) and 10m dewpoint temperature were also measured. Precipitation was measured from a rain gauge near the base of the tower. The upper level wind system was mounted on a mast nominally 12 feet above the microwave tower: the rain gauge was located approximately 50 feet north of the low level tower.
Vertical temperature gradient can be used to determine atmospheric stability class. With a vertical separation distance between upper and lower temperature sensors of approximately 30m, the following Delta-T ranges should be used for the stability classes.
(1985 - 1996 June)
Stability Class Delta-T Range (C)
A dT -
0.57 B
-0.57 < dT -
0.51 C
-0.51 < dT -
0.45 D
-0.45 < dT -
0.15 E
-0.15 < dT 0.45 F
0.45 < dT 1.2 G
1.2 < dT 2.3.3.3 June 13, 1996 through Present On June 13, 1996 a taller, replacement meteorological tower became operational onsite. It is located approximately 1300 feet southwest of the CNS Unit 1 Vent, on the same hilltop as the previous meteorological/microwave tower. Figure 2-17 depicts the relative elevations of the meteorological instruments at both upper and lower heights on the new meteorological tower,
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 6 (09 OCT 2016) with respect to tower grade, station grade, and unit vent release elevation. The new tower is instrumented with wind speed, direction and temperature sensors at approximately 10m and 60m heights above ground level. The dew point temperature sensor is located at the 10m level.
The rain gauge remained at the previous location and is near the base of the 60m tall meteorological tower.
Above Grade FEET METER Grade = 629.67 ft. elev.
0.0 0.0 Lower Aspirator = 660.25 ft. elev.
30.58 9.32 Lower Wind Speed & Direction = 663.5 ft. elev.
33.83 10.31 Upper Aspirator = 827.25 ft. elev.
197.58 60.22 Upper Wind Speed & Direction = 830.5 ft. elev.
200.83 61.21 Delta-T values are calculated based on the temperature measurements between the upper and lower levels. The vertical separation distance between the temperature sensors is 50.9m. This is a change from the previous tower's delta-T separation distance, therefore, the delta-T ranges for determining stability class are different. The new ranges are listed below for the new separation distance of 50.9m (i.e. 167 feet).
Stability Class Delta-T Range (C)
A dT -0.97 B
-0.97 < dT -0.87 C
-0.87 < dT -0.76 D
-0.76 < dT -0.25 E
-0.25 < dT 0.76 F
0.76 < dT 2.04 G
2.04 < dT Instrument accuracies and schedules for calibration and maintenance are listed below. The ambient and T sensors were specified to have a minimum accuracy of 0.1C which is considered adequate for the intended monitoring function.
Operational measurements consist of near real-time digital outputs for all parameters in addition to the analog system. The digital outputs are averaged over 15-minute periods for use in a near real-time puff-advection model to calculate offsite dose during potential radiological emergencies. The Operator Aid Computer computes 15 minute running averages from a sampling interval of one minute, with the exception of upper and lower wind direction. They have a sampling interval of five seconds, and the lower wind direction is also used for a 15 minute running average calculation of sigma theta. Precipitation is sampled each minute and a value accumulated during the current quarter hour and the previous quarter hour.
Instrument accuracies for satisfying regulatory requirements are:
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 7
- 1. Wind Direction
- a. Time-averaged digital accuracy +/- 5° of azimuth
- b. Time-averaged analog accuracy +/- 7.5° of azimuth, or (1.5 x digital accuracy)
- c. Starting threshold 0.3 m/sec (0.7 mph)at 10° initial deflection, or less than 0.4 m/s (1.0 mph)
- d. Damping ratio 0.4 at 10° initial deflection
- e. Distance constant 1.1 m
- 2. Wind Speed
- a. Time-averaged digital accuracy +/- 0.5 mph for range of 0 to 90 mph
- b. Time averaged analog accuracy +/- 0.75 mph for range of 0 to 90 mph
- c. Starting threshold 0.6 mph, or less than 1 mph (0.4 m/s)
- d. Distance constant 1.5 m, or less than 2m
- 3. Temperature
- a. Operating Range: -20.0C to 40.0C
- b. Time-averaged digital accuracy +/- 0.5°C
- c. Time-averaged analog accuracy +/- 0.5°C
- 4. Delta-Temperature
- a. Operating Range: -4.00°C to 8.00C
- b. Time-averaged digital accuracy +/- 0.15°C
- c. Time-averaged analog accuracy +/- 0.15°C
- 5. Dew Point
- a. Required Range: -30C to 30C
- b. Time-averaged digital accuracy 1.5C
- c. Time-averaged analog accuracy 1.5C
- 6. Precipitation
- a. Digital accuracy +/- 0.1 inches at 3 inches /hr rate
- b. Analog accuracy +/- 0.1 inches of total accumulation at 3 inches /hr rate
- c. Resolution 0.01 inches Schedule for Calibration and Maintenance of Meteorological Instruments Weekly The following field checks are to be performed each week before the digital media is replaced and the recorders are serviced. These checks ensure proper maintenance of the system.
- 1. Wind Direction
- a. Recorder time accuracy
- b. Channel zero and full scale
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 8 (09 OCT 2016)
- 2. Wind Speed
- a. Recorder time accuracy
- b. Channel zero and full scale
- 3. Temperature, Delta-temperature, Dew point
- a. Recorder time accuracy
- b. Channel zero and span
- 4. Rainfall
- a. Recorder time accuracy
- b. Channel zero and full scale Semiannually The following checks are to be performed to ensure semiannual calibration of the instruments:
- 1. Temperature and Delta-Temperature
- a. Electronic simulation to processor (over total range of temperature)
- b. Processor calibrated to NIST-Traceable RTD sensors
- c. Field loop accuracy checks performed
- 2. Dew Point
- a. Dew point control unit is calibrated to NIST Traceable Dewpoint Sensor.
- b. Field loop accuracy checks performed
- 3. Wind Speed
- a. Sensor is laboratory certified
- b. Field loop accuracy checks performed (includes calibration of processor)
- 4. Wind Direction
- a. Sensor is laboratory certified
- b. Field loop accuracy checks performed
- c. True direction alignment
- 5. Precipitation
- a. Electronic simulation to processor over total range of rainfall
- b. Volumetric check of channel
- c. Field loop accuracy checks performed Other Structural inspection of the guyed tower is normally made every 3 years, per industry practice.
2.3.4 Short-Term (Accident) Diffusion Estimates HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 9 2.3.4.1 Objectives Conservative and realistic estimates of atmospheric dilution factors at the exclusion area boundary and at the outer boundary of the low population zone are provided in this section for appropriate time periods to 30 days after an accident. Data collected onsite from December 17, 1975 through December 16, 1977 provide the basis for the dilution factor, X/Q (sec/m3) estimates for an inadvertent release of radioactive material. These two years of on-site meteorological data are shown to be a representative data base for X/Q estimates with respect to long-term conditions.
2.3.4.2 Calculations Hourly dilution factors are developed for each 22.5° sector by computation from the widely accepted gaussian equation (Reference 13)
)
/
CA
(
u 1
Q
/
X z
y
where X/Q =
normalized concentration at plume centerline (sec/m3) u =
mean wind speed through the vertical extend of the plume (m/sec) y
=
crosswind concentration distribution standard deviation (m) z
=
vertical concentration distribution standard deviation (m)
C =
containment structure shape factor = 0.5 A =
cross-sectional area of containment structure normal to the wind = 1616 m2 Crosswind and vertical standard deviations are those suggested by D. B. Turner (Reference 10).
The factor
)
CA
(
z y
is a measure of plume spread. This factor is restricted to be no greater than
)
3
(
z y
as recommended in the Regulatory Guide 1.145, "Atmospheric Dispersion Models for Potential Accident Consequence Assessments at Nuclear Power Plants."
To account for effects of plume meander, especially during low wind speed stable atmospheric conditions, calculation is also made from the following empirically devised formulation to be applied for atmospheric stability classes D, E, F, and G when low level wind speeds (10 m) are less than 6 m/sec, as recommended in Regulatory Guide 1.145.
z y
u 1
Q
/
X
Where y
= crosswind concentration distribution standard deviation during the specified conditions (m)
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 10 (09 OCT 2016) y
=
y M for distances < 800 m y
=
y y
)1 M
(
for distances > 800 m where M is determined as a function of stability class and wind speed as follows:
For D stability class M =
sec)
/
m 2
u
(
for 2
M =
sec)
/
m 2
u sec
/
m 6
(
for u
10
.3 632
For E stability class M =
sec)
/
m 2
u
(
for 3
M =
sec)
/
m 2
u sec
/
m 6
(
for u
00
.6 00
.1
For F stability class M =
sec)
/
m 2
u
(
for 4
M =
sec)
/
m 2
u sec
/
m 6
(
for u
59
.9 26
.1
For G stability class M =
sec)
/
m 2
u
(
for 6
M =
sec)
/
m 2
u sec
/
m 6
(
for u
6.
18 63
.1
The effective hourly dilution factor for hours qualifying for calculation from both Eqn. 2.3.4-1 and Eqn. 2.3.4-2 is taken to be the lesser value from these two equations.
To provide the necessary wind and stability information, a joint stability-wind distribution is generated which displays the joint frequencies of wind direction and speed by atmospheric stability type as they were observed onsite at the 10 m level (see Table 2-35). The joint data recovery for this two-year period of record is 92.0 percent.
Stability categories are determined by vertical temperature gradient according to the following schedule:
Stability Class Vertical Temperature Gradient G
greater than +2.1°F in 100 ft F
+0.9 to +2.1°F in 100 ft E
-0.2 to +0.8°F in 100 ft D
-0.8 to -0.3°F in 100 ft B-C
-1.0 to -0.9°F in 100 ft A
less than -1.0°F in 100 ft Selection of hourly dilution factors have traditionally involved a probability deemed appropriate for the population as a whole, commonly accepted at 5 percent and 50 percent for conservative
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 11 and realistic estimates respectively. In order to preclude any individual (one sector) from receiving a disproportionate share of the 5 percent or 50 percent probabilities, a level of 1/16 of the probability accepted for population determination is taken as appropriate for sector probabilities. That is, the corresponding probabilities for sector assessment would be 1/16 of 5 percent and 1/16 of 50 percent.
The following procedure is used to ensure that this is the case. An effective probability is calculated for each sector which in effect establishes a probability of 1/16 of the corresponding probabilities accepted for populations (5 percent and 50 percent):
S
)
n
/
N
(
P Pe where Pe
= effective probability for the sector P
= population related probability N
= total number of hours of valid wind and stability data n
= number of hours of wind flow into the sector S
= total number of sectors The cumulative frequency distribution for hourly dilution factors in each sector results from summation of percentage values from the joint stability-wind distribution in decreasing order of normalized concentration for selected wind speed class intervals and stability categories. A dilution factor at the effective probability level (Pe) is determined for each sector for each population related probability (5 percent and 50 percent); percentage values in the joint stability-wind distribution are normalized to 100 percent. The highest of these sector dilution factors in each case is used to represent the respective population related probability levels (5 percent and 50 percent).
Hourly dilution factors selected in this way ensure that probabilities related to an individual are equal to or less than 1/16 of the population related probabilities.
Table 2-36 displays hourly dilution factors at the 5 percent and 50 percent probability levels for each 22.5° sector at the exclusion area boundary (EAB). The minimum distance for all direction sectors is taken as 2400 ft (see Figure 2-4). The highest dilution factor at the 5 percent probability level is 5.5 x 10-4 sec m-3 in the northeast sector; the highest dilution factor at the 50 percent probability level is 1.3 x 10-4 sec m-3, also in the northeast sector. Figure 2-18 displays the distribution in the northeast sector at the EAB for all probability levels. For all other sectors, these values represent probabilities less than 5 percent and 50 percent respectively.
Table 2-37 displays hourly dilution factors at the 5 percent and 50 percent probability levels for selected 22.5° sectors at the low population zone boundary (LPZB). This distance is taken in all cases as 20,000 ft (see Figure 2-8). Calculation at the LPZB is only made for sectors with high dilution factors at the EAB. The highest dilution factor at the 5 percent probability level is 3.0 x 10-5 sec m-3 in the northeast sector; that at the 50 percent probability level is 6.6 x 10-6, also in the northeast sector. Figure 2-19 displays the distribution in the northeast sector at the LPZB for all probability levels. For all other sectors these values represent probabilities less than 5 percent and 50 percent respectively.
Estimates of diffusion for longer time periods up to 30 days are developed from dilution factors for 0-8 hours, 8-24 hours, 1-4 days, and 4-30 days following an accident. These dilution factors result from a gaussian diffusion model which stores and accumulates successive hourly X/Q
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 12 (09 OCT 2016) values at angular intervals of five degrees at the low population zone boundary. Successive hourly values are calculated to crosswind distances of +/- 20 degrees from observed wind directions. Points beyond +/- 20 degrees for any one hour are assumed at zero concentration for that hour. Computation by hour then is by Eqn. 2.3.4-1 or Eqn. 2.3.4-2 according to all provisions outlined above for the calculation of hourly dilution factors. In the case of Eqn. 2.3.4-1 an exponential factor is applied of the form:
Exp
CA Y
2 1
2 y
2 where y = crosswind distance from plume centerline (m) and the building wake parameters, C and A, are entered as suggested by Davidson (Reference 13). In the case of Eqn. 2.3.4-2 an exponential factor is applied of the form:
Exp
2y 2
Y 2
1 All calm hours are included in the averages by assuming a wind direction as indicated and a wind speed of 0.45 m/sec. This minimum value derives from precedent in other licensing reviews. It is also taken as acceptable, along with use of indicated wind directions, in view of the low incidence of wind speeds less than 0.45 m/sec.
The hourly dilution factors are then combined to form cumulative frequency distributions of X/Q for the required averaging times; that is, 8 hours9.259259e-5 days <br />0.00222 hours <br />1.322751e-5 weeks <br />3.044e-6 months <br />, 16 hours1.851852e-4 days <br />0.00444 hours <br />2.645503e-5 weeks <br />6.088e-6 months <br />, 72 hours8.333333e-4 days <br />0.02 hours <br />1.190476e-4 weeks <br />2.7396e-5 months <br /> (3 days), and 624 hours0.00722 days <br />0.173 hours <br />0.00103 weeks <br />2.37432e-4 months <br /> (26 days). Successive averaging times overlap except for the first H hours of one average and the final H hours of the next. For example, the midnight to 8 a.m. average and the overlapping 2 a.m. to 10 a.m. average are considered independent members of the 8-hour average frequency distribution with H equal to 2 hours2.314815e-5 days <br />5.555556e-4 hours <br />3.306878e-6 weeks <br />7.61e-7 months <br />. The value of H can increase with increasing averaging time without significantly altering the resultant frequency distributions. Values of H of 2 hours2.314815e-5 days <br />5.555556e-4 hours <br />3.306878e-6 weeks <br />7.61e-7 months <br />, 2 hours2.314815e-5 days <br />5.555556e-4 hours <br />3.306878e-6 weeks <br />7.61e-7 months <br />, 6 hours6.944444e-5 days <br />0.00167 hours <br />9.920635e-6 weeks <br />2.283e-6 months <br />, and 24 hours2.777778e-4 days <br />0.00667 hours <br />3.968254e-5 weeks <br />9.132e-6 months <br /> are selected for averaging times of 8 hours9.259259e-5 days <br />0.00222 hours <br />1.322751e-5 weeks <br />3.044e-6 months <br />, 16 hours1.851852e-4 days <br />0.00444 hours <br />2.645503e-5 weeks <br />6.088e-6 months <br />, 72 hours8.333333e-4 days <br />0.02 hours <br />1.190476e-4 weeks <br />2.7396e-5 months <br />, and 624 hours0.00722 days <br />0.173 hours <br />0.00103 weeks <br />2.37432e-4 months <br /> respectively.
Dilution factor selection here is aimed at probability levels for the individual (one sector) at or below the corresponding population related probabilities of 5 percent and 50 percent for conservative and realistic estimates respectively. The following procedure is taken to approximate these levels.
Cumulative frequency distributions for the averaging times above are developed without regard to wind direction. In other words at the end of a given "window" calculation, the highest concentration is stored for that window period regardless of its direction. It is assumed that these distributions represent the respective sector distributions giving the highest dilution factor at the specified probability level. It is also assumed that these distributions occur in the highest frequency direction sector, for the purpose of determining an effective probability level (Pe).
Accordingly, for each distribution the highest 5 percent and 50 percent dilution factors are taken to be represented by effective probabilities of 2.3 percent and 23 percent respectively. This assumes a direction frequency of 13.92 percent which is the highest frequency for low level winds (see Table 2-35). This approximation is taken as acceptable since hourly dilution factors show similarity in frequency distribution by sector and the highest value in each probability category occurs in the highest direction frequency sector (see Table 2-36).
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 13 Table 2-38 through Table 2-44 display the cumulative frequency distributions at integer percentile levels for averaging times of 8 hours9.259259e-5 days <br />0.00222 hours <br />1.322751e-5 weeks <br />3.044e-6 months <br />, 16 hours1.851852e-4 days <br />0.00444 hours <br />2.645503e-5 weeks <br />6.088e-6 months <br />, 72 hours8.333333e-4 days <br />0.02 hours <br />1.190476e-4 weeks <br />2.7396e-5 months <br />, and 624 hours0.00722 days <br />0.173 hours <br />0.00103 weeks <br />2.37432e-4 months <br /> at the LPZ.
These distributions are combined to represent conditions for periods of 8 hours9.259259e-5 days <br />0.00222 hours <br />1.322751e-5 weeks <br />3.044e-6 months <br />, 24 hours2.777778e-4 days <br />0.00667 hours <br />3.968254e-5 weeks <br />9.132e-6 months <br />, 4 days and 30 days following an accidental release. For the 8 hour9.259259e-5 days <br />0.00222 hours <br />1.322751e-5 weeks <br />3.044e-6 months <br /> averaging time, the worst, 5 percent, and 50 percent dilution factors are 3.8 x 10-5, 1.8 x 10-5, and 6.2 x 10-6 sec m-3 respectively. For the 16 hour1.851852e-4 days <br />0.00444 hours <br />2.645503e-5 weeks <br />6.088e-6 months <br /> averaging time, the worst, 5 percent, and 50 percent dilution factors are 2.1 x 10-5, 1.2 x 10-5, and 5.4 x 10-6 sec m-3 respectively. For the 72 hour8.333333e-4 days <br />0.02 hours <br />1.190476e-4 weeks <br />2.7396e-5 months <br /> averaging time, the worst, 5 percent, and 50 percent dilution factors are 5.7 x 10-6, 4.3 x 10-6, and 2.5 x 10-6 sec m-3 respectively. For the 624 hour0.00722 days <br />0.173 hours <br />0.00103 weeks <br />2.37432e-4 months <br /> averaging time, the worst, 5 percent, and 50 percent dilution factors are 1.2 x 10-6, 1.2 x 10-6, and 9.7 x 10-7 sec m-3 respectively. Linear interpolation between integer level values was used in calculating a 2.3 percent effective probability value; the 23 percent effective probability value was read directly from the table.
Effluent control systems are evaluated for design basis accidents with respect to dose at the exclusion area boundary and the low population zone boundary. Two hour dose estimates are based on the highest effective 95 percentile hourly X/Q at the exclusion area boundary (5.5 x 10-4 (sec m-3) in the northeast sector), see Table 2-47. Dose estimates to 30 days at the low population zone boundary are accumulated end-to-end from separate exposure periods of 8 hours9.259259e-5 days <br />0.00222 hours <br />1.322751e-5 weeks <br />3.044e-6 months <br />, 16 hours1.851852e-4 days <br />0.00444 hours <br />2.645503e-5 weeks <br />6.088e-6 months <br />, 3 days and 26 days. Average dispersion factors for each time period represent the respective highest effective 95 percentile X/Q (assumed to occur in the northeast sector in all cases) with values of 1.8 x 10-5, 1.2 x 10-5, 4.3 x 10-6 and 1.2 x 10-6 (sec m-3) for averaging times of 8 (0-8) hours, 16 (8-24) hours, 3 (1-4) days and 26 (4-30) days, respectively; see Table 2-47.
2.3.5 Long Term (Routine) Diffusion Estimates HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.3.5.1 Objectives Realistic estimates of annual average atmospheric dilution factors (X/Q) and deposition factors (D/Q) for unrestricted areas are provided in this section. Four separate analyses comprise this section:
- 1. A spatial distribution of annual average X/Q values is generated assuming advection and diffusion are the primary plume dispersion and transport processes.
- 2. A value for Man -X/Q is calculated as a population weighted annual average value within a 50 mile radius of the site (2000 population estimate; see Table 2-6 and Table 2-13).
- 3. Annual average X/Q values for computing radioiodine dosage through inhalation are produced considering the role of dry deposition in plume depletion in addition to advection and diffusion of the plume.
- 4. Annual average D/Q values for computing radioiodine dosage through milk and leafy vegetable pathways are produced considering the role of dry deposition in plume depletion in addition to advection and diffusion of the plume.
Onsite data from December 17, 1975 through December 16, 1977 provide the basis for the diffusion and deposition estimates. These estimates are assumed to be representative of annual average X/Q and D/Q values with respect to long-term conditions.
Changes in numerical values under Revision 1 result from the correction of a coefficient pertaining to stable plume rise.
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 14 (09 OCT 2016) 2.3.5.2 Calculations Average dilution factors, X/Q (sec m-3), are computed covering the stated period of record for angular intervals of five degrees at ten distances to 50 miles utilizing a computer program to store and accumulate successive hourly values.
The model for annual average X/Q is identical to that described for the calculation of dilution factors for intermediate averaging times up to 30 days with the following exceptions:
- 1. Calculation is solely from Eqn. 2.3.4-1.
- 2. Releases from the 38 meter vent stacks are considered partially elevated and partially ground-level releases. The fraction of the plume material which remains elevated depends on the ratio of exit velocity to wind speed at release height. This fraction has been calculated from equations 7 and 8 of NRC Regulatory Guide 1.111, Revision 1 (Reference 14).
5.1 5.1 W
1 for U
W 58
.1 58
.2 Fg o
o
and 0.5 U
W 5.1 for U
W 06
.0 3.0 Fg o
o
where Fg
= fraction of the time the release is considered to occur at the ground Wo
= exit velocity (m/sec)
U
= wind speed from the 40 m sensor (m/sec)
Fe
= 1 - Fg where Fe
= fraction of the time the release is considered to be elevated.
We selected this conditionally elevated treatment of effluent release as the most appropriate for Catawba against the basic criteria of obtaining most realistic estimates without substantial underestimation. The approach incorporated results from a gaseous tracer study at Millstone Nuclear Power Station (Reference 15). There is, admittedly, some legitimate question with respect to extrapolation rationale where other building geometries and release conditions may be concerned.
Defining the appropriate release conditions involves analysis of relevant qualitative and quantitative research results. Notable work in this area regarding prototype tests are the recent tests at Millstone Nuclear Power Station referenced above and the ongoing research by the Environmental Research Laboratories, Air Resources Laboratories at Idaho Falls, Idaho.
Physical modeling studies at the Colorado State University also addressing this general problem are underway.
We examined the Millstone Tests, being the most applicable to Catawba with regard to quantitative prototype tests of somewhat similar release conditions and building geometry. First, they are viewed by comparison to Catawba as to probable tendency for building wake entrainment. On the basis of building geometry and release conditions (Figure 1-9 and Figure 1-15 and Reference 15), the configuration at Catawba appears more favorable than at Millstone; that is, considering general characteristics of flow separation and wake behavior for the two
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 15 geometries (Reference 13) and considering the relative position and nature of the release points to expected wake boundaries.
Dominant factors are:
- 1. Building geometry. It is sharp-edged at Millstone where the leading edge for the test series is significantly upwind from the primary release point. It is domed at Catawba with wall parapet approximately 65 feet laterally from and 17 feet below the dome apex, where the release point is 12 feet below and approximately 70 feet laterally from the dome apex.
- 2. Release conditions. Effluent velocity is about 10 m sec-1 at Millstone from stack of cross-sectional area about 3.5 m2. Effluent velocity is about 22 m sec-1 at Catawba from a stack of cross-sectional area about 3.5 m2.
We also examined the Millstone tests against qualitative evidence from other tests to further judge the suitability of the Millstone results for use at Catawba. The Air Resources Laboratory at Idaho Falls, Idaho recently investigated building wake effects at the Rancho Seco Nuclear Power Station (Reference 16).
The experimental procedure involved release of two quantitative tracers and one visual tracer at various points on the reactor building complex. Release conditions for some runs (oil fog released at the wall parapet) in this series parallels the Catawba physical arrangement. The reactor building itself is a domed structure similar to Catawba and release of oil fog at the building parapet essentially reproduces release conditions at Catawba in that the release point is below the apex of the dome in a nearly identical sense. Releases at Rancho Seco, however, involve little or no vertical momentum or buoyancy. Photographic documentation of oil fog releases in this configuration reveals a very substantial tendency for plume material to loft above wake turbulence. Summary statements by the investigators also indicate that the bulk oil fog material released at the wall parapet remains aloft both for stable and unstable conditions. This behavior must be regarded as an approximation to observed plume characteristics at Millstone. Accordingly, the quantitative treatment derived from the Millstone experiment satisfies the condition of best methodology for most realistic estimates with the provision that significant underestimation is unlikely.
Plume height for elevated releases is calculated from Sagendorff (Reference 17). Effective stack height is determined from pr s
h h
H
H = effective stack height (m) hs = physical stack height (m) hpr = plume rise (m)
Plume rise is calculated using formulas from Briggs (Reference 18).
The station is assumed to have a cold plume, so the heat emission rate is zero, and the plume rise is calculated from the momentum equations. For neutral or unstable conditions, D
D X
U W
44
.1 h
3
/
3
/
o pr 1
2
(a) where
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 16 (09 OCT 2016)
X
= downwind distance (m)
D
= internal stack diameter (m).
When the exit velocity is less than 1.5 times the wind speed, a correction for downwash is subtracted (Reference 19):
D U
W 5.1 3
C o
where C is the value to be subtracted, and the other terms are defined above. The result is compared with D
U W
3 h
o pr
(b) and the more conservative value is used.
For stable conditions, the result from (a) or (b) is compared with the results from the following two equations:
4
/
m pr 1
S F
4 h
6
/
1 3
/
m pr S
U F
5.1 h
1
and the smallest value of hpr is used. Above, Fm is the momentum flux parameter and S is a stability parameter where:
m F =
2 D
W 2
2 o
S =
T g
and
vertical gradient of potential temperature (°K/m) g
9.8 m/sec2 and T = Temperature (°K)
S =
)
(sec 10 45
.2 and 10 75
.1, 10 7.8 2
3 3
4
for E, F, and G stabilities, respectively.
Plume rise is computed from the exit velocity (22.4 m sec-1) and stack diameter (2.1 m),
employing the wind speed from the 40 meter sensor. The effect of terrain on effective plume height is included according to Egan (Reference 20). If all heights are referenced to plant grade, H is the effective plume height without terrain correction, and ht is the height of the terrain feature; then the corrected plume height is 2
/
h H
t
above local terrain. An exception
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 17 noted is that plume height is constrained to remain between H and H/2 above local terrain.
The ht values represent the highest terrain in the vicinity of the receptor within the 22.5° sector after Sagendorf (Reference 17); ht values are taken to be equal to or greater than zero according to NRC Regulatory Guide 1.111, Revision 1 (Reference 14).
The equation employed for each hourly X/Q calculation for the ground release portion is
/
CA 2
y exp CA u
F
)
Q
/
X
(
2 y
2 1
z y
1 g
g which is Eqn. 2.3.4-1 with the crosswind exponential.
The equation employed for the elevated portion is
2 z
2 2
y 2
2 z
y 2
e e
2 H
2 y
exp u
F
)
Q
/
X
(
which is Eqn. 2.3.4-1 with the crosswind and elevated release exponentials, where CA=O.
1 u and 2
u are the low-level and high-level average wind speeds, respectively (m sec-1). A minimum value of 0.45 m/sec is assumed. The minimum value derives from precedent in other licensing reviews. It is also taken as acceptable, along with use of indicated wind directions, in view of the low incidence of wind speeds less than 0.45 m/sec. High level winds (40 m) seem appropriate for elevated material in that momentum plume rise tends to balance the minor effect of disparity in absolute elevation.
y1 and y2 are the lateral distances of the receptor from the wind direction vectors 1
u and 2
u,
respectively (m).
H is the plume height considering all corrections as discussed above (m).
The factor
)
CA
(
z y
is a measure of plume spread. This factor is again restricted to be no greater than
)
3
(
z y
- 3. The (X/Q)g and (X/Q)e values, calculated for annual average X/Q input to radioiodine dosage, are modified to account for plume depletion by dry deposition of elemental radioiodine. Factors employed are from NRC Regulatory Guide 1.111, Revision 1, (Reference 14) which discusses the bases for their development.
- 4. Output is summarized in terms of sector averages from the 5° grid point values. (The grid point values are averages derived from hourly values over the period of record. These grid point values in turn are averaged within each 22.5° sector to give sector average values.)
NRC Regulatory Guide 1.111, Revision 1, (Reference 14), suggests that long-term X/Q and D/Q values be adjusted to account for variations in plume trajectory over time scales on the order of one day, which would otherwise not be considered by the straight-line trajectory models. The adjustment factor for this station is taken to be effectively 1.0; that is, no such variations are significant. This fact is demonstrated in the following analysis:
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 18 (09 OCT 2016)
Factors which would cause an adjustment factor greater than 1.0 are a) systematic flow reversals, b) stagnant pooling of air, c) systematic curved trajectories such as terrain-induced channeling, and d) randomly curved trajectories under some conditions.
Flow reversals would yield higher doses because a repeated passage of the effluent would effect longer dwell times and would cause higher air concentrations by introducing contaminated background air. Nocturnal downslope flows at the plant site could be a mechanism for such recirculation. Inspection of Table 2-34 reveals that no significant bias in wind direction during stable conditions is evident to support appreciable occurrence of such a flow. Downslope wind possibilities are assessed for near vent levels as most releases (about 95%) are elevated in the calculation of annual average X/Q. Drainage activity affecting these levels should be reflected in data from the 10 m wind system.
Stagnation of contaminated air would cause higher doses since the model assumes contamination only in the downwind direction. Stagnation at the plant would result from winds at stack release height which are persistently low. The frequency of wind speeds less than 1.0 mph without regard to persistence for all stability conditions is 0.17 percent. This low frequency occurrence of low wind speeds at stack height should not significantly affect annual average dose from stagnation. Wind speeds equal to or greater than 1.0 mph at the 40 m level are assumed associated with parcel trajectories which on average have a net away-from-the-plant component (for a period of one hour) that is constant with downwind distance. The time mean mass distribution in the along-wind direction, then, is constant and equal to u
/
Q (Ci/m). For low wind speeds greater than 1.0 mph, it is certainly recognized that individual parcels would be expected to follow an exaggerated meandering path. Even so, the motion field described precludes a pooling effect in time averages. The position is of course a matter of judgement and subject to equally valid alternative opinions.
Systematic curved trajectories would effect higher doses in some direction if flow, induced by terrain or any other source, exposes a receptor more frequently or to higher concentrations than the straight-line trajectory assumption. Channeling of winds by the valley walls at Catawba, or pronounced drainage winds at night could cause such an underestimation by the model. It is evident that the gentle terrain variations within the valley do not channel the winds. Also, the absence of significant drainage is addressed above.
With respect to curvature which is random by direction, when direction frequency is inhomogeneous and dispersion conditions are homogeneous from one sector to another, the effect of course is to reduce the annual average X/Q or D/Q value in the high frequency sector.
Only when there exist severe differences in direction frequency, or a positive correlation between poor dispersion conditions and high direction frequency is evident, are noticeable changes likely to occur in the long-term X/Q or D/Q fields. Inspection of the high frequency southwest wind directions indicates neither the relative frequency of wind direction nor the relative proportion of poor dispersion conditions from one sector to another is unduly biased in the sense discussed above (see Table 2-34 and Table 2-43). In the high frequency northerly wind directions, there is no significant bias.
In summary, there exists no apparent cause for systematic flow reversals, systematic trajectory curvature or stagnation of contaminated air; and the conditions for which random curvature is a problem do not exist at the site.
Table 2-41 displays annual average X/Q values by sector at 10 distances to 50 miles. The highest value at the EAB is 2.5 x 10-7 sec m-3 in the north northeast sector; the annual average X/Q weighted by population within 50 miles is 5.0 x 10-8 sec m-3 (Table 2-48). Table 2-42 displays annual average X/Q values, reflecting the effect of plume depletion of elemental radioiodine by sector at 10 distances to 50 miles. The highest value at a residence is 3.5 x 10-7
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 19 sec m-3 at 0.7 mi north-northeast of the plant interpolated average values; the highest value at the EAB is 2.4 x 10-7 sec m-3 north-north east of the plant (Table 2-48).
Average deposition factors, D/Q (m-2) are computed, covering the stated period of record, utilizing a second computer program to combine wind and stability frequencies with relative deposition rates (m-1) to yield sector average relative deposition/area, D/Q (m-2). The resulting estimates of sector average D/Q are assumed to represent annual conditions.
For each sector, normalized wind and stability frequencies are extracted from Table 2-43 (40 m wind system) to provide weighting factors for relative deposition rates (m-1) obtained from NRC Regulatory Guide 1.111, Revision 1 (Reference 14), in which the methodology used to generate these values is discussed. In the weighting process, a fractional breakdown of elevated and ground-level plume contributions is assumed with calculation by the method employed for average X/Q values; plume rise also is calculated as for input to X/Q values. Where, in this case, the fractional breakdown and plume rise applies to a wind speed category, the midpoint value in the wind speed class is selected for input to these calculations. The effective plume height is adjusted for effect of terrain according to Egan (Reference 20) as already described for application to average X/Q values. Sector average relative deposition rates (m-1) so determined are calculated for each sector at 10 downwind distances. Relative deposition/area, D/Q (m-2) then is calculated by multiplying each average relative deposition rate (m-1) by the respective sector direction frequency and dividing by the appropriate sector width. In effect this takes the crosswind integrated average relative deposition rate (m-1) and apportions the release rate based on direction frequency, with account for lateral spread by sector width, to generate an average relative deposition factor D/Q (m-2). Adjustment considerations with respect to the straight-line trajectory aspect of this model are addressed above.
Table 2-44 displays average D/Q values by sector at 10 distances to 50 miles. The highest values through the milk pathway are for a cow at 2100 m northwest of the plant and a goat at 2200 m northwest of the plant (Table 2-48). They are 5.7 x 10-10 m-2 and 5.4 x 10-10 m-2 respectively. The highest value through the leafy vegetable pathway is for a garden 2400 m south-southwest of the plant at 7.7 x 10-10 m-2 (Table 2-48). Again, values at these receptors are interpolated from the sector average results at the 10 radial receptors (Table 2-44).
2.3.6 Other Diffusion Considerations HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED Annual average X/Q values for receptors on or near plant structures are provided in this section.
Onsite data from December 17, 1975 through December 16, 1977 also provide the basis for these diffusion estimates.
By virtue of the near-field mixing characteristics, a crosswind integrated form of equations 2.3.5-1 and 2.3.5-2 is used.
Treatment with respect to fractional breakdown of plume height and plume rise is identical to that in 2.3.5. Calculation for each receptor is in terms of a sector average value utilizing a third computer program with computation from the following equations:
The higher of
2
/
1 2
2 o
2 z
2 1
2
/
g g
4 D
W u
f f
3 R
)
2
(
F
)
Q
/
X
(
1
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 20 (09 OCT 2016) or and Most parameters and variables are defined in 2.3.5. Others are considered below.
The ground release equations again place an upper limit on the effect of building turbulence to enhance spread rate of the plume. As applied to vertical diffusion above, the factor 3 is taken as appropriate for the onedimensional case (Reference 14). In-stack dilution is accounted for here with calculation analogous to building wake mixing. Effective plume height is adjusted to account for receptor height by straight subtraction of the highest building height in the vicinity of the receptor; this allows for effects of entrainment along side walls of roof protuberances, etc.
That is, if all heights are referenced to plant grade, with H the effective plume height without terrain correction and hb the highest building height near the receptor; the corrected plume height is H-hb. In assessing effective sector widths for these crosswind integrated forms, some degree of arbitrariness is necessary in view of the nature of building wake phenomena.
Consideration of an appropriate sector width is coupled with a judgement on the range of direction azimiths affecting a given receptor. For this application, sector width, (rad), is assumed at 22.5° for elevated plume portions; the width for ground-level portions is the greater of 22.5° or the angle intercepted by the reactor building as seen by an observer at the receptor.
With regard to the range of wind directions affecting a given receptor, the factor f is the frequency of a wind and stability category for one or more 22.5° sectors as presented in Table 2-43 (40 m wind system); the summation is performed over all categories. For simplicity the same number of 22.5° sectors are presumed to contribute to f for both elevated and ground release portions. The number of sectors selected for each receptor is based on consideration of the ratio of reactor building-receptor distance, R(m), to the reactor building diameter, d(m),
according to Murphy (Reference 21). This approach is taken as a reasonable approximation in conjunction with assumptions for sector width. The calculation of Fg, Fe and plume rise is made from the midpoint wind speed in each wind speed-stability category. The midpoint value is also used for 2
u in the expression for (X/Q)e. For 1
u input, the following relation applies:
P 2
1 2
1
)
h
/
h
(
u u
, where h1 is the height of 1
u and h2 is the height of 2
u (Reference CA u
f f
R
)
2
(
F
)
Q
/
X
(
1 2
/
1 2
z 1
2
/
1 g
g
4 D
W u
4 D
(
2 H
exp f
f R
)
2
(
F
)
Q
/
X
(
2
/
1 2
2 o
2 z
2 2
2 2
z 2
2
/
1 e
e
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 21 22). P has values of 0.50 and 0.25 for stable (G, F and E) and unstable (A, B-C and D) stability classes respectively. The reactor building diameter, d, is 40m from Figure 1-15.
Values of X/Q computed for various intake vents are presented in Table 2-46.
Control room /Q values associated with design basis accidents also are presented in this section. Averaging periods of 0-2 hours, 2-8 hours, 8-10 hours, 10-24 hours, 24-96 hours, and 96-720 hours following an accident are taken. Onsite meteorological data spanning five years form the basis for these diffusion estimates.
The ARCON96 computer code (Reference 30) was used to calculate all control room /Q values reported here. For all these calculations, the ground release option was used.
The results of the ARCON96 calculations are presented in Table 2-105. All values are associated with transport of fission products from one release point to one control room outside air intake. Values are presented for the time periods 0-2 hr, 2-8 hr, 8-10 hr, 10-24 hr, 24-96 hr, and 96-720 hr. The 0-2 hr control room /Q value is used for the 2 hr time span of maximum release of radioactivity to the enviromment. This 2 hr period is taken to be the time period of maximum 2 hr Total Effective Dose Equivalent (TEDE) at the Exclusion Area Boundary (EAB) as predicted in the calculations of post accident radiation doses. The 2-8 hr control room /Q value is used for the remaining intervals of the first 8 hr after the initiating event.
Catawba Nuclear Station is equipped with two control room outside air intakes. These intakes are open. Their post accident function is to stay open to allow the Control Room Ventilation System to develop and maintain an overpressure in the control room with respect to all abutting volumes. There are no automatic controls to close these outside air intakes. However, the one control room outside air intake may be closed under administrative controls. If an intake is closed these administrative controls direct the control room operators to declare a pressurized filter train of the Control Room Area Ventilation System inoperable. For this reason, the calculations of radiation doses following a design basis accident include the following assumption:
- 1) Both control room outside air intakes initially are open for a design basis accident with a single failure. Or
- 2) One control room air outside intake initially is closed for a design basis accident with no single failure. It is assumed that the control room operators open the closed intake within 10 hours1.157407e-4 days <br />0.00278 hours <br />1.653439e-5 weeks <br />3.805e-6 months <br /> after unit trip.
The airflow into the control room outside air intakes may not be balanced. The design basis value of asymmetry in the airflow split is 60/40. Composite values for control room /Q values for transport of fission products to two control room outside air intakes with an airflow imbalance is calculated below positions. If both intakes are in one wind direction window, the composite control room /Q value is calculated with
Q X /
max
2 1
2 1
2 1
2 1
/
/
min min
/
/
max Q
X Q
X q
q Q
X Q
X q
q
In the above, (/Q)1 and (/Q)2 are associated with transport of radioactivity to the control room outside air intakes of Unit 1 and Unit 2, respectively. Also, q1 and q2 are the relative airflow rates in the intakes of Unit 1 and 2. Their limiting values are q1 = 0.6 and q2 = 0.4 giving
= 0.6 max ((/Q)1,(/Q)2) + 0.4 min((/Q)1,(/Q)2)
If at most one control room outside air intake is in any wind direction window, composite control room X/Q values for transport of radioactivity to both outside air intakes are calculated with
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 22 (09 OCT 2016)
2 1
2 1
/
/
max max
/
Q X
Q X
q q
Q X
For Catawba this reduces to
= 0.6 max ((/Q)1,(/Q)2)
Both control room outside air intakes are in the same wind direction window for a release from the Refueling Water Storage Tank and the outboard steam generator doghouse. This is reflected in the design basis values for the control room /Q value associated with these release points. For all other potential release points, only one outside air intake is in any wind direction window.
The calculations of the post accident control room /Q value as described above conform to the germane regulatory positions (Reference 31). The control room /Q values associated with transport of fission products from post accident release points to one control room outside air intake are presented in Table 2-105.
2.3.7 Summary of Offsite Diffusion Estimates HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED.
Table 2-46 and Table 2-47 depict dilution factors for each type of release at appropriate receptors and percentile values.
STAR processing of Charlotte Airport data has been accomplished for the period of onsite data (1976-1977) in addition to a five-year period (1969-1973); see Table 2-49. Comparison of wind direction and speed, and of stability type forms the basis of judging the representativeness of data for the period 1976-1977, with respect to long-term conditions, as for the period 1969-1973; see Table 2-50. The 1976-1977 period is taken on balance to be reasonably representative of long-term conditions at the site. Some conservatism, with respect to accident analysis calculations, is indicated by a higher incidence of low wind speed, stable conditions.
2.3.8 References
- 1. U.S. Department of Commerce, Environmental Science Services Administration, Climate of the States: Climate of North Carolina (Climatography of the United States No. 60-31, Washington, D.C., Revised Ed. 1970), pp. 11-13.
- 2. Tattleman, Paul and Gringortem, Irving I., Estimated Glaze Ice and Wind Loads at the Earth's Surface for the Contiguous United States (Air Force Surveys in Geophysics, No.
227, Bedford, Mass., Air Force Cambridge Research Laboratories, 1973), p. 24.
- 3. U.S. Department of Commerce, Weather Bureau, Tropical Cyclones of the North Atlantic Ocean (Technical Paper No. 55, Washington, D.C., 1965), p.30.
- 4. U.S. Department of Commerce, Weather Bureau, Tornado Occurrence in the United States (Technical Paper No. 20, Washington, D.C., 1960).
- 5. Thom, H.C.S. "Tornado Probabilities," Monthly Weather Review, Oct. - Dec., 1963, pp. 730-736.
- 6. U.S. Department of Commerce, Weather Bureau, Mean Number of Thunderstorm Days in the United States (Technical Paper No. 19, Washington, D.C., 1952).
- 7. U.S. Department of Commerce, Weather Bureau, Severe Local Storm Occurrence, 1955-1967 (Technical Memorandum WBTM-FCST #12, Washington, D.C., 1969).
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.3 - 23
- 8. Holzworth, George C., Mixing Heights, Wind Speeds, and Potential for Urban Air Pollution Throughout the Contiguous United States (Office of Air Programs Publication No. AP-101, Research Triangle Park, N.C., Office of Technical Information and Publications, Office of Air Programs, Environmental Protection Agency, 1972), pp. 26-35.
- 9. Ryan P.J. and Harleman, D.R.F., An Analytical and Experimental Study of Transient Cooling Pond Behavior (Ralph M. Parsons Laboratory for Water Resources and Hydrodynamics Report No. 161, Cambridge, Mass., Department of Civil Engineering, School of Engineering, Massachusetts Institute of Technology, 1973), pp. 59, 83.
- 10. Turner, D. Bruce, Workbook of Atmospheric Dispersion Estimates (Office of Air Programs Publication No. AP-26, Research Triangle Park, N.C., Office Technical Information and Publications, Office of Air Programs, Environmental Protection Agency, Revised ed. 1970),
pp. 8-9.
- 11. U.S. Department of Commerce, Environmental Science Services Administration, Climate of the States: South Carolina (Climatography of the United States No. 60-38, Washington, D.C., Revised ed. 1970), p. 11.
- 12. U.S. Department of Commerce, Environmental Science Services Administration, Environmental Data Service, Climatic Atlas of the United States, Washington, D.C., 1968, pp. 57-58.
- 13. Slade, D.H., ed., Meteorology and Atomic Energy - 1968 (U.S. Atomic Energy Commission Report TID-24190, Springfield, Virginia, National Technical Information Service, U.S.
Department of Commerce, 1968), pp. 99, 112, and 221-232.
- 14. U.S. Nuclear Regulatory Commission, Methods for Estimating Atmospheric Transport and Dispersion of Gaseous Effluents in Routine Releases from Light-Water-Cooled Reactors (U.S. Nuclear Regulatory Commission Regulatory Guide 1.111, Revision 1, Washington, D.C., Division of Document Control, U.S. Nuclear Regulatory Commission, 1977), pp. 9-12.
- 15. Johnson, W.B., et al., Gas Tracer Study of Roof-Vent Effluent Diffusion at Millstone Nuclear Power Station (Stanford Research Institute Project 3588, Menlo Park, California, Stanford Research Institute, 1975).
- 16. Start, G.E., et al., Rancho Seco Building Wake Effects on Atmospheric Diffusion (National Oceanic and Atmospheric Administration Technical Memorandum ERL ARL-69, Idaho Falls, Idaho, Air Resources Laboratory, 1977).
- 17. Sagendorf, J.F., A Program for Evaluating Atmospheric Dispersion From a Nuclear Power Station (National Oceanic and Atmospheric Administration Technical Memorandum ERL ARL-42, Idaho Falls, Air Resources Laboratory, 1974), pp. 5-7.
- 18. Briggs, G.A., Plume Rise (U.S. Atomic Energy Commission Report TID-15075, Springfield, Virginia, Clearinghouse for Federal Scientific and Technical Information, National Bureau of Standards, U.S. Department of Commerce, 1969), pp. 27, 59.
- 19. Gifford, F.A., "Atmospheric Transport and Dispersion Over Cities," Nuclear Safety, 13, Sept-Oct., 1972, pp. 391-402.
- 20. Egan, B.A., "Turbulent Diffusion in Complex Terrain," in Lectures or Air Pollution and Environmental Impact Analyses, Boston, Mass., American Meterological Society, 1975, pp.
112-135.
- 21. Murphy K.G. and Campe, K.M., "Nuclear Power Plant Control Room Ventilation System Design for Meeting General Criterion 19," in 13th AEC Cleaning Conference (CONF 740807, Springfield, Virginia, National Technical Information Service, 1974), pp. 401-430.
UFSAR Chapter 2 Catawba Nuclear Station 2.3 - 24 (09 OCT 2016)
- 22. Smith, Maynard E., ed., Recommended Guide for the Prediction of the Dispersion of Airborne Effluents, New York, The American Society of Mechanical Engineers, 1968, p. 55.
- 23. Murphy, K.G., U.S. Atomic Energy Commission, personal communications, 1974.
- 24. U.S. Weather Bureau, U.S. Department of Commerce, 1963 - 1980: "National Summary",
Vol. 13, No. 13 - Vol. 31, No. 13, U.S. Weather Bureau, Washington, D.C.
- 25. U.S. Weather Bureau, U.S. Department of Commerce, 1956 - 1958: "National Summary",
Vol. 7, No. 13 - Vol. 9, No. 13, U.S. Weather Bureau, Washington, D.C.
- 26. National Oceanic and Atmosphere Administration, U.S. Department of Commerce, 1959 -
1980: "Storm Data", Vol. 1, No. 1 - Vol. 22, No. 12, National Climatic Center, Asheville, N.C.
- 27. Fujita, T. T. 1973: "Results of FPP Classification of 1971 and 1972 Tornadoes", Eighth Conference on Severe Local Storms, American Meteorological Society, Boston, Mass.
- 29. Duke Power Company, "Cooling Tower Effects on Area Fog Events - Catawba Nuclear Station - Clover, South Carolina", Report, April 1988.
- 30. J.V. Ramsdel, Jr. and C.A. Simonen, Atmospheric Relative Concentrations in Building Wakes, NUREG/CR-6331, Revision 1, May1997.
- 31. USNRC, Atmospheric Relative Concentrations for Control Room Radiological Habitability Assessments at Nuclear Power Plants, Regulatory Guide 1.194, June 2003.
THIS IS THE LAST PAGE OF THE TEXT SECTION 2.3.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 1 2.4 Hydrologic Engineering 2.4.1 Hydrologic Description 2.4.1.1 Site and Facilities The location and description of Catawba Nuclear Station, presented in Section 2.1, includes reference to figures showing the general arrangement, layout, and relevant elevations of the station. Yard grade is 593.5 ft mean sea level (msl) (180.9 meters), 1.1 ft (0.3 m) above the probable maximum flood static water elevation of 592.4 ft (l80.6 m) ms1. The mezzanine floor elevation in the turbine, auxiliary and service buildings is 594 ft (l8l.l m) ms1. Exterior accesses to these buildings are at elevation 594 ft (181.1 m) ms1.
The crest of the Standby Nuclear Service Water Pond (SNSWP) Dam is at elevation 595 ft (181.4 m) ms1, and the NSW and SNSW pumps are located in a concrete structure, not subject to flooding or wave effects as discussed in detail herein and in Section 3.4.2.
The statFion has an all-weather access road from county road 1132. Changes to the natural drainage of the original site are shown on Figure 2-20.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.1.2 Hydrosphere The main hydrologic features influencing the plant site are the Catawba River and Lake Wylie.
The headwaters of the Catawba are at the Blue Ridge Divide (Eastern Continental Divide) near Old Fort, North Carolina. The river flows east and then south where it joins the Wateree River at Lake Wateree near Camden, South Carolina. The Catawba is approximately 240 mi (386 km) long and has a drainage area of approximately 4750 sq mi (12,298 km2) above Wateree Dam.
Lake Wylie, created in l904 with the construction of a dam on the Catawba River for hydroelectric power production was increased to its current size in 1925 with the rebuilding and raising of Wylie Dam. The lake extends north from Wylie Dam up the Catawba River 28 mi (45 km) to Mountain Island Dam. The impoundment also extends approximately five miles up the South Fork of the Catawba River. At full pond elevation 569.4 ft (173.6 m) ms1, Lake Wylie has a surface area of 12,455 Ac (50 km2), a shoreline of about 325 mi (523 km) a volume of 281,900 Ac-ft (3.46 x 108 m3) and a mean depth of 22.5 ft (6.9 m). Its total watershed is approximately 3020 sq mi (7819 km2) yielding an average flow of 4212 cfs (119 m3/sec) giving a retention time of 34 days.
There are eleven hydroelectric reservoirs and dams along the Catawba River. Table 2-51 shows the drainage area, ownership, date of first operation, spillway design flow and seismic design criteria for these structures, and Figure 2-21 shows the major hydrologic features. No additional water control structures are planned for the Catawba River.
Rock Hill and Belmont, North Carolina take their raw water supplies from Lake Wylie. Table 2-52 lists the owner, location, and use rate of the surface water withdrawals from Mt. Island Lake to Wateree Dam. Figure 2-22 shows the location of surface water withdrawals. Groundwater use is discussed in Section 2.4.13.
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 2 (09 OCT 2016) 2.4.2 Floods Note:
This section of the FSAR contains information on the plant level bases and criteria on flooding from external sources. Additional information that may assist the reader is provided in the Plant Design Specification for Flooding From External Sources (CNS-1465.00 0011).
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.2.1 Flood History The maximum flow recorded for the Catawba River at USGS gauge number 1460 near Rock Hill, South Carolina is 151,000 cfs (4273 m3/s) on May 23, 1901. The period of record for this gauge is 1895 to 1903 and 1942 to the present. Two major floods not recorded by the gauge are the flood of 1916 with an estimated flow at Wylie Dam of 299,400 cfs (8473 m3/s) and the flood of 1940 with an estimated flow of 169,160 cfs (4748 m3/s). Six reservoirs exist on the Catawba River upstream from the station and Lake Wylie. They have a combined usable storage of approximately 1.5 million Ac-ft (1.85 x 109 m3). Because of such a large volume of storage, the floods of record are well modified and the annual flood peaks on the main stem of the Catawba do not represent the uncontrolled flood potential of the basin. Table 2-53 shows the return period of annual peak floods for the Catawba River at the USGS gauge near Rock Hill.
The flood of August 1940 caused Lake Wylie to reach a maximum surface water elevation of 575.0 ft (175 m) msl, 5.6 ft (1.7 m) above full pond.
2.4.2.2 Flood Design Considerations Flood levels for the site are analysed for the following flood producing phenomena:
- 1. Probable Maximum Flood (PMF) resulting from the probable maximum precipitation in the drainage area.
- 2. A 25 year frequency flood passing through Lake Wylie combined with a seismic failure of Cowans Ford Dam, the largest upstream reservoir.
- 3. A Standard Project Flood (SPF) passing through Lake Wylie combined with the failure of one of the upstream dams due to an Operating Basis Earthquake (OBE). The SPF is considered equal to one-half of the PMF.
The effect of wind on wave height and runup at this site is also analyzed.
The maximum static water elevation of 592.4 ft (180.6 m) msl occurs during a SPF combined with the failure of Cowans Ford Dam. The station yard elevation is at elevation 593.5 ft (180.9 m) msl.
Conservative engineering analysis, such as those presented in Regulatory Guide 1.59 "Design Basis Floods for Nuclear Power Plants", are used in the PMF analysis. Appendix A of Regulatory Guide 1.59 was used in making the flood study evaluation at Catawba. In summary, descriptions for determining probable maximum flood, hydrologic characteristics, flood hydrograph analysis, precipitation losses and base flow, runoff model, probable maximum precipitation estimates, channel and reservoir routing, seismically induced floods, water level determinations, and coincident wind-wave activity are provided in Section 2.4.3.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 3 2.4.2.3 Effects of Local Intense Precipitation HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.2.3.1 General The plant site is provided with a surface water drainage system that is designed and constructed to protect all safety-related facilities from flooding during a local probable maximum precipitation (PMP). The drainage system consists of 1) catch basin inlets which are connected by corrugated metal pipes to form several networks and 2) graded areas which permit free surface outflow to Lake Wylie when ponding in the powerhouse yard reaches elevation 593.5 feet. All pipe networks and graded areas discharge at elevations which are higher than Lake Wylie full surface elevation (Elev. 569.4). A discussion of the PMP critically centered over Lake Wylie and the resulting lake elevations are discussed in Section 2.4.3.4.
During a local intense PMP, water will pond in the powerhouse yard to a maximum elevation which is below the entrances to any safety-related structure (594.0 feet). The following sections describe the detailed analyses which were performed to verify the adequacy of the site drainage system for controlling runoff during a local intense PMP.
2.4.2.3.2 Probable Maximum Precipitation and Runoff Models Based upon the Hydrometerological Report No. 33 (Reference 1), the site index PMP is 30.1 in.
(76.5 cm.) within a six-hour period, 32.5 in. (82.6 cm.) within a twelve-hour period, and 35.2 in (89.4 cm.) within a twenty-four-hour period. To provide for the imperfect fit of storm isohyetal patterns to the shape of a particular basin, the Corps of Engineers recommended a transposition coefficient for reducing the PMP when the storm area is larger than the drainage basin area. In accordance with the Corps Engineering Circular EC 1110-2-27, the minimum transposition coefficient of 0.8 is applicable to the site index PMP. The following equation is used to determine the coefficient:
]
TRSDA
/
3008
.0 1[
TPSPC 17718
.0
where TPSPC
= transposition coefficient (0.80 minimum)
TRSDA
= site drainage area = 0.20 sq. mi.
The PMP used in the analyses is therefore 24.1 in. (6l.2 cm.) for the peak six-hour period, 26.0 in. (66.0 cm.) within a twelve-hour period, and 28.2 in (71.6 cm.) within a twenty-four-hour period. The peak six-hour precipitation (24.1 in.) is distributed according to the U. S. Army Corps of Engineers procedure (Reference 2) as follows:
Time (Ending Hour) Incremental PMP Accumulative PMP Percent Inches Percent Inches 1
10 2.4 10 2.4 2
12 2.9 22 5.3 3
15 3.6 37 8.9 4
38 9.2 75 18.1 5
14 3.4 89 21.5
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 4 (09 OCT 2016)
Time (Ending Hour) Incremental PMP Accumulative PMP Percent Inches Percent Inches 6
11 2.6 100 24.1 To obtain time dependent inflow for the PMP, the site is divided into sub-areas as shown on Figure 2-23. Inflow to the power block area is due to precipitation which falls on 129 acres, which include the powerhouse yard, buildings, and the construction yard. The switchyard and cooling tower yard sub-areas have been bermed/curbed to route water away from the power block area. Elevations of these berms/curbs are conservatively set by neglecting outflow from the switchyard and cooling tower yard storm drainage systems. Inflow to the switchyard and cooling tower yard sub-areas is due to precipitation which falls directly on the yards and structures, excluding the cooling towers. Sections 2.4.2.3.4 and 2.4.2.3.5 present the flood routing for the switchyard and cooling tower yard, respectively.
The inflow hydrograph for the power block area is determined by combining the inflow hydrograph for the construction yard and powerhouse yard sub-areas. Each hydrograph is based on a Soil Conservation Service (SCS) dimensionless unit hydrograph. The HEC-1 computer program (Reference 21) was used to develop a total hydrograph for a given lag time with lag time (L) being defined as the time in hours from the center of mass of rainfall excess to the peak discharge. Lag time is calculated using the following equation (Reference 22):
5.0 7.0 8.0 y
1900
)1 S
(
L
where
=
= hydraulic length in feet y
= slope in percent S
= maximum retention = 0.20 For the construction yard, a hydraulic length of 1815 feet and an average slope of 1.18 percent is obtained from site topography as shown on Figure 2-23. The maximum retention is based on the SCS Curve Number 98 and results in a total rainfall loss of 0.12 inches for the 24-hour PMP.
Lag time for the construction yard sub-area is therefore l3.4 minutes. The SCS has determined that the time of concentration is approximately 1.67 times the lag time; therefore, the time of concentration for the construction yard is 22.4 minutes. Lag time for the powerhouse yard sub-area is conservatively assumed to be zero to approximate an instantaneous time of concentration.
Since the time of concentration used to develop the inflow hydrograph is small, the peak one-hour precipitation (9.2 in.) is distributed according to the U. S. Army Corps of Engineers procedure for five minute durations as shown below:
Time (Ending Hour)
Incremental PMP Accumulative PMP Percent Inches Percent Inches 5
3 0.28 3
0.28 10 4
0.37 7
0.65 15 5
0.46 12 1.11
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 5 Time (Ending Hour)
Incremental PMP Accumulative PMP Percent Inches Percent Inches 20 6
0.55 18 1.66 25 9
0.83 27 2.49 30 17 1.56 44 4.05 35 25 2.30 69 6.35 40 11 1.01 80 7.36 45 8
0.74 88 8.10 50 5
0.46 93 8.56 55 4
0.37 97 8.93 60 3
0.27 100 9.20 The following sections present the methodology and resulting water surface elevations for routing the inflow hydrographs through the powerhouse yard, switchyard, and cooling tower yard.
2.4.2.3.3 Powerhouse Yard 2.4.2.3.3.1 Yard Characteristics Figure 2-24 shows detailed drainage features for the power block area. Included on the figure are locations of catch basins, locations of permanent and temporary facilities, locations and elevations of perimeter fences, location and elevation of cable trenches, roads, and railroads.
During a local intense PMP, water will pond in the power block area and on the roof of the service building. With the exception of the reactor building, the roofs of safety-related structures are designed with no obstructions so that water flows directly off roofs and there is no accumulation. A gutter drain system catches the water and routes it to collection points which discharge directly into the yard drainage system. The reactor building roof drainage system is designed for a rainfall intensity of 5.0 in./hr. (12.7 cm./hr.). Intensities in excess of 5.0 in./hr.
(12.7 cm./hr.) result in ponding; however, once the water level reaches El. 7ll.34 feet (2l6.8 m) msl, the water flows directly off the roof. The reactor building roof is designed to safely carry live loading due to ponding, as discussed in Section 3.8. In determining the effect of a local intense PMP on the powerhouse yard, it is assumed that water flows directly off the reactor building and service building without ponding or discharging through the roof drainage system. Water which ponds in the power house yard will discharge into catch basins and over the northeast and south ends of the yard.
Two types of catch basin inlets are designed for the site as shown on Figure 2-25. Type II inlets consist of slotted catch basin covers with an effective opening of 0.69 square feet. Type I inlets have no slotted cover, but are protected by the steel grating on the four sides and top. The total effective opening in the grating on any one side is at least equal to the effective opening of the pipe inlet (1.48 sq. ft.), but may be as much as 3.9 times the effective opening. The open area provided by all four sides of the Type I inlet varies from 4 to 15.6 times the pipe opening, virtually eliminating the possibility of complete blockage by debris accumulation. Catch basin inlets are shown in Figure 2-24. Eighty-nine Type I inlets are provided in the power block area,
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 6 (09 OCT 2016) but only eighty are required to be operable at any given time. All inlets are connected to corrugated metal pipes which are fully coated with a paved invert.
The yard drainage pipes, individually and within a network, are designed using Manning's equation for pipe flowing full. Accumulative totals are used throughout the networks to determine pipe sizes. All pipe gradients are 0.5 percent or greater. Energy equations are used to verify that each network is capable of discharging inflows during a PMP, conservatively assuming all inlets, including those in the switchyard, are 100% efficient. Invert elevations at pipe discharge points are shown on Figure 2-23.
Water in the power block area is assumed to rise and fall as a "level pool." Inspection of Figure 2-23 indicates conservatism in the level pool assumption due to the proximity of the free surface outflow areas to safety-related structures. Obstructions in the powerhouse yard, which may reduce the volume of water near safety-related facilities, have been conservatively neglected.
One obstruction is the gravel or concrete berms which are located at the base of sections of the administration and protected area fences. Figure 2-3 indicates that these berms are located in a manner which adds conservatism to the level pool assumption. Details of the conditions at the base of the fences are shown on Figure 2-26. Another obstruction in the powerhouse yard is the cable trenches, with top elevation 594.0, which are located near the Turbine Building. These trenches will not impede water flowing away from the safety-related structures.
All railroad tracks and roads in the powerhouse yard are at elevation 593.5, except as shown on Figure 2-3.
2.4.2.3.3.2 Water Level Determination Water levels in the power block area are predicted by routing the inflow hydrograph through the area using the modified Puls routing method (Reference 20). The HEC-1 computer program or an Excel Spreadsheet was used to perform the routing with a five minute time increment. Input for the program includes a storage versus elevation relationship and outflow rating curves for catch basin inlets and sheet outflow areas.
Storage in the power block area is determined by assuming the area is an inverted pyramid with a top plane of 37.9 acres and a corresponding apex depth of 1.27 feet below the yard high point (Elev. 593.5 ft.). This assumption accounts for the depressed drainage area around each catch basin inlet. The top area represents 54.6 acres in the power block area less 16.7 acres which are occupied by the permanent and temporary structures as shown on Figure 2-3. Storage on the roof of the service building and reactor building is neglected. The following elevation-storage data is used in the analysis:
Elevation, Ft.
Storage, Acre-Ft.
592.23 0
592.50 0.7 592.75 2.7 593.00 5.9 593.25 10.3 593.50 16.0 593.75 25.5 594.0 35.0
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 7 The outflow rating curve for the catch basin inlets is determined using the orifice equation:
gH 2
ca Qo where Qo
= orifice outflow, cfs c
= orifice coefficient = 0.60 a
= sum of effective opening of all catch basin inlets g
= acceleration due to gravity = 32.2 ft/sec2 H
= depth of ponding above average catch basin inlet elevation (592.23 feet)
Three orifice rating curves, based on 0%, 50%, and 100% clogging of Type II inlets, are used.
As previously discussed, Type I inlets are designed to prevent clogging of the vertical pipe. The following table lists discharge data for 80 Type I inlets and 6l Type II inlets:
Elevation (feet) 100% Type II Discharge, cfs 50% Type II 0% Type II 592.2 0
0 0
592.5 401.5 348.9 296.2 592.8 583.4 506.9 430.4 593.1 720.8 626.3 531.7 593.4 835.9 726.3 616.7 593.7 936.9 814.1 691.2 594.0 1028.1 893.3 758.5 Once ponding reaches the yard high point (Elev. 593.5), sheet outflow over the northeast and south ends of the yard begins. Discharge data for the sheet outflow is determined using the weir equation:
2
/
3 CLh Qw where Qw
= weir outflow, cfs C
= weir coefficient = 2.70 L
= length of weir = 700 feet h
= depth of ponding above yard high point A check on the hydraulic conductivity of the area below the outflow weirs is made using Manning's equation for open channel flow with 1) a Manning Coefficient, n = 0.060, for natural channels flowing sluggish with weedy or deep pools, and 2) a ground slope of 0.333 (3H:1V).
Resulting flow is supercritical, therefore confirming that the area will not adversely affect water levels in the power block area.
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 8 (09 OCT 2016)
Figure 2-28, Figure 2-29, and Figure 2-30 present inflow hydrographs, outflow hydrographs, and staging curves from flood routings for 0%, 50%, and 100% blockage of Type II inlets, respectively. Maximum water elevation in the powerhouse yard will be below the entrances to any safety related structures (594.0 feet) for all three cases (0%, 50%, and 100% blockage of Type II inlets).
2.4.2.3.3.3 Modifications and Inspections Any modifications that could have an impact on the minimum 80 required Type 1 catch basin inlets or associated buried piping are evaluated and implemented in accordance with the Nuclear Station Modification program. Reference Section 2.4.2.3.3.1 for a description of the yard drainage characteristics. Inspections of the required catch basin inlets were done prior to Unit 1 fuel loading and are conducted annually. Adverse conditions, which could impact the flood vulnerability of the major QA Condition 1 structures, are corrected.
2.4.2.3.4 Switchyard The PMP discussed in Section 2.4.2.3.2 is routed through the switchyard using the modified Puls method for an instantaneous time of concentration. It is conservatively assumed that no outflow occurs from the switchyard drainage system. A curb (Elevation 632+8) is provided on the North, South, and East ends of the switchyard as shown in Figure 2-23 to prevent water from flowing onto the powerhouse yard. In addition, trench plugs are provided at the curb location in all trenches that communicatet between the switchyard and the powerhouse yard.
The flood routing is performed with the HEC-1 computer program using storage and discharge data which are based on the switchyard topography.
Storage in the switchyard is determined by characterizing the yard as an inverted pyramid with a top area of 16.5 acres and a corresponding apex depth of 1 foot below elevation 632. The following elevation-storage data is used in the analysis:
Elevation, Ft.
Storage, Acre-Ft.
631.0 0
631.2 0.04 631.5 0.69 631.8 2.82 632.0 5.51 632.5 13.77 633.0 22.04 When ponding in the switchyard reaches elevation 632.0, sheet outflow over the west side of the yard begins. The natural topography below the overflow area will route discharge away from the site. Outflow is determined using the weir equation:
2
/
3 CLh Qw where Qw
= weir outflow, cfs
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 9 C
= weir coefficient = 2.70 L
= length of weir = 350 feet h
= depth of ponding above weir crest Figure 2-31 presents the inflow hydrograph, outflow hydrograph, and staging curve for the routed flood. Water will pond to a maximum elevation which is below the curbs.
2.4.2.3.5 Cooling Tower Yard The PMP discussed in Section 2.4.2.3.2 is routed through the cooling tower yard using the modified Puls method for an instantaneous time of concentration. It is conservatively assumed that no outflow occurs from the cooling tower yard drainage system. The area occupied by the cooling towers is neglected for the inflow and storage calculations since all precipitation which falls on the towers will be contained within the system. An earth berm (Elevation 620+6) is provided on the northwest end of the cooling tower yard as shown in Figure 2-23 to prevent water from flowing onto the powerhouse yard. In addition, trench plugs are provided at the berm location in all trenches that communicate between the cooling tower yard and the powerhouse yard. The flood routing is performed with the HEC-1 computer program using storage and discharge data which are based on the cooling tower yard topography.
Storage in the cooling tower yard is determined by characterizing the yard as an inverted pyramid with a top area of 32.4 acres and a corresponding apex depth of 1.5 feet below elevation 620.0. The following elevation-storage data are used in the analysis:
Elevation, Ft.
Storage, Acre-Ft.
618.5 0
619.0 0.46 619.5 3.65 620.0 12.31 620.5 24.62 621.0 36.93 When ponding in the cooling tower yard reaches elevation 620.35, sheet outflow over the southwest side of the yard begins. The natural topography below the outflow area will route water away from the site. Outflow is determined using the weir equation:
2
/
3 CLh Qw where Qw
= weir outflow, cfs C
= weir coefficient = 2.70 L
= length of weir = 1000 feet h
= depth of ponding above weir crest
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 10 (09 OCT 2016)
Figure 2-32 presents the inflow hydrograph, outflow hydrograph, and staging curve for the routed flood. Water will pond to a maximum elevation which is below the earth berm.
2.4.2.3.6 Site Evaluation Using HMR 51 and 52 The preceeding analysis conforms to General Design Criteria 2 of Appendix A to 10CFR 50, NRC Standard Review Plan 2.4.2, PSAR commitments and is conservative based on expected return periods for the rainfall events evaluated. The analysis indicates a maximum ponding elevation below 594.0 for the Catawba site during a Probable Maximum Precipitation (PMP) event as defined in HMR No. 33 using time distributions from COE Engineering Manual 1110 1411. The maximum ponding elevation is below exterior door entrances of safety related buildings, resulting in no adverse effect on the safe operation of the plant.
However, at the request of the NRC, Duke has evaluated site drainage using the following PMP values and rainfall distribution from HMR No. 51 (Reference 23) and HMR No. 52 (Reference
- 24) respectively.
Time Incremental PMP Accumulative PMP (Ending Hour)
Percent Inches Percent Inches 1
2.3 0.7 2.3 0.7 2
5.3 1.6 7.6 2.3 3
16.6 5.0 24.2 7.3 4
63.2 19.0 87.4 26.3 5
8.3 2.5 95.7 28.8 6
4.3 1.3 100.0 30.1 No reduction was taken to account for the imperfect fit of isohyetal patterns to the basin shape.
Rainfall percentages given in HMR No. 52 for the five, fifteen, thirty, and sixty minute intervals were used to estimate rainfall for each five minute interval during the peak one hour with the resulting rainfall distributed according to the US Army Corps of Engineers procedure as shown below.
Time Incremental PMP Accumulative PMP (Ending Minute)
Percent Inches Percent Inches 5
1.8 0.35 1.8 0.35 10 4.2 0.80 6.0 1.15 15 5.6 1.05 11.6 2.20 20 6.1 1.16 17.7 3.36 25 8.5 1.61 26.2 4.97 30 10.6 2.02 36.8 6.99 35 32.5 6.18 69.3 13.77 40 7.9 1.49 77.2 14.66
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 11 Time Incremental PMP Accumulative PMP (Ending Minute)
Percent Inches Percent Inches 45 8.4 1.60 85.6 16.26 50 6.0 1.14 91.6 17.40 55 5.0 0.95 96.6 18.35 60 3.4 0.65 100.0 19.00 Inflow hydrographs for the powerhouse yard, switchyard, and cooling tower yard were developed using the SCS method and routed using the modified Puls routing method as discussed in Sections 2.4.2.3.3, 2.4.2.3.4 and 2.4.2.3.5. Berms around the switchyard and cooling tower yard are high enough to prevent water from flowing from these yards onto the powerhouse yard.
Calculations based on the distribution above indicate that water will pond on site to a maximum elevation of 594.74 feet. Since some of the entrances to safety related structures are (0.74 feet lower) at elevation 594.0, water could enter these buildings. Doors where water could enter safety related buildings are shown on CN-1103-7.5. Concrete curbs have been installed (top el.
594+71/2) at the entrances to the Unit 1 and 2 Diesel Generator Buildings, thus minimizing inflow during a PMP event. Other exterior door entrances considered in this analysis include doors entering the Auxiliary Service Building, New Fuel Buildings, UHI Buildings, Exterior Doghouses, and the Unit 1 and 2 Turbine Building. (See FSAR Figure 1-6 for door locations).
The Turbine Building is not a safety related structure and security hardware is not required at the exterior entrances. All the doors entering the Turbine Building are conservatively assumed open during the entire PMP event. The inleakage through each door is calculated by the weir equation:
2
/
3 CwLH Q
where:
Cw
=
Weir Coefficient (3.0)
L
=
Total Door Widths (Ft.)
H
=
Driving Head, the vertical distance measured from the water surface to elevation 594.0 As water enters the Turbine Building at elevation 594+0 it is intercepted by the floor drain system and numerous large openings in the floor slab. All the water is routed to the basement level of the Turbine Building at elevation 568+0. Neglecting any pumping, the water ponds to a maximum elevation of 568'+6. A 12 ft. high concrete flood barrier, located on column line 34, retains all the water within the Turbine Building (see FSAR Figure 1-5). Therefore, no water enters the safety related areas from the Turbine Building. For increased conservatism against blockage of catch basins during a design basis rainfall, 80 catch basin inlets are modified to provide a minimum of 4 times and as much as 15.6 times the required discharge capacity. The modified inlets, identified as Type I on Figure 2-25, are designed to eliminate the possibility of complete blockage. The resulting water elevations for postulated conditions of blockage associated with unmodified catch basins is discussed in FSAR Section 2.4.2.3. We have committed in FSAR Section 2.4.2.3 to inspect the catch basins in accordance with Regulatory
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 12 (09 OCT 2016)
Guide 1.127. Additionally, the site is graded to provide 700 feet of sheet outflow area and will provide berms around the switchyard and cooling tower yard to divert PMP storm runoff from the powerhouse yard.
All doors entering Units 1 and 2 Auxiliary Building (Elec. Pen. Room), Outside Doghouse, former UHI Building, New Fuel Building and the Auxiliary Service Building are assumed closed during the thirty-five minute period water remains above elevation 594.0. All doors entering safety related buildings are pressure doors (designed for 3 psi). All normally used exterior doors are equipped with automatic closures, except for equipment access doors AR6 and AR5. These doors are controlled from inside the Auxiliary Service Building, and cannot be opened from the outside. Station Security by procedure ensures that the Auxiliary Building and Auxiliary Service Building exterior doors are closed in the event of severe weather. The leakage into the building around cracks at each door is calculated by the orifice equation:
gh 2
A C
Q D
where:
C D
=
Coefficient of discharge (.61)
A
=
Crack size x door width (Ft.2), includes crack at the sides of each door as the water elevation increases (see CN-1103-7.5).
g
=
gravitational acceleration (32.2 Ft/Sec2) h
=
driving head, measured to the bottom of the orifice (Ft)
(Note: The crack size used to determine the inleakage are consistant with manufacturer's drawings and specifications. Field measurements verify that the actual crack sizes are much smaller than those used in the analysis).
As water enters the building it spreads across the floor and is intercepted by the floor drain system (all openings in the floor slabs of safety related structures are sealed to serve as fire barriers). The floor drain system routes the entire volume of water to four floor drain sumps and a floor drain tank, all located in the Auxiliary Building at elevation 543+0. (See FSAR Figure 1-3 for locations). Conservatively assuming no pumping, water will pond as documented in CNC-1203.03-00-0142 with no adverse effect on plant equipment required for shutdown.
Deleted per 2009 Update.
Thus, using the conservative PMP values and time distributions from HMR No. 51 and No. 52, no safety related equipment will be affected by the resulting water levels from the PMP event, and the plant can be safely shutdown by established normal shutdown procedures.
2.4.3 Probable Maximum Flood (PMF) on Streams and Rivers HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.3.1 Probable Maximum Precipitation A search, made of historically great storms which occurred near the Catawba River basin is used to obtain the hypothetical flood characteristics of peak discharge, volume, and hydrograph shape considered to be the most severe "reasonably possible" at the Catawba site. The storms, listed on Table 2-55, are believed to be the greatest to occur in the southeastern part of the
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 13 country. Table 2-55 lists the locations of storm centers and maximum rainfall depths and durations for a drainage area equal to that above the Wylie Dam.
The greatest storm over the Wylie drainage area is recorded for the period July l3-l7, 1916.
However, greater amounts of precipitation occurred in Elba, Alabama, and Bonitoy and Yankeetown, Florida as shown in Table 2-55. It is of note that the later storms all occurred immediately along the coastal area and are expected to produce diminishing amounts of precipitation by transposing these storms inland some 200 mi (322 km) to the Wylie watershed.
Maximum depth duration of rainfall, from a study made by the Hydrometeorological Section of the Weather Bureau for the Savannah River above Hartwell dam site, is included in Table 2-55 for comparison purposes since the location is very close to the Catawba River watershed. The Savannah River study uses the storm of July 13-17, 1916 as a guide for the determination of maximum rainfall. However, to arrive at a maximum possible precipitation and transposing the storm over the Savannah River basin, an adjustment is made to increase the precipitation values by 42 percent. To arrive at the Probable Maximum Precipitation over the Catawba River basin, the July 13-17, 1916 storm is selected based upon meteorological and physiographic considerations as a guide to the determination of time and area rainfall distribution pattern. The following adjustments are made to this storm to increase its magnitude and intensity to such values considered to be equal to the Probable Maximum Precipitation over the Catawba River basin.
- 1. Rainfall depth-duration values are distributed in accordance with that of the 1916 storm.
- 2. Storm position is transposed over a limited distance within the Catawba River basin to produce a maximum concentration of precipitation over a selected area.
- 3. Precipitation amounts are increased 40 percent.
In the area of the Wylie watershed, snow melt is not a consideration because of its southern location. An isohyetal map for the PMP is shown on Figure 2-34. Hourly incremental rainfall and rainfall excess for the 54 hour6.25e-4 days <br />0.015 hours <br />8.928571e-5 weeks <br />2.0547e-5 months <br /> period of the Probable Maximum Precipitation are available for inspection in the Regulatory Compliance Licensing Files as former FSAR Figure 2-35 and Figure 2-36.
2.4.3.2 Precipitation Losses The topography of the Catawba River basin is gentle to moderate, sloping toward the river in a southeasterly direction. The soil designated according to the National Cooperative Soil Survey Classification of 1967 (Reference 3) is Ultisoil U5-3. These soils are low in bases and have subsurface horizons of clay accumulation; usually moist, but during the warm season of the year some are dry.
Initial loss for conditions usually preceding major floods in humid regions normally range from about 0.2 (0.5 cm) to 0.5 in. (1.3 cm) and is relatively small in comparison with the flood runoff volume. A value of 0.5 in. is taken for initial loss in the study.
Infiltration rates vary throughout the storm period from a high rate at the beginning to a relatively low and uniform rate as the precipitation continues. It is common practice to assume infiltration rates from 0.05 (0.13 cm) to 0.10 in. (0.25 cm) per hour, depending upon antecedent field moisture conditions, slope, and soil type. A hydrologic study made by the Corps of Engineers (Reference 4) of the Saluda River Basin above Chappels, South Carolina indicates an infiltration rate of 0.12 in. (0.30 cm) per hour. The Saluda River Basin is 1290 sq mi (3340 km2) and lies approximately 80 mi (129 km) southwest of the Catawba River. The topography, soil group, and climate of both basins are very similar. Based upon this study, an infiltration rate of 0.10 in.
(0.25 cm) per hour is used.
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 14 (09 OCT 2016) 2.4.3.3 Runoff and Stream Course Models To obtain time dependent inflow to the Catawba River from the probable maximum precipitation, each reservoir's drainage area is divided into subareas depending on the number of larger tributary streams flowing into each reservoir as shown on Figure 2-35. Synthetic unit hydrographs are derived by method of dimensionless unit hydrographs (Reference 5). Synthetic unit hydrograph coefficients are derived from the storm of September 29, 30, 1958 (Hurricane Gracie) for nine tributary streams from which gaging records are available. The relationship between subarea hydrologic characteristics and lag time obtained from this study is shown on Figure 2-36. Normally, unit hydrographs are not consistent for all magnitudes of floods; peaks increase and lag times shorten with larger floods. However, constant unit hydrographs are used in this study since peak flows are not as critical as volume runoff. The lag time is reduced in this study so that it may more closely represent conditions during a large flood runoff. This reduction shown by the adjusted lag time curve on Figure 2-36 and the unit hydrographs thus derived are shown on Figure 2-37. The subdivision of any reservoir into principal subareas is necessary for the purpose of reflecting more accurately the nonuniformity and vary-intensity of rainfall over any reservoir drainage area.
Hourly precipitation amounts are distributed to existing precipitation stations by the Thiessen polygon method. Existing precipitation stations and Thiessen polygons are shown on Figure 2-
- 35. The area of each polygon falling within a subarea is expressed as a percentage of the total subarea. These percentages are the rainfall-subarea coefficients assigned to each precipitation station rainfall. A tabulation of these coefficients is shown on Table 2-58.
The steps used to synthesize flood flow into reservoirs are summarized below :
- 1. The applicable portion of each rainfall station's precipitation is converted directly into inflow for that portion of the polygon covered by reservoir water surface.
- 2. The runoff (rainfall less losses) from each rainfall station is applied to the percentage which the precipitation station Thiessen polygon bounded. These values for each precipitation station are summed for each subarea, resulting in the average hourly rainfall excess for the subareas. The average hourly rainfall excess in inches is then applied to each subarea unit hydrograph, resulting in a storm hydrograph of local inflow for each subarea for each hour of runoff.
- 3. The total inflow to each reservoir consists of local inflow from each subarea of the reservoir local drainage area, plus local inflow due to reservoir surface rainfall, plus upstream flow, plus base flow.
The flood resulting from the PMP is routed through the Catawba River system to Wylie Dam by means of a flood routing program (Reference 6).
Local reservoir inflow for each hour of the storm and for each reservoir is calculated as indicated. Reservoir elevations are computed from reservoir inflows and with the discharges corresponding to these elevations. The method of computing reservoir elevations and discharges is one of successive approximations to satisfy the relation that inflow is equal to outflow plus storage. The outflow from dams with gravity overflow spillways is uncontrolled and a function of the reservoir elevation. Outflow from dams with gated structures is dependent on operation of the gates as described below. The discharge from the first reservoir allowing for lag time, where applicable, is added to the local inflow for the next reservoir and the routing procedure is repeated for all reservoirs for each hour of the storm plus any additional hours needed to cover the complete runoff.
Reservoirs with gated spillways are operated on the basis of first filling or attempting to fill the reservoirs to a predetermined control elevation and thereon assuming outflow equal to inflow
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 15 until the gate capacity is exceeded. These levels are shown in Table 2-59 and are based on historical water level records for late summer and early fall. Discharges are then limited to gate capacities at given elevations and the reservoirs begin to rise if inflow continues to exceed gate capabilities. When reservoir elevations start to fall, which occurs when inflows become less than outflows, discharges are continued at the capacity of the gates until such time as the reservoirs reach their predetermined control elevations. Subsequently, the reservoir levels are held constant for the remaining period of the flood; outflow becomes equal to flow during this period.
Discharges from generation of power are assumed to continue throughout all gate operation procedures unless reservoir levels overtop bulkheads protecting powerhouses or switchyards, at which time the discharge through the powerhouse is stopped for the remainder of the storm period. Discharges through the turbines are shown in Table 2-59. The effect of not having any power generation releases is also calculated.
Operation of all floodgates is done at the direction of personnel from the System Operations Center. The decision to operate the floodgates is based on water elevations, rainfall, power plant operations, and weather forecasts over the entire Duke System. The spillway gates are opened as necessary to maintain the reservoirs at or close to the normal operating level. The reservoir levels are monitored hourly. The relationship between elevation, storage, and discharge for each lake is shown in Table 2-60.
Details of each dam upstream of Wylie Dam are presented in the Catawba Nuclear Station PSAR, Appendix 2F ("Report on Flood Studies for Proposed Catawba Nuclear Station", Chas.
T. Main, Inc., June, 1972). Details of access provisions and facilities for floodgate operation for Oxford, Cowans Ford, and Wylie Dams are as follows:
Oxford There are ten gates on this spillway operated by a single gantry crane. Access is normally from the powerhouse, which is attended at all times, but access is also available from both ends of the dam. The top of the bulkhead is at elevation 950 ft (290 m) ms1, which will not be overtopped in the first 37 hours4.282407e-4 days <br />0.0103 hours <br />6.117725e-5 weeks <br />1.40785e-5 months <br /> of the storm causing the maximum flooding in this reservoir (McGuire FSAR, Appendix 2F, Plate IV). Access to the gantry for gate operation is therefore assured. All facilities to operate the gates are maintained on-site. Power is supplied normally by a motorgenerator set which is supplied from the 6600/575 V station service transformers.
Alternate dc power can be supplied by making a temporary connection from No. 1 Unit exciter.
Cowans Ford There are 11 gates on this spillway operated by either of two hoists which travel along the top of the structure. Access is normally through the power house, but access is also available from both ends of the dam. The top of the bulkhead is at elevation 770.0 ft (235 m), which will have a freeboard of 2.1 ft (0.6m) during the storm causing the maximum flooding in this reservoir (McGuire FSAR, Appendix 2F, Plate VI). Access to the gantry for gate operation is therefore assured.
The plant superintendent and one serviceman live on the property, and other personnel live in the immediate area. All facilities to operate the gates are maintained onsite. There are two normal station service transformers and two separate feeders either of which can supply power to the hoists. There is also a standby station service transformer supplied from an external 44 kV source, which can automatically replace either normal station service. It is possible to operate the hoists by hand with a crank if necessary.
Wylie
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 16 (09 OCT 2016)
There are 11 gates on this spillway, five of which are operated by one overhead trolley-type crane, and the remaining six are operated by another identical crane. Access is normally through the powerhouse, which is attended at all times, but access is also available from both ends of the dam. The top of the bulkhead is at elevation 584.4 ft (178 m), which will have a freeboard of 4.4 ft (l.3 m) during the Probable Maximum Flood in this reservoir. Also, the bulkhead will not be overtopped in the first 31 hours3.587963e-4 days <br />0.00861 hours <br />5.125661e-5 weeks <br />1.17955e-5 months <br /> of the 25-year storm with Cowans Ford Dam failure. Access to the gantry for gate operation is therefore assured. All facilities to operate the gates are maintained on-site. Power is supplied by a motor-generator set which is supplied from the 6600/575 V station service transformers.
In cases where reservoir elevations overtopped the bulkheads of reservoirs, it is assumed the bulkheads do not fail. Appropriate discharges are provided for in the routing program considering these structures as broad-crested weirs.
Where earth embankments are overtopped, it is assumed that there are failures for a small amount of overtopping and the discharges followed that of broad crested weirs. However, when the flow depth reaches two feet or more, it is assumed that progressive failures of the embankments by erosion take place along the overtopped crests and exposed ends.
2.4.3.4 Probable Maximum Flood Flow Routing the PMF through the Catawba River system results in the inflow and outflow hydrographs for Lake Wylie as shown in Figure 2-38, Run No. 27. The maximum inflow is calculated to be 591,300 cfs (15,374 m3/s) with a maximum discharge of 512,700 cfs (14,509 m3/s).
An examination of the Catawba River profile shows that the entire reach from Wylie Dam to the headwater of Bridgewater Reservoir, a distance of 154 mi (248 km), is completely controlled by dams and reservoirs with the exception of one reach of 18 mi (29 km) in length. This reach, outside of the influence of any reservoir, does not seriously hinder the passage of a flood wave.
Considering the passage of a flood wave, it is readily concluded that such a wave is gradually dissipated by reservoir storage and to some degree by valley storage in the uncontrolled reach.
Since Duke normally operates these reservoirs below full pool levels, there is generally a sizable amount of storage available. In addition, surcharge storage is also available. As a conservative measure, no attenuation of a flood wave is provided for in routing through this one channel reach outside the influence of any reservoir.
Several trial positions of the Storm Center are made to determine the most critical postion for producing the maximum flood over Wylie Reservoir. The storm center is positioned over each of the reservoir drainage areas in turn and then routed through the Catawba River system into Wylie Reservoir. Figure 2-35 illustrates the isohyetal map of the adjusted 1916 storm over the drainage basin subareas used in computing the PMF. The location which produced the highest reservoir elevation at Wylie Dam of 580.0 ft (177 m) msl results in centering the storm over Wylie drainage area as shown in Table 2-61.
In determining the PMF that would produce the highest maximum water surface elevation on Wylie Reservoir it is assumed that normal operational procedures allowing the use of the flood gates at Wylie Dam will be followed. All of the flood gates at Wylie Dam are inspected and physically operated yearly by FERC regulation. These gates are operated by electric hoist which are powered by two independent power sources, Wylie Hydroelectric Station and/or from the Duke transmission system power grid. There are two sources available at Wylie Station; ac motor driven dc generator units or dc excitor units. It should be noted that Lake Wylie has never exceeded an elevation at the dam of 570.10 ft (100.10 Duke datum), and that during the flood of August 1940 the maximum elevation was 570.0 ft. with a peak discharge of 129,807 cfs. Based
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 17 on the above discussion faulty gate operation coincident with the PMF on Lake Wylie is not considered valid and consequently has not been evaluated (see Section 2.4.3.3).
The 18 mile (29 km) reach of river between Bridgewater and Rhodhiss Reservoirs passes the peak outflow of 188,100 cfs (5323 m3/s) from Bridgewater Reservoir as the result of routing the PMF. The maximum depth of flow is computed to be 31 ft (9.4 m) and the average velocity of 6.9 ft/s (2.1 m/s). This reach of river does not flow through any populated area, and flood damages are therefore at a minimum.
For increased conservatism a trial calculation of the PMF level in Lake Wylie is computed assuming no hydroelectric units operating and conservative assumptions of initial rainfall loss of 0.5 in. (1.27 cm) in. and infiltration loss of 0.05 in. (0.127 cm) per hour. The reservoir inflow, outflow, and elevations for these conditions are given in Table 2-62. The PMF level of 583 ft (178 m) msl is 1.87 ft (0.57 m) higher than the PMF under similar operating conditions, and a higher infiltration rate of 0.1 in. (0.25 cm) per hour.
Floods resulting from dam failures are discussed in Section 2.4.4. The maximum calculated water level for Lake Wylie is 592.4 ft (180.6 m) msl, which assumes a seismic failure of Cowans Ford Dam coincident with a Standard Project Flood.
2.4.3.5 Water Level Determination The method of computing reservoir elevation is based on the relation that inflow minus outflow equals the change in storage. The total inflow to each reservoir consists of local inflow from the reservoir drainage area, rainfall on the lake surface, upstream inflow plus base flow. Reservoir outflow is dependent upon reservoir elevation and gate operation as previously described.
Routing of the PMF through the Catawba River system results in a maximum elevation of 580.0 ft (176.8 m) msl for Lake Wylie, as shown in Figure 2-38. The PMF is also routed through the system assuming the remote possibility that none of the hydroelectric generating units of Wylie and upstream stations are operating. Under this condition the peak elevation of Lake Wylie for the PMF is 581.1 ft (177.1 m) as shown in Figure 2-39. Both Figure 2-38 and Figure 2-39 are based on initial loss conditions as described in Section 2.4.3.2.
The South Fork Catawba River, with a drainage area of 660 sq mi (1709 km2), enters Lake Wylie near its northern end about 7 mi from the Catawba site. The maximum flood level in Lake Wylie due to the PMP is experienced when the storm is located on the Lake Wylie drainage area (the drainage area between Mountain Island and Wylie Dams is 1160 sq mi (3003 km).
Therefore, the effect of a local PMP on the South Fork drainage area is not materially different to that of the PMP on the larger Lake Wylie drainage area.
A flood inflow hydrograph developed for PMP occurring locally over the Big Allison Creek drainage area is also considered to coincide with Standard Project Flood over the Lake Wylie drainage area along with a seismic failure of Mountain Island Dam. Such a highly unlikely combination of events, with the most critical pyramiding of inflow causes a flood elevation of 578.1 ft (176.2 m) msl in Lake Wylie, lower than the surcharge caused by PMF.
2.4.3.6 Coincident Wind Wave Activity Procedures used in evaluating wind set-up and waves are described in Corps of Engineers Shore Protection Manual. The calculation is performed assuming deep water waves, a lake surface elevation of 591.8 ft (180.4 m) msl and a 40 mph (64 km/hr) overland wind. The run-up is calculated for three locations, the plant yard at the end of the intake canal, the discharge structure, and the SNSWP dam. Figure 2-40 shows the effective fetch. The results are presented in Table 2-63 and show that there is no overtopping for the above conditions.
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 18 (09 OCT 2016)
Assuming a static water elevation of 592.2 ft (180.5 m) msl, corresponding to the SPF and failure of Cowans Ford Dam, and using the same fetch, there would be minor wave run-up of 0.4 ft (0.12 m) on the plant yard, below openings for Class I structures.
2.4.3.7 Regulatory Guide 1.59 Refer to Section 1.7 for the discussion of Regulatory Guide 1.59.
2.4.4 Potential Dam Failures 2.4.4.1 Dam Failure Permutations Seven reservoirs on the Catawba River from the headwaters to and including Wylie Dam influence conditions at Catawba Nuclear Station. The reservoirs and a brief description of each are presented in Table 2-51. Additional information on drainage areas above the reservoirs and a description of the structures, ownership, date first operated, seismic design criteria and spillway design criteria is also presented.
There are no additional dams proposed for the Catawba River that would affect the Catawba Nuclear Station.
All of the dams above Wylie are composed of part earth embankment and part concrete gravity type structures. In the event of a seismic condition of sufficient magnitude as to cause structural failure to the dams, it is assumed that where the structure is composed of both earth embankment and concrete sections, the weaker portion is the earth embankment and its failure analyzed. The discharge from an abrupt earth embankment breach is computed from the equation:
5.1 b
b 28
.0 2
/
D W
K g
29
.0 max Q
1
where:
Q max
= the peak discharge at the dam in cfs, g
= the acceleration of gravity, Wb
= the width of breach in feet, Db
= the depth of water above the breach in feet, K
Db Yo Wb Wd
Wd
= the width of the dam in feet, Yo
= the depth of water before the breach in feet.
Records show that breaches in earth dams caused by the Chilean earthquakes of March 26, 1965 (Reference 7) varied in length from 45 m (148 ft) to 100 m (328 ft). Therefore, the initial breach in an earth embankment is assumed to be 300 ft (91 m) in length. In addition, the erosion rate of the exposed embankment face is taken as 1.4 ft (0.43 m) per minute. This rate of erosion is in agreement with the test results obtained from the test of a full size fuseable plug (Reference 8). All of the earth embankments are constructed on bedrock surface or on a shallow overburden; it is therefore assumed that shortly after an embankment breached, the invert would scour to bedrock. Seismic failure of an earth embankment is assumed to take place in the section having the greatest height. All of the Catawba side slopes are gently sloping and
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 19 there is little danger in a seismic shock causing a landslide into the reservoir pool causing a flood wave.
The seismic failure for each dam is timed to coincide with the Standard Project Storm centered over its drainage area. At the hour which the reservoir reaches its maximum level, it is assumed that seismic failure of the dam occurs. The seismic breach and resulting flood wave discharge is assumed to conform with that already described. As a measure of conservatism for the Bridgewater Dam failure analysis, no attenuation of discharge is considered in the channel flow of 18 miles between the Bridgewater and Rhodhiss Dams. A time lag of one hour is assumed for this reach. This travel time agrees with flood wave travel times of actual dam failures, particularly the failure of the South Fork Dam near Johnstown, Pennsylvania (Reference 9). The flood routing is computed for hourly intervals by means of a flood routing program with the procedure described in Section 2.4.3.
The results of these computations showing maximum lake elevations at each of the reservoirs are listed in Table 2-64.
For the simultaneous failure of Cowan's Ford Dam and the SPF, several storm locations are evaluated to find the highest flood level. This occurs when the storm is centered over the Wateree drainage basin, as shown in Table 2-65. The flood elevation resulting from the failure of Cowan's Ford Dam and the SPF over the Wateree drainage area is also analyzed for the special case where the control elevation for Lake Norman (Cowan's Ford) is set at elevation 761 ft (232.0 m) msl. This is a 1 ft (0.3 m) rise over the 760 ft (231.6 m) mls elevation shown in Table 2-59 and used in all other flood analyses. The results from this case are presented in Table 2-65, and the flood inflows, outflows and water elevations for Lake Wylie are presented in Figure 2-41.
The critical flood hydrograph associated with an upstream dam breach is developed for the flood resulting from the PMP occurring locally over the Big Allison Creek drainage area, combined with the SPF over the Lake Wylie drainage area and a seismic failure of Mountain Island Dam. Such a highly unlikely combination of events, with the most critical pyramiding of inflow will cause a flood elevation of 578.1 ft (176.2 m) msl in Lake Wylie, which is lower than the surcharge caused by the PMP on the larger Lake Wylie drainage area.
Multiple failures of two or more contiguous dams from a single seismic event are highly unlikely.
Such an event will not result in a higher pond elevation at the site than due to possible failure of only one of the dams. The minimum distance between two upstream dams on the Catawba River is 10 mi (16 km). A multiple failure of the two nearest dams results in two smaller flood waves passing through the site in quick succession. By the time the effect of failure of the upper dam reaches the site, peak flow due to the breach in the lower dam is already past the site. The duration of flood at the site is longer in case of a simultaneous failure, but the peak flood elevation is not higher than those given in Table 2-64.
Similarly, the effect due to failure of more than two dams simultaneously (all the more unlikely to happen) does not create more severe conditions at the site.
Assuming the failure of the earth embankment of Wylie Dam, the normal source of cooling water is lost above elevation 550.0 ft (167.6 m) msl. This is the elevation of the top of rock for that section of the dam. If this occurs, the plant would have to shut down using the Standby Nuclear Service Water Pond as a cooling water source. The SNSW Pond dam, described in Section 2.5, is designed to withstand sudden drawdown on the downstream face.
2.4.4.2 Unsteady Flow Analysis of Potential Dam Failures See Section 2.4.4.1.
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 20 (09 OCT 2016) 2.4.4.3 Water Level at Plant Site Maximum water level elevation at the station occurs with the simultaneous failure of Cowans Ford Dam, the SPF over the Wateree drainage area, and control elevation for Lake Norman set at elevation 761 ft (232.0 m) msl. Several storm locations are used to find the highest flood level as shown in Table 2-65. The maximum static water surface elevation is 592.4 ft (180.6 m) ms1.
The maximum water surface elevations when combined with a 40 mph (64 kph) wind are shown in Table 2-63.
2.4.5 Probable Maximum Surge and Seiche Flooding 2.4.5.1 Probable Maximum Winds and Associated Meteorological Parameters The Probable Maximum Hurricane, defined by the Weather Bureau report HUR 7-97, is used to determine the probable maximum meteorological winds at the site. Two hurricane tracks are considered possible of occurring and are included in the study to maximize the surge and wave effects at the plant site.
Case A assumes the classic hurricane track wherein the storm center moves shoreward in a northwesterly direction, crosses the coastline near Savannah, Georgia, and then gradually recurves toward the northeast to pass to the east of Lake Wylie. This track is hypothesized so as to place the radius of maximum wind coincident with the effective fetch at the plant site.
Case B is patterned after the track of Hurricane "Ginger" of the 1971 season. This storm follows a generally northwesterly course from the ocean, and goes ashore on the North Carolina coast between Wilmington and Cape Hatteras. It then follows a westerly course to a point between Charlotte and Raleigh where it curves northeasterwardly into Virginia. This track is also modified so as to put the radius of maximum wind coincident with the effective fetch at the plant.
The maximum wind speeds calculated are 101.5 mph for Hurricane A and 116.0 mph for Hurricane B.
The rise in water level due to the low barometric pressure that accompanies the eye of the hurricane is analyzed for both Cases A and B, and these values are:
Case Central Pressure Index Ht. From Diff. Pressure A
26.78" 2.25' B
26.89" 2.18' 2.4.5.2 Surge and Seiche Water Levels Surge and seiche water levels generated by various wind mechanisms (as described in Weather Bureau Report HUR 7-97 and "Shore Protection Manual", Department of the Army Corps of Engineers, 1973) are described below:
The wind tide, wave height, and runup as well as the total water level height for the significant wave and the one percent wave for the two hurricane cases are:
Wind Velocity Wind Tide Wave Ht.
Runup Total Ht.1 Hurricane A 101.5 mph 0.33' 7.4' 4.0' 6.6'a 0.33' 12.4' 5.5' 8.1'b
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 21 Wind Velocity Wind Tide Wave Ht.
Runup Total Ht.1 Hurricane B 116.0 mph 0.44' 8.6' 5.1' 7.7'a 0.44' 14.4' 5.8' 8.4'b Notes:
- 1. Total Heights for Hurricanes A and B include the component due to differential pressures.
- a. = significant wave: the average height of the one third highest waves of a given wave group
- b. = one percent wave:the average of the highest one percent of all waves in the group. 1.67 times the height of the significant wave The wind tide, wave height, and runup are also calculated for a 40 mph wind as previously detailed.
As detailed in the Catawba ER-OLS Section 2.4.1.1, the 100 year recurrence interval high water for Lake Wylie would be impractical to determine because of the six upstream dams (1.5 million acre-feet of storage). It should be noted that the hypothetical 1 in 100 year flood could be passed by the Wylie Dam with no increase in water surface elevation.
2.4.5.3 Wave Action For discussion of wave action resulting from hurricanes, refer to Section 2.4.5.2. Refer to Section 2.4.3.6 for discussion and method of calculation of waves generated by a 40 mph wind.
2.4.5.4 Resonance The configuration of the reservoir is such that resonance is not considered to be applicable.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.5.5 Protective Structures The slope of the yard fill at the north end of the plant yard is protected from wave action by a 2.0 ft (0.61 m) thick riprap underlain by a 12 in. (30 cm) thick subgrade of filter material. This protection is more than adequate for the maximum waves to which the slope is exposed. The SNSWP dam is also protected from wave action by 4 ft (1.2 m) thick riprap as shown on Figure 2-141.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.6 Probable Maximum Tsunami Flooding The Catawba Nuclear Station is located more than 150 miles (241 km) from the nearest coastal area and at an elevation of 594 ft (181.1 m) ms1. Therefore, tsunami effects are not a consideration.
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 22 (09 OCT 2016) 2.4.7 Ice Effects The climate of the site vicinity is temperate. Minimum monthly mean water temperature for Lake Wylie is in the low 40's. Therefore, significant ice formation is not considered a problem. The effect of ice formation on the drainage system is discussed in Section 2.4.2.3.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.8 Cooling Water Canals and Reservoirs The Standby Nuclear Service Water Pond (SNSWP) is a Nuclear Safety Related impoundment constructed by placing a dam across a small cove of Lake Wylie, as shown in Figure 2-20.
Table 2-66 presents pertinent information about the pond and dam. The SNSWP is the only new reservoir or canal built for the station.
The Probable Maximum Precipitation centered critically over the SNSWP is developed using the procedure outlined in the Bureau of Reclamation publication titled, Design of Small Dams (Reference 5). The PMP for a ten square mile area (26 km) and a six hour duration, taken from Figure 15 on page 48, is 30.1 in. (76.5 cm). This concurs with chart 50 in Technical Paper No.
40, Rainfall Frequency Atlas of the United States (Reference 10), published by the Weather Bureau.
The drainage area for the SNSWP is 410 Ac (1.66 km2). Though no distinction is made between a ten square mile and a point rainfall due to the unlikelihood of a perfect strike of a storm center on the drainage area in question (Reference 5), the graph lines in Figure 2, zone 6 are extended to obtain conservative values for a 1.0 sq mi (2.6 km2) drainage area (Reference 5).
Using the zone C curves in Figure 18 (Reference 5) the PMP is divided into its hourly increments and these increments are arranged in a sequence which produces a flood greater than the one which is produced by assuming that the maximum rainfall occurs in the first hour (Reference 5). The hydrologic soil-cover complex number of the watershed is determined by using Appendix A (Reference 5). The direct runoff is estimated using Figure A-4 (Reference 5)and the hydrologic soil-cover complex number. The incremental runoffs are 2.4 (6.1), 2.7 (6.9), 3.3 (8.4), 14.9 (37.8), 4.6 (11.7) and 2.5 in. (6.4 cm) per hour for hours one through six respectively. Determination of the runoff for a period of longer duration is not considered necessary because the first six hours provides the most intense flooding conditions.
The time of concentration is calculated by the formula given on Figure 30 (Reference 5) and is 40 minutes. The peak rate in cfs for 1.00 in. (2.54cm) of runoff is calculated using the equations of Figure 29 (Reference 5). This value is adjusted for the incremental runoff.
Inflows for hours one through six, the most intense periods, are 70 (2.0), 450 (12.7), 750 (21.2),
3650 (103.3), 2800 (74.2), and 900 cfs (25.5 m3/s), respectively. Figure 2-42 gives the PMF inflow, outflow, and water surface elevation hydrographs for the SNSWP. Figure 9-54 gives the areavolume curves for the SNSWP.
Discharge from the pond is through a 60 in. (152 cm) diameter pipe, as shown on Figure 2-143.
The pipe is set in a concrete wall with the invert elevation at 571.0 ft (174 m) msl. A weir, top elevation 574.0 ft. (174.96 m) msl, is installed at the discharge pipe headwall to control the water level in the pond. The pipe is approximately 560 ft (171 m) long extending across a small peninsula and discharging into Lake Wylie. Discharge capability of the pipe is calculated using methods described in the Handbook of Hydraulics (Reference 11).
A graphical routing of the flood results in a maximum reservoir elevation of 583.5 ft (177.9 m) during the sixth hour of rainfall. This excess inflow into the reservoir is 620 Acft (7.6 x 105m3) with a surcharge on the pond of 9.5 ft (2.9 m).
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 23 The maximum wave height is calculated using formulae developed by Saville, McLendon and Cochran as outlined in Water and Wastewater Engineering (Reference 12). The maximum wind velocity for the area, recorded at Charlotte, North Carolina, is 57 mph (92 kph) for a one minute average from the northeast.
Translating the velocity into its component for the direction of maximum fetch (east), the maximum velocity is 40 mph (64 kph) and the fetch length is 0.42 mi (0.68 km). The significant wave height is calculated to be 1.3 ft (0.4 m). A ratio of wave height to runup (
s r H
/
H
),
calculated by the methods previously described, is 0.8. The maximum height of the wave on the slope including the runup is 1.0 ft (0.3 m).
Thus, the maximum flood elevation including wave and runup becomes 584.5 ft msl (178.2 m).
The crest elevation is set at 595.0 ft (181.4 m) to safely accommodate the highest flood surcharge downstream for Lake Wylie. Therefore, there is an 10.5 ft (3.2 m) freeboard on the upstream side to contain the PMF surcharge, wave height, and wave runup.
The analysis of a PMF peak on Lake Wylie occurring coincident with a PMF peak on the SNSWP was not evaluated because it was not considered credible that the two PMF peaks would coincide.
The SNSWP dam is protected from wave action by riprap as shown in Figure 2-140.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED The total loss of volume in the SNSW pond due to sedimentation during the 40 year plant life is calculated by use of the universal soil loss equation (Eq (1) Reference 19) to be 5.2 ac-ft.
The resulting distribution of this sediment within the pond is dependent on several factors. An analysis of the settling characteristics of the expected sediment load indicates that about two-thirds of the sediment will settle in delta type formations. This was determined by employing the relationship L = v/s x d (Reference 19)
Where:
L
= length required for particle to settle d
= depth of settling zone v
= forward velocity of flow s
= fall velocity of particle The forward velocity in the pond is taken conservatively as 0.1 ft/sec, and the depth of settling zone is taken to be 20 ft. The fall velocity of the sediment is found from Stoke's Law:
)
(
D 18 1
v s
2
Where:
D
= diameter of particle
= viscosity of water (2.359 x 10-5 lb.5/ft2) s
= density of sediment (165 lb/ft3)
= density of water (62.4 lb/ft3)
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 24 (09 OCT 2016)
Using the grain size analysis for site soils, it is calculated that approximated two-thirds of the sediment entering the pond would settle rapidly (L < 310 ft). This would lead logically to the formation of delta deposits in the head water areas of the pond. This conclusion is substantiated by data from existing resevoirs, presented in Reference 20. The one-third or approximately 1.7 ac-ft of sediment which does not deposit in the upper regions of the pond can be assumed to distribute uniformly in the form of bottom set deposits in the vicinity of the dam and SNSW intake structure. This amount of deposition should have no adverse impact on the operation of the SNSW intake structure. Soundings will be taken around the SNSW intake structure prior to fuel loading and at 5 year intervals thereafter to assure that sediment deposits will not adversely affect the operation of the NSW System. In addition, a license renewal commitment was made to perform a survey of the SNSW pond as part of the SNSW Pond Volume Monitoring Program on a 3-year frequency to ensure the SNSW pond volume for the period of extended operaion.
With the exception of approximately 30 acres, the entire 450 acre drainage basin of the SNSW pond is owned by Duke Power Company.
The 30 acre tract of private property is located along the western edge of the drainage basin, over 3,000 feet from the pond. The effect on sediment load delivered to the SNSW Pond due to disturbance on this property would be minimal because of its relatively flat terrain and its remote location with respect to the pond.
Operation of the pond as the ultimate heat sink is described in Section 9.2.5.
2.4.9 Channel Diversions The source of cooling water for Catawba Nuclear Station is Lake Wylie. There are seven reservoirs on the Catawba River upstream from Wylie Dam, all of which are owned and operated by Duke Power Company. Within limits set by the operating licenses of these dams and dam leakage, the minimum discharge to Lake Wylie is controlled by Duke Power. No present means exist to divert or reroute other than minor amounts of water used for municipal supply.
In the event of the loss of Lake Wylie, the Catawba Station could be safely shut down using the Standby Nuclear Service Water Pond.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.10 Flooding Protection Requirements The SNSWP dam top elevation of 595.0 ft (181.4 m) msl, is adequate for the maximum flood level of 592.4 ft (180.6 m) msl. It is designed to withstand wave wash due to a coincidental wind wave runup by an overland wind of 40 mi (64 km) per hour.
The north end of the plant yard at the end of the water intake cove is protected from wave action by a riprap cover. The rip rap is class IA, a 2 ft. (0.6m) thick layer of rock ranging in size from 3 in. (7.6 cm) to 24 in. (61 cm) in. diameter with greater than 50 percent larger than 18 in. (46 cm);
this is underlain with a 12 in. (30 cm) thick filter. See Section 2.5.6.4.3.
2.4.11 Low Water Considerations HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 25 2.4.11.1 Low Flow in Streams The primary water source for Catawba Nuclear Station is Lake Wylie. The lake inflow is from Mountain Island Lake (Catawba River), the South Fork Catawba River, and tributary creeks which contribute approximately 50, 25, and 25 percent, respectively, of the total flow. The total drainage area for Lake Wylie is 3020 square miles of which only 1860 square miles is directly from the Catawba River. The FPC license for Catawba-Wateree Project No. 2232 requires a minimum average daily release of 314 cfs from Mountain Island Dam and 411 cfs from Wylie Dam. To calculate the low flow entering Lake Wylie, it is assumed the Catawba River contributes the 314 cfs required release from Mountain Island Dam and the remainder of the basin contributes a flow equivalent to the 7Q10 yield per square mile for the South Fork Catawba River. The total flow entering the lake based on these assumptions is about 516 cfs as shown on Table 2-67.
Lake Wylie has 107,200 acre-feet of storage between full pond elevation of 569.4 ft msl and the maximum drawdown elevation of 559.4 ft msl. Assuming no inflow to Lake Wylie while maintaining the required 411 cfs discharge from the lake and forced evaporative losses resulting from Allen Steam Station and Catawba Nuclear Station, plus natural evaporation, it takes 100 days for the lake to drop from full pond to maximum drawdown as shown on Table 2-67.
Non safety-related water supplies will be adequate during a 1 in 100 year drought because of the upstream control on the Catawba River provided by six dams with approximately 1.5 million acre-feet of storage (refer to Catawba ER-OLS, Section 2.4.1.1).
2.4.11.2 Low Water Resulting from Surges, Seiches, or Tsunami The effects of seiche, tsunami, and ice are not applicable to the site.
2.4.11.3 Historical Low Water The minimum water elevations for Lake Wylie are shown in Table 2-68 and are discussed in Catawba ER-OLS, Section 2.4.1.1.
2.4.11.4 Future Control The minimum flow downstream of Mountain Island Dam into Lake Wylie is 314 cfs (8.9 m3/s) as stated in Subdivision 2.4.11.1. No future projects affecting the minimum flow are proposed on the Catawba River or its tributaries between Mountain Island and Wylie Dams.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.11.5 Plant Requirements During two unit operation at full load, a nuclear service water flow rate of approximately 12,000 gpm (45,425 l/min) is required to maintain pump miniflow requirements and supply the safety and non safety related components with cooling water. Table 9-3 shows the nominal nuclear service water flow rates to essential and non essential header components during various modes of operation. The NSW Pump Structure has two pumps per unit, each with a rated capacity of 20,900 gpm (79,123 l/min). The sump invert elevation is 545.8 ft (166 m) msl. The sump configuration is shown in Figure 2-43. The minimum lake surface elevation required for operation of the pumps is 559.4 ft (169.3 m) msl.
Condenser cooling water is supplied by closed cycle wet mechanical draft cooling towers.
Makeup water for cooling tower evaporation and drift is from Lake Wylie. With two units at 100 percent load, makeup requirements average of 89 cfs (2.6 m3/s). The makeup water intake
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 26 (09 OCT 2016) structure bottom sill is at elevation 542 ft (165.2 m) msl as shown in Figure 2-44. The minimum lake surface elevation required for operation of the pumps is 559.4 ft (170.5 m) msl. It is impractical to consider whether or not the cooling water pumps will be able to supply station water during a one in 100 year drought because of the upstream control on the Catawba River provided by the largest six dams (with approximately 1.5 million acre - feet of storage). Refer to Catawba ER - OLS, Section 2.4.
Conventional service water (also referred to as Low Pressure Service Water) provides make up to the condenser cooling water system, and is discharged to the lake after once-through cooling.
The conventional service water flowrate is approximately 42,171 millions of gallons per day (159,700 m3/day). Refer to the Catawba NPDES Permit issued 4/30/01.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.11.6 Heat Sink Dependability Requirements Under normal operating conditions nuclear service water is supplied from Lake Wylie. An alternate source is the Standby Nuclear Service Water Pond which is used if the normal supply is not obtainable. The nuclear service water system is described in Section 9.2.1. The design basis hydrometeorology, water losses, and temperature analysis are described in Section 9.2.5.
The nuclear service water pond and dam are described in Section 2.4.8 and Section 2.5.6.
2.4.12 Dispersion, Dilution and Travel Times of Accidental Releases of Liquid Effluents in Surface Waters In the event of an accidental liquid radwaste release from the Catawba Nuclear Station, the effluent could enter Lake Wylie. A discussion of this accident is presented in Section 15.7. The point of entry would be approximately 5.0 mi (8.0 km) uplake from the Wylie dam and hydro station. The nearest surface water intake that would be affected by such an accidental release is the Rock Hill municipal water intake located on a cove of Lake Wylie just south of the Wylie hydro forebay, approximately 5.3 mi (8.5 km) downlake from the discharge structure at Catawba Nuclear Station. Surface water intake locations are presented in Table 2-52. An analysis of a postulated ground water accidental release is discussed in 2.4.13.3.
Lake Wylie has a full pond volume of 281,900 Ac-ft (3.48 x 108m3) and an average flow through of 4212 cfs (119.3 x m3/s) which provides a large volume of water for dilution. Initial dilution of the postulated accidental release will occur prior to entering Lake Wylie as it will mix with the 120 cfs (2.94 x 105m3/day) normal service water effluent. Further dilution will occur with lake water as the spill travels toward the Wylie dam and hydro station; and the Rock Hill municipal water intake. With all four units operating at best efficiency point, Wylie hydro can pass 12,500 cfs (354 m3/s).
Travel times to the downlake water intake structure are dictated by wind induced surface currents (Reference 13). A critically high 12 to 15 hr duration wind speed of 30 mph was assumed in the direction of the dam. Based on these assumptions, the travel time to the downlake intake was approximately 0.6 days. A more realistic critical travel time of 3.0 days was computed assuming the average 24 hr duration wind speed of 7.5 mph. Documentation for the time of travel calculations can be found in the following Catawba Nuclear Station calc: CNC-1105.05-00-0001, Dispersion of Accidental Liquid Release to Lake Wylie.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 27 2.4.13 Groundwater 2.4.13.1 Description And On Site Use HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.13.1.1 Regional Groundwater Conditions The Catawba site lies within a groundwater region which is part of the Piedmont Groundwater Province. Groundwater recharge in this area is derived entirely from infiltration of local precipitation. The surface materials in many locations are relatively impermeable with the result that only 10 in. (25 cm) to 15 in. (38 cm) of the average 45 in. (114 cm) of annual precipitation percolate to the water table (Reference 14). Groundwater is contained in the pores that occur in the weathered material (residual soil-saprolite) above the relatively unweathered rock and in the fractures in the igneous and metamorphic rock. The depth to the water table depends on climate, topography, rock type and rock weathering. The water table varies from ground surface elevation in valleys to more than 100 ft (30 m) below the surface on sharply rising hills.
Shallow dug wells are supplied from surface deposits or from the upper decomposed parts of the bedrock. Many drilled wells of moderate depth are supplied from joints in the crystalline rocks. Groundwater rarely occurs at depths more than 300 ft (91 m) below ground surface, and usually occurs in significant quantities at depths less than 150 ft (46 m). The water quality is excellent, generally low in minerals, except iron. The quantities of water available are generally small. There are numerous localized perched water tables, restrained by less pervious zones of rock or saprolite, as well as very localized artesian aquifers. These artesian aquifers are generally related to isolated but well developed rock cracks, generally in the higher ground.
2.4.13.1.2 Station Groundwater Use All process water required for the operation of the station is obtained from the SNSW Pond or Lake Wylie. Potable water for the station is obtained from the York County, SC water system.
Several wells are maintained and used by Duke Power at or near the site. One well, formerly used for potable water (at the training center) is now available for irrigation of shrubbery and grass. Two other wells, one at the conference center on the other side of Concord Road from the station and the other at the recreation area (Catawba Park), are used for potable water.
One other well located at the firing range is used only for flushing and cleaning purposes.
2.4.13.2 Sources HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.13.2.1 Groundwater Users A survey, made of the wells which provide domestic water supply in the general site area (approximately 1 mi (1.6 km) radius), yielded the locations of 12 wells and one spring as shown on Figure 2-47 and the available information given in Table 2-69. Interviews with local well drillers, conducted to augment the information on domestic wells in the area, indicate that domestic wells are normally less than 100 ft (30 m) deep, 6 in. (15 cm) or less in diameter, and have a flow rate in the range 3 (11) to 150 gallons per minute (568 l/m). Table 2-70 and Table 2-71 list regional groundwater users. Figure 2-48 and Figure 2-49 show their locations.
Present groundwater use in the area is limited to domestic use. Due to the relatively low yield of wells in the area and the proximity of Lake Wylie, the character of future groundwater use is not
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 28 (09 OCT 2016) expected to differ significantly from that of present use. Future groundwater uses in the region are not seen as having any significant effect upon groundwater conditions at the site.
2.4.13.2.2 Program of Investigation To evaluate groundwater conditions at the site, observation wells and water level recorders are installed; tests are performed to determine hydrologic properties of the water-bearing materials; water levels are periodically measured in observations wells and core borings. A water level recorder is used to monitor groundwater changes and continuous records of water levels are obtained. Water level recorder information is presented in Section 2.4.13.2.4 and a hydrograph is shown on Figure 2-53. Water samples from selected core borings and private wells are obtained and analyzed.
Packer permeability tests are performed in borings to determine permeabilities of the rocks underlying the plant site. Constant head field permeability tests are performed in selected borings to determine horizontal permeabilities of saprolite soils. Variable head laboratory permeability tests are performed on selected undisturbed samples to determine vertical permeabilities of saprolite soils. Two constant discharge pumping tests are performed at the site to assess the permeability and storage coefficient (effective porosity) of the water table aquifer over a relatively large area. A total porosity of 0.3 was assumed for the residual site materials based on tests for unit weight, moisture content, and specific gravity as described in Section 2.5.4.
2.4.13.2.3 General Geologic Conditions of the Site The oldest and most abundant rock type at the site is adamellite, an intrusive igneous rock. This rock type is fairly uniform in texture and mineralogy, as determined from observing rock outcrops in the field and by studying petro graphic thin sections of rock core samples. The adamellite is medium grained and is composed primarily of feldspars with lesser amounts of quartz and biotite.
Mafic dikes constitute a subordinate rock type and are discontinuous and irregular across the site. The mafic dikes consist of fine-grained plagioclase feldspar, hornblende and biotite. Both the mafic dikes and the adamellite at the site have undergone metamorphism which imparted a moderate foliation to the mafic dikes and a faint foliation to the adamellite. Additional discussion of site geologic conditions is provided in Section 2.5.1.2.2.
The rock at the site is covered by a mantle of weathered material. With increasing depth, these materials range from a thin near-surface veneer of moderately-plastic, fine-grained soil underlain by saprolite soils of varying degrees of weathering. The saprolite generally grades with depth into weathered rock and eventually to continuous rock. The thickness of saprolite at the site is variable and greatest beneath the higher elevations of the original ground surface. A general maximum thickness of saprolite is on the order of 100 ft (30 m). The saprolite generally consists of silty sand weathered from the adamellite.
2.4.13.2.4 Groundwater Levels In the site area, groundwater is generally encountered under water table conditions in weathered rocks and residual soils that overlie less weathered rocks. Observations of groundwater elevations, made at about 60 locations in the immediate vicinity of the site, are used to make the contour map of the preconstruction water table, as shown on Figure 2-50.
This map shows that the preconstruction elevation of the groundwater varies from about 10 (3) to 40 ft (12 m) below natural ground surface near the location of the Reactors and that it
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 29 approaches the surface elevation of Lake Wylie near the lake shore. Thus, groundwater movement is from the plant area toward the lake coves which cut into the peninsula to the north and to the south of the site. A cross section through the site illustrates the relation between the topography, groundwater elevation, lake elevation, rock elevation and plant elevation (see Figure 2-51).
Groundwater recharge at the site is controlled by local precipitation and infiltration. Natural discharge is through evapotranspiration by plants and by discharge to Lake Wylie. Experience in the region is that the groundwater level normally declines during the late spring, summer, and early fall months as a result of evapotranspiration by plants, and in the fall when rainfall is low.
The groundwater level rises in late fall, winter, and early spring when the evaporation potential is reduced.
The USGS monitors the groundwater level in a drilled well across Lake Wylie from the plant.
This well is 700 ft (213 m) in depth with the upper 50 ft (50 m) cased. The surface elevation is about 600 ft (183 m) msl. The well location is shown on Figure 2-47. Figure 2-52 shows the monthly groundwater hydrograph from October, 1973 to September, 1977 for the well. The annual groundwater fluctuations generally follow the winter-spring high and summerfall low cycle. The maximum groundwater change since the well has been monitored (1971) is 6.64 ft (2.02 m).
Groundwater levels in core borings at the site varied several feet during the period of subsurface exploration. The amount of groundwater level variation in the borings is typified by the water level recorder results given below. A water level recorder was monitored in boring A-33 and then moved to boring A-62. The groundwater level fluctuation at boring A-33 was about 2 ft (0.6 m) as is shown on Figure 2-53, along with the lake level and daily rainfall data. The water level generally rose during the period monitored beginning in the late fall and ending in early spring, correlating well with regional experience. The fluctuation at boring A-62 was about 2.5 ft (0.76 m) over a period of 8 months.
Regionally, there is characteristic overall change in the water table which begins to decline in late winter and spring owing to the increasing amount of evaporation and transpiration of plants.
In the late summer, fall, or into early winter, when much of the vegetation has become dormant, the local precipitation begins to make-up the summertime soil-moisture deficiency and the water table begins to rise as recharge occurs. Thus, the variation in water levels during the subsurface exploration correlates with this regional experience for seasonal trends.
2.4.13.2.5 Permeability Tests Various types of tests are performed to evaluate the hydraulic characteristics of the subsurface materials at the site. All tests measure either the permeability or transmissivity of the aquifer, and the pumping tests also provide a measure of the aquifer storage. Permeability is defined as the amount of water which would flow across a unit area of aquifer in a unit time under a unit gradient. Permeability is controlled by the distribution of fractures in the bedrock and by the size and distribution of the pores in the saprolite and soil above bedrock. The transmissivity of an aquifer is equal to the permeability of the aquifer multiplied by its thickness. The coefficient of storage of an aquifer is the volume of water released from storage per unit surface area of the aquifer per unit decline of head.
The tests performed at the site have individual purposes and merits; when the results are collectively examined, a meaningful evaluation can be made of the hydrologic character of the subsurface conditions. Packer field permeability tests are useful in estimating rock permeability within short intervals or borings; however, results are not always indicative of permeability values over a large area because the test stresses only a small portion of the aquifer, and thus
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 30 (09 OCT 2016) measures localized values of permeability in the vicinity of the borings. Constant head field permeability tests are useful in determination of permeability in soil-saprolite and weathered rocks, but evaluation of results may also be limited to relatively small areas or influenced by localized conditions. Laboratory variable-head permeability tests provide a quantitative determination of the coefficient of permeability of onsite soils; however, these tests also measure only small samples of the materials.
Large-scale field pumping tests are performed at the site to provide quantitative determinations of two principal hydrologic properties of the aquifer, the capacity to transmit and to store water.
Constant discharge pumping tests are utilized for evaluation of hydrologic conditions over a relatively large area at locations where the weathering profile is similar to that in the powerhouse area.
2.4.13.2.5.1 Packer Permeability Tests Test results from 62 packer permeability tests performed in 13 borings are presented in Table 2-
- 72. These tests measure the permeability of rocks underlying the site. Figure 2-54 shows the arrangement of the field test equipment and a brief description of the procedure used in determining the rock permeability test results.
The effective permeability of the rock mass is low. Test results show the permeability to range from 0.0 to less than 500 feet (152 m) per year, with an exception being observed in boring A-54 where a permeability of 1761 ft (537m) per year is measured. This higher permeability is measured in a zone of fractured rock. At this location there was a loss of drilling fluid and a section of core was lost during drilling. In addition a number of seams of quartz pegmatite intersect the boring at the depth at which the higher permeability values are measured. In general the higher values of permeability are associated with cored sections classified as moderately hard, jointed adamellite with water stained joints and weathered seams.
2.4.13.2.5.2 Constant-Head Permeability Tests Horizontal soil permeability is measured by field measurements in borings using sealed piezometers. Figure 2-55 shows the arrangement of the field test equipment along with a brief description of the procedure used in determining the soil permeability test results. Test results are from 8 borings as presented in Table 2-73. These test results indicate a very low permeability and are less than the packer tests in rock.
2.4.13.2.5.3 Variable-Head Permeability Tests Vertical soil permeability is measured by laboratory tests according to ASTM D2434 on undisturbed samples. The laboratory tests can be divided into two categories: (1) Tests made under conditions designed to simulate the overburden pressure present under field conditions, and (2) Tests made without overburden pressure. Test results from 11 borings are presented in Table 2-73.
The tests made under simulated field conditions yield permeabilities significantly lower than those tests for which overburden pressure is not applied. Also, the vertical permeabilities determined under simulated field conditions are significantly smaller than the horizontal permeabilities measured in the field. The higher values of soil permeability, which range up to 378 ft (115 m) per year, are measured in residual soils having the texture of a silty fine to medium sand. The lower values are measured in clayey fine to medium sandy silt and partially weathered rock and were less than 1.0 ft (0.3 m) per year. The higher values of soil permeability are less than the higher values measured by the packer tests in rock.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 31 2.4.13.2.6 Constant-Discharge Pumping Tests Constant-discharging pumping tests are conducted to determine average values of horizontal permeability, transmissivity, and storage coefficient of the aquifer. The locations and layout of two pumping tests are shown on Figure 2-56 and Figure 2-57. These sites have weathering profiles similar to that of the powerhouse area but are away from the construction activities. For these tests, the wells are drilled to a depth of 70 ft (21 m) and 4 in. (10 cm.) PVC pipe is installed. The pipe is slotted with 0.030 in. (0.076 cm) openings in the lower 65 ft (20 m) of the well. The annular space between the drilled well and the slotted pipe is filled with pea gravel.
Eight observation wells, installed at distances of 10 (3) to 60 ft (18 m) from the test well, are constructed of 2 in. PVC pipe to a depth of 70 ft (21 m), with the lower 65 ft (20 m) of pipe slotted. Table 2-74 and Table 2-75 show the results of the pumping tests. The cone of depression for these wells is shown in Figure 2-58 and Figure 2-59.
2.4.13.2.7 Groundwater Quality The chemical and bacteriological groundwater quality in the vicinity of the site is high and satisfactory for domestic use without treatment. Chemical and physical tests were conducted on water from 6 wells located around the site. The analysis shows the water to be low in mineral content and slightly acidic. The mineral content in these wells is as low or lower than average values found in the surrounding area. The results of the chemical and physical tests are shown in Table 2-76. The locations of the wells from which the samples were taken are shown on Figure 2-47. In addition, quality tests were made on samples taken from 4 borings on the site.
The locations of the borings are shown on Figure 2-78. The results of these tests are in the same range as those from the wells, except for those parameters which are affected by the drilling operation, such as hardness and turbidity.
2.4.13.3 Accident Effects 2.4.13.3.1 Contaminant Transport Model The movement and dilution of postulated, accidentally released, radioactive contaminants is a function of the hydrologic and hydraulic parameters of the groundwater system and a function of radioactive decay and ion exchange of the contaminants. A contaminant transport model incorporating these factors is used to evaluate the movements of radionuclides in the groundwater system between the point of postulated contamination and Lake Wylie.
2.4.13.3.2 Ion Exchange Capacity Standard methods of chemical analysis are used to determine the cation exchange capacity of the soil at the site. Several samples are selected from borings near the center of the site and tested for their ion exchange capacity. The results of these tests are presented in Table 2-77.
The distribution coefficient, Kd, provides a measure of the exchange characteristics of the soil. It has been shown (Reference 15) that the distribution coefficient depends on the concentration of the contaminant, the pH of the transporting solution, and on the presence of additional ions in the transporting solution. Comparision of the ion exchange capacity of the soil and the chemical characteristics of the groundwater at the proposed site with values obtained from laboratory test (see Prout) suggest a value of Kd for the site in the range 10-100 ml/g for strontium.
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 32 (09 OCT 2016) 2.4.13.3.3 Transport of Postulated Contaminants In general, flow of groundwater is normal to groundwater contours. The quantity of groundwater movement is controlled by the slope or gradient of the water table and the permeability of the area through which it moves. The velocity of flow is controlled by the gradient, the permeability, and the porosity. The shortest groundwater path between the site and Lake Wylie is by way of the coves in the lake shore located north and south of the site.
The groundwater velocity and travel time along the critical path is computed as follows:
i S
K V
y h
Where:
V
= Groundwater Velocity, Kh
= Horizontal Permeability, Sy
= Drainable Porosity, i
= Groundwater Gradient.
Based on the maximum measured rock permeability value of 1761 ft/yr (537 m/yr) from boring A-54, a minimum average drainable porosity of 0.05 from the pumping test and a groundwater gradient of 0.03, the groundwater velocity is computed to be about 1100 ft/yr (335 m/yr). The travel time is computed to be about one year by dividing the length of the assumed critical path (750 ft) (229 m) by the groundwater velocity. This time is half that reported in the PSAR because of the lower porosity value of 0.05 selected on the basis of the two pumping tests. This calculated travel time is quite conservative because of using the steeper groundwater gradient for the preconstruction conditions, not the gradient that will be established by permanently lowering the water table in the plant vicinity.
The rate of movement of radioactive contaminants depends on the composition of the waste, composition of the soil, and the rate of movement of groundwater. The radioactive contaminant will move less rapidly than the groundwater because it will be absorbed, to some degree, by soil particles. A relationship (Reference 16) has been developed which provides an estimate of the effect of ion absorption on the travel time of a radioactive contaminant. This relationship may be expressed as:
w d
c t
P K
)
P 1(
B 1
t
Where:
tc
= time of travel for contaminant, B
= bulk density (g/ml),
P
= porosity, Kd
= distribution coefficient (ml/g),
tw
= time of travel for groundwater.
A conservative travel time for strontium is estimated to be approximately
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 33
w c
t]
3.0 10 3.
1 925
.1 1[
t
w c
t]
45 1[
t
i.e., the conservative travel time for strontium is about 46 times longer than the travel time for water or approximately 46 years. If the larger value of the distribution coefficient is used, the travel time would be increased by a factor of 45l instead of 46 or about 450 years. These calculations are based on a value of density equal to l.925 g/ml (l20 lbs/ft3) and a porosity of 0.3.
Strontium is used in these calculations because it frequently represents the most critical potential nuclear contaminant.
The distribution coefficient for cesium, like that for strontium, varies with pH and the concentration of the isotope solution. At a molal concentration of 5 x 10-8, the soil tested by Prout had a distribution coefficient for strontium that varied from about 10 (pH=3) to a value of approximately 900 (pH=7). For the same concentration and range of pH, the distribution coefficient for cesium at low values of pH can be predicted to be considerably greater than that for strontium. As a result, the travel time for cesium will also be greater, possibly in the range of about two times to as much as twenty times longer than for strontium.
In determining the above travel times, the groundwater underdrain system is assumed to be not functioning with a resulting groundwater level at plant yard elevation. In the normal case where the groundwater underdrain system is functioning, any release of radioactive liquid into the groundwater would seep through the soil to the zoned wall filter and eventually flow to the Auxiliary Building groundwater sumps. The contaminated liquid would then be pumped to the storm drain system which discharges directly into Lake Wylie. A discussion of an accidental radioactive liquid release into Lake Wylie is presented in Section 2.4.12 and Section 15.7.
Since test borings were filled in or covered by earthwork and construction activities, and observation wells are capped, they provide no pathway for the release of radioactive liquid into the groundwater. The test borings and observation wells are shown on Figure 2-78.
2.4.13.4 Monitoring or Safeguard Requirements Only those private wells outside the 2500 ft (762 m) radius exclusion area will remain during operation of the plant. Of the remaining local private wells, most are associated with cottages along Lake Wylie and are used primarily during the summer months. Volume of usage from these wells is small and is not predicted to increase greatly in the future. The station dewatering system does not affect groundwater levels more than a few hundred feet away from the station.
Groundwater does not move from the plant area toward any private wells that will remain in service after construction. Further, the topography and groundwater conditions are such that no reversal of groundwater gradients can be expected after station construction.
2.4.13.5 Design Basis for Subsurface Hydrostatic Loading As shown in Figure 2-50, the preconstruction groundwater elevations vary from about 10 ft. (3.0 m) to 40 ft (12.2 m) below natural ground surface near the location of the Reactor Buildings and approaches the surface elevation of Lake Wylie near the lake shore. Seasonal groundwater fluctuations are discussed in Subdivision 2.4.13.2.4. These preconstruction groundwater levels, if left unchanged, would result in uplift and hydrostatic loads on the Auxiliary and Reactor Building substructures.
A permanent Category I groundwater drainage system (which fully complies with the NRC's Proposed Positions for Review and Acceptance of Underdrain Pressure Relieving Systems) is
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 34 (09 OCT 2016) installed as shown on Figure 2-60 and Figure 2-61 create and permanently maintain a normal groundwater level at or near the base of the foundation mat and basement walls, thus eliminating the uplift and hydrostatic forces. This groundwater drainage system consists of foundation underdrains and continuous exterior wall drains. There are some very low pits in the Reactor and Auxiliary Buildings for which hydrostatic and uplift loads are not relieved by this groundwater drainage system. All such low areas not relieved of this groundwater pressure are designed to withstand the resultant uplift and hydrostatic loads. Similarly, no underdrain channels exist directly below the exterior Doghouse structures and therefore these structures are designed for the resultant uplift forces. It is unlikely, however, that any uplift force can develop on the exterior Doghouse structures since there are underdrain systems at lower elevations in the adjoining structures.
The foundation underdrain consists of a grid of interconnected flow channels at the top of excavated rock or at the top of fill concrete on approximately 20 foot centers. This grid of flow channels provides a drainage path for the entire foundation areas of the Auxiliary and Reactor Buildings, except for the above mentioned exceptions which are designed for the subsequent uplift forces.
Each flow channel of the foundation underdrain has a minimum cross sectional area of 0.157 square feet (0.014 m2) and is constructed of lumber treated with the preservative pentachlorophenal in accordance with American Wood Preservers Association (AWPS)
Standard C2-73.
If the flow channels are placed on top of fill concrete, rather than directly on top of excavated rock, 2-5/8 in. (6.7 cm) diameter holes are drilled through the concrete into the rock, at a maximum spacing of eight feet on center. These drilled holes penetrate the rock a minimum depth of three feet, and are located directly below a flow channel. The purpose of these drilled holes is to provide a means for groundwater to flow from beneath the fill concrete into the foundation underdrain. The three-foot penetration of these drilled holes insures that interception of rock joints is obtained. The 3 ft (0.91 m) depth is determined from the dip and spacings of the joints in the rock. The jointing in the powerhouse area is demonstrated in construction photographs shown in Figure 2-65.
Combined with the foundation underdrain is a continuous exterior wall drain. This exterior wall drain consists of a 2 ft (0.61 m) minimum thickness zoned sand and stone filter placed vertically from the bottom of the excavation up to an elevation of 589+0 ft (179.5 m) msl (4.5 ft (1.37 m) below yard grade), and a 12 in (30.5 cm) diameter perforated corrugated metal pipe which is continuous horizontally around the exterior wall near the bottom of the zoned filter. The zoned filter consists of an outer layer of fine material which surrounds an inner layer of coarse material.
Both filter layers conduct seepage as well as surround the perforated drain pipe.
Figure 2-63 shows the gradation limits of the fine and coarse filter material derived from field gradation tests. Also shown on Figure 2-63 is the backfill design gradation. Figure 2-64 shows the gradation limits of the coarse filter material placed a minimum of 1'-0" around the perforated drain pipe. The gradations shown satisfy the following criteria:
- 1.
5
)
Base
(
85 D
)
Filter
(
15 D
- 2.
4
)
Base
(
15 D
)
Filter
(
15 D
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 35
- 3.
1 Diameter Hole Pipe Perforated
)
Filter
(
85 D
The following values were used in computing the filter criteria:
Maximum D15 (Coarse)
= 8.2mm (Figure 2-63)
Minimum D15 (Coarse)
= 1.4mm (Figure 2-63)
Minimum D85 (Coarse)
= 9.5mm (Figure 2-64)
Average D15 (Fine)
= 0.32mm (Figure 2-63)
Average D85 (Fine)
= 1.7mm (Figure 2-63)
D85 (Backfill)
= 0.63mm (Figure 2-63)
D15 (Backfill)
= 0.013mm (Figure 2-63)
Perforated Pipe Hole Diameter
= 3/8 in. = 9.5mm The following calculations verify that the filter criteria are satisfied:
Backfill and Fine Filter 5.0 mm 63 mm 32
.0
)
Backfill
(
85 D
)
Fine
(
15 D
6.
24 mm 013
.0 mm 32
.0
)
Backfill
(
15 D
)
Fine
(
15 D
Fine Filter and Coarse Filter:
8.4 mm 7.1 mm 2.8
)
Fine
(
85 D
)
Coarse
(
15 D
4.4 mm 32
.0 mm 4.1
)
Fine
(
15 D
)
Coarse
(
15 D
Coarse Filter and Perforated Pipe:
0.1 mm 5.9 mm 5.9 diameter Hole Pipe Perforated
)
Coarse
(
85 D
The fine and coarse filter materials are placed in layers not exceeding 12 in. (30.5 cm) in vertical thickness and compacted to a minimum of 80 percent relative density. Maximum relative density is determined as per ASTM D 2049. ASTM D 2049 has been withdrawn and replaced with ASTM D 4253 and ASTM D 4254. ASTM D 4253 and ASTM D 4254 may be used in lieu of ASTM D 2049. Field measurements of gradation and density are performed at a minimum frequency of once per day or once for every 400 tons (3.6 x 105 kg) of material placed,
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 36 (09 OCT 2016) whichever is greater. Materials that fail the gradation tests are rejected. Materials that fail the density tests are recompacted or removed.
Since the zoned filter wall drain system is confined by building walls and the compacted backfill (or rock excavation at the foundation level) the wall drain system will remain passive during earthquake as will the underdrain system. Since the top of the zoned wall filter is 41/2 feet below plant yard grade there is no credible flood that will affect the underdrain system.
Remedial work as described in the Selected Licensee Commitments would be initiated when the zoned wall filter is exposed to sources of water not considered in the design.
A similar underdrain system has been used successfully at Cowans Ford Hydroelectric Station since 1963.
The capacity of the drainage system is based on the calculated seepage flow rate into the Reactor/Auxiliary Building foundation area. The seepage flow rate is computed from the following equation (Eq. (6) Reference 18):
X 2
L E
L 2
)
he H
(
K q
2 2
Where:
q
=
Flow rate per unit width in gallons per day per foot.
K
=
Coefficient of permeability in gallons per day per square foot.
H
=
Height of groundwater above impervious boundary in feet.
he
=
Height of drainage system above impervious boundary in feet.
L
=
Influence distance from drainage system to groundwater table in feet.
E
=
Infiltration rate in gallons per day per square foot.
X
=
Distance from the influence distance to the point at which the flow rate is to be calculated in feet.
With X equal to the influence distance(L) and multplying by the perimeter length of the area to be dewatered, the equation becomes:
2 EL L
)
he H
(
2 K
P Q
2 2
Where:
Q
= Flow rate in gallons per day.
P
= Perimeter length of the area to be dewatered in feet.
The calculation for seepage flow rate into the Reactor/Auxiliary Building foundation area is made using the above equation and measured soil and rock permeabilities.
The permeability of the aquifer is taken as 600 feet per year, which is greater than the mean plus one standard deviation of all the soil and rock permeability test values for the site (Reference Section 2.4.13.2.5).
The perimeter length of the area to be dewatered is taken as 1,600 feet.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 37 The average elevation of the underdrain system is 540 feet msl. The bottom of the permeable zone is taken as elevation 530 feet msl. The post construction groundwater elevation is assumed to be at elevation 585 feet msl. at the drawdown radius of influence. This is a very conservative assumption for the groundwater level in certain areas of the plant yard, and completely ignores the lower post construction groundwater level that exists west of each unit due to the presence of the Turbine Buildings.
Groundwater recharge from rainfall is calculated by taking the maximum monthly rainfall reported for Charlotte, N.C. (Reference 14) with 32 percent of the rainfall infiltrating to the groundwater, and recharge uniformly distributed in time. The resulting computed infiltration rate is 0.0724 gallons per day per square foot (gpd/sq. ft.).
The radius of influence is computed from the following equation with the assumption of steady conditions developed during the month of maximum rainfall (Reference 18):
2
/
2 2
1
)
he H
(
E K
L
Using the infiltration rate rounded to 0.1 gpd/sq. ft. results in a computed radius of influence of about 600 ft.
The above equations and values result in a calculated flow of about 57 gallons per minute (gpm) into the Reactor/Auxiliary Building foundation area.
Using the computed inflitration rate of 0.0724 gallons per day per square foot (gpd/sq.ft), results in a computed radius of influence of about 705 ft. Using this radius of influence (705 ft.) results in a computed flow of about 57 gallons per minute (gpm) into the Reactor/Auxiliary Building foundation area.
It might be hypothetically possible that groundwater will "mound" under the elevated area of the Cooling Tower Yard, east of the Reactor/Auxiliary Buildings. A groundwater elevation of 615 feet msl., 5 feet below yard grade, is assumed. The radius of influence was calculated to be 934 feet. The resultant effect on the computed seepage flow rate is an additional 11 gallons per minute (gpm). A flow rate of 68 gallons per minute (gpm) into the Reactor/Auxiliary Building foundation area is used in the design of the groundwater drainage system.
The foundation underdrains and the exterior wall drains discharge into three sumps located adjacent to the Auxiliary Building (Sumps A and B are 10 feet by 10 feet by 15 feet deep. Sump C is 17 feet by 17 feet by 12 feet deep). Although the computed design flow used for the groundwater drainage system is considered conservative, the final system design will be based on measured flows in the three sumps. Groundwater discharged from the sumps will be measured periodically to verify that the design flow is not exceeded and that a steady state condition has been achieved.
The latest measured groundwater discharge from the three sumps was 35 gallons per minute (gpm) on September 12, 1981. From September 1980 to September 1981, the groundwater discharged from the three sumps has averaged 34 gallons per minute (gpm), which is half of the groundwater drainage system design flow.
Groundwater collected in these sumps is pumped to the yard storm drainage system. Two 300 gallon per minute pumps are installed in both Sump A and Sump B. However, only Auxiliary Building Sump Pumps A1 and B2 are classified as QA Condition 1. Two QA Condition 4 pumps, each capable of discharging 75 gallons per minute are installed in Sump C. These pumps, each capable of handling the total design flow of the groundwater drainage system, maintain the water level automatically in each sump. One pump will automatically start when the water level
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 38 (09 OCT 2016) reaches elevation 536 msl. In the unlikely event a pump fails to start and water rises above elevation 538 msl, an alarm will alert the operator and the second pump will automatically start.
In the unlikely event two pumps become inoperable in any one sump, groundwater would flow through the many redundant channels to the other sumps. Since all three sumps are interconnected by the foundation underdrains and exterior wall drains, all six pumps are available to discharge groundwater.
Multiple redundancy of vital system components will assure the ability of the system to function over the life of the plant. In the unlikely event that a single flow channel or wall drain becomes blocked, groundwater will flow to the sumps through any of the many redundant drain routes available. Monitoring of pump operation provides assurance that the zoned wall filter, drains and pumps are properly functioning.
The design and anticipated operation of the groundwater drainage system meet the requirements of Branch Technical Position HGEB-1, Rev. 2 NUREG 0800, Section 2.4.12.
In order to estimate the response characteristics of the aquifer, should a failure of the Category 1 underdrain system occur, the underdrain system was modeled by using a two-dimensional finite difference solution to the unsteady flow equations (Reference 17). Aquifer parameters were assumed based upon the results of the field investigation. Simulations were made with a permeability equal to 600 feet per year, and an aquifer storage coefficient equal to.05. The bottom of the aquifer was assumed to be at elevation 530 feet msl, ten feet below the elevation of the underdrain system. An infiltration rate was estimated by assuming that 32 percent of the maximum monthly rainfall at Charlotte, North Carolina, (Reference 14) infiltrates to the groundwater system, and that this recharge is uniformly distributed in time. These assumptions yield a recharge rate of 0.0724 gpd/sq. ft.
Initial conditions for the simulation were derived by assuming that the water table was at steady state at time t=o (the time of underdrain failure). The edge of the structure was assumed to represent an impermeable boundary, and a line source was placed at a distance of 605 feet from the structure to simulate the long-term effects of recharge and of groundwater flow from areas distant from the structure. The line source was held at a constant elevation of 585 feet msl. (assumed site area groundwater level) throughout the simulation.
The results of these simulations are presented in Figure 2-62 shows water level at the structure as a function of time after underdrain failure. These curves neglect all storage in the underdrain system, and consider only storage in the aquifer.
The Reactor and Auxiliary Buildings were statically analyzed to withstand soil pressure along with hydrostatic and uplift forces resulting from a groundwater rebound to yard grade (El.
593+6). This analysis did not consider any other loadings, was not a design condition, and was not considered in the loading combinations listed in Section 3.8.
An analysis of postulated pipe breaks in the yard and the subsequent flooding of the groundwater drainage system concluded that a rupture in the Nuclear Service Water pipe will induce the greatest quantity of water into the system. The Nuclear Service Water system is a moderate energy fluid system and is evaluated for rupture per NRC Branch Technical Positions MEB 3-1, and APCS 3-1 "Moderate Energy Fluid System Piping." Using these positions, a through wall leakage crack 21 inches long by 0.25 inches wide is assumed to develop in the pipe and induce 2360 gallons per minute of water into the groundwater drainage system. In an assumed time of 15 minutes before the isolation valves are closed, 35,400 gallons of water is pumped into the system. After the valves are closed, an additional 86,000 gallons from the pipe will drain into the system, during an estimated 40-minute time period.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 39 Water from the postulated pipe break flows through the zoned wall filter where it is intercepted by the flow channels and perforated pipe which carries the water to all three sumps. The total capacity of the six pumps is 1350 gallons per minute. This compares to the postulated induced flow rates of 2360 gallons per minute for the 15 minutes prior to shutting the isolation valves and the 2150 gallons per minute after the valves are closed. The net difference between the amount of water draining into the system and the amount of water pumped out of the sumps is approximately 47,150 gallons. The storage capacity of the three sumps (below pump level) is 48,000 gallons, thus the sumps alone could store the 47,150 gallons excess. The storage capacity of the flow channels in the underdrain system is approximately another 16,000 gallons, giving an added margin of storage capacity.
A postulated break in the Condenser Circulating Water piping is also investigated. The Auxiliary/Reactor Building groundwater drainage system is isolated from the CCW piping by a nominal 17-foot (5.2 m) minimum thickness impermeable layer of compacted earth backfill as shown in Figure 2-66. If a break in the CCW piping induces the flow into the Turbine Building groundwater drainage system, no effect on the Auxiliary/Reactor Building system will be realized since the two systems are totally isolated.
A failure of the Condenser Circulating Water System inside the Turbine Building is considered highly unlikely. However, a rupture of two water boxes (one on each unit) in this system and the subsequent flooding is analyzed. Such a postulated rupture could induce 2,069,000 Ft3 of water into the Turbine and Service Buildings. The top of this water would be at elevation 576.95 ft.
(175.85 m) msl. A watertight barrier with a top elevation of 577.5 ft. (176.02 msl) is designed to contain this water within the Turbine and Service Buildings and prevents water from infiltrating the Auxiliary Building.
The groundwater environment in the immediate vicinity of the site will be substantially changed by construction. The effects will be to decrease the slope of the water table and thus to increase the transit time of contaminants moving from the site to any discharge point. Since the bottom elevations of the structures are below the natural water table, an underdrain system has been installed to lower the water table. This will result in a minimum groundwater level of about elevation 540 feet msl. in the Reactor/Auxiliary Building area and a depression of the water table with groundwater flow toward the Reactor/Auxiliary Building area from all directions inside the radius of influence. The groundwater level will approach the surface elevation of Lake Wylie near the lake shore. A groundwater level monitoring program has verified the expected construction stage drawdown and will establish the stabilized elevation of groundwater outside the Reactor/Auxiliary Building walls.
Table 2-78 shows the latest results of the permanent groundwater monitoring. All wells are dry with the exception of Groundwater Monitor Well 4 which has groundwater at elevation 548.5 feet msl (1.5 feet below the top of the 4.0 feet thick foundation mat). A 4.0 feet thick concrete mat will experience no resultant uplift load until submerged more than 9 feet. Therefore, the ground water monitoring verifies the assumption that the groundwater drainage system eliminates uplift and hydrostatic forces on subsurface walls and foundation mats.
Twelve permanent groundwater wells are installed in the zoned wall filter around the perimeter of the Reactor/Auxiliary Building walls. Continuous monitoring devices will be installed in six of the twelve wells to monitor the groundwater level in the zoned wall filter. In addition to the continuous monitoring devices, each of the six wells have three points of alarm to alert the plant operator to a rise in groundwater. The first alarm is set at the elevation of the top of the adjacent floor slab in the building. The second alarm point is set five feet and the third fifteen feet above the top of the adjacent floor slab. Any alarm will alert the plant operator to a groundwater rise.
The remaining six wells without monitoring devices will be available to dewater the zoned wall
UFSAR Chapter 2 Catawba Nuclear Station 2.4 - 40 (09 OCT 2016) filter in the unlikely event of a rise in groundwater. Remedial action as described in the Selected Licensee Commitments would be in when groundwater rises above the elevation of the floor slab in the wall drain. The location of the twelve groundwater wells is shown in Figure 2-60.
In addition to the twelve groundwater wells around the perimeter of the Reactor/Auxiliary Building walls, sixteen temporary observation wells are located at strategic locations to monitor groundwater levels at the site. The locations of the observation wells are shown in Figure 2-78.
The latest results of the temporary observation well monitoring are provided in Table 2-79.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.4.14 Selected Licensee Commitments and Emergency Operation Requirements With the exception of the Selected Licensee Commitments describing remedial action required in the event of a rise in groundwater as described in Section 2.4.13.5, the hydrologic design bases developed in the preceding sections do not necessitate emergency procedures to ensure safety-related plant functions. In the event of an accidental radioactive release, industries and municipalities that utilize Lake Wylie and the Catawba River downstream as a water supply are notified of possible contamination.
2.4.15 References
- 1. U.S. Weather Bureau, Report No. 33 - Seasonal Variation of the Probable Maximum Precipitation East of the 105th Meridian for Areas of 10 to 100 Square Miles and Durations of 6, 12, 24, and 48 hours5.555556e-4 days <br />0.0133 hours <br />7.936508e-5 weeks <br />1.8264e-5 months <br />. Hydrometeorological Section, Hydraulic Services Division, in cooperation with the U.S. Army Corps of Engineers, 1956.
- 2. Department of the Army, Office of the Chief of Engineers, Civil Engineering Bulletin No. 52-8, Standard Project Flood Determinations, Washington, D.C., 1952.
- 3. A Forest Atlas of the South, Forest Service, U.S. Department of Agriculture, 1969, p. 8.
- 4. Engineering Manual for Civil Works, Part III, Chapter 5, April 1946.
- 5. Design of Small Dams, U.S. Department of the Interior, Bureau of Reclamation, Second Ed.
1973.
- 6. Flood Study and Description of Flood Routing Analysis by Computer for the Catawba River Hydroelectric System, November 1, 1968 by Chas. T. Main, Inc.
- 7. "Seismic Failures of Chillean Tailings Dams" by Ricardo Doyey and Leonardo Alvarez, Journal of the Soil Mechanics and Foundation Division, A.S.C.E., Vol. 93, No. SM6, November, 1967.
- 8. "Mechanics of Washout of an Erodible Fuse Plug," E. Roy Tinney and H. Y. Hsn, A.S.C.E.,
Vol. 87, No. HY3, May 1961.
- 9. Johnstown: The Day the Dam Broke by Richard O'Connor, J. B. Lippencott Co.,
Philadelphia, Pennsylvania, 1957.
- 10. Rainfall Frequency Atlas of the United States, Technical Paper No. 40.
- 11. Handbook of Hydraulics, Ernest Brater and Horace King, McGraw-Hill, 1963.
- 12. Water and Wastewater Engineering, G. M. Fair, et al, Wiley, 1966.
- 13. "Prediction of Near-Surface Drift Currents from Wind Velocity," Jim Wu, A.S.C.E. Journal of Hydraulics Division, Vol. 99, No. HY9, September 1973.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.4 - 41
- 14. Annual Summary 1974, Local Climatological Data for Charlotte, North Carolina, National Oceanic and Atmospheric Administration, 1974.
- 15. Prout, W. E., "Adsorption of Radioactive Wastes by Savannah River Plant Soil," Soil Science, Vol. 86, No. 1, July 1958, pp. 13-17.
- 16. Iome, Y., and Kaufman, W. J., "Studies of Injection Disposal," Proceedings of Second Ground Disposal of Radioactive Wastes Conference, Chalk River, Canada, 1961, pp. 303-321.
- 17. Prickett, T.A. and Lonnquist, C. G., "Selected Digital Computer Techniques for Groundwater Resource Evaluation," Illinois State Water Survey, Urbana, Illinois, Bulletin 55, 1971.
- 18. Harr, M.E., Groundwater and Seepage, McGraw-Hill Book Company, New York, 1962, p.
43-44.
- 19. National Research Council, Transportation Research Board, NCHRP Report 70, Design of Sedimentation Basins, Washington, D. C., 1980.
- 20. Chow, V. T., Handbook of Applied Hydrology, McGraw-Hill Book Company, New York, 1964, p. 17-23.
- 21. U. S. Army Corps of Engineers, HEC-1: Flood Hydrograph Package, Computer Program 723-X6-L2010, 1981.
- 22. McCuen, R. H., A Guide to Hydrologic Analysis Using SCS Methods, Prentice Hall, Inc.,
New Jersey, 1982.
- 23. Hydrometeorological Report No. 51, "Probable Maximum Precipitation Estimates, United States East of the 105th Meridian," National Oceanic and Atmospheric Administration, June 1978.
- 24. Hydrometeorological Report No. 52, "Application of Probable Maximum Precipitation Estimates, United States East of the 105th Meridian," National Oceanic and Atmosphere Administration, August 1982.
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Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 1 2.5 Geology and Seismology HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE UPDATED]
2.5.1 Basic Geologic and Seismic Information Geologic and seismic investigative studies for Catawba Nuclear Station, begun in 1970 with preparation of the Preliminary Safety Analysis Report, include the following:
- 1. A comprehensive review of geologic and seismic data including published and unpublished reports and maps.
- 2. Examination of gravity and aeromagnetic data, topographic maps and aerial photographs.
- 3. Geologic mapping of the site and reconnaissance geologic mapping of an area within a 10-mile radius of the site.
- 4. Consultation with recognized experts.
- 5. Completion of more than 100 soil test and rock core borings and 9500 ft of seismic refraction traverses.
- 6. Study of test pits and trenches.
- 7. Study of thin sections of rock and rock core samples.
- 8. Radiometric dating.
Beginning in 1975, a detailed geologic investigation was performed at the site after the discovery of brecciated zones in some excavations. The results of the investigations of all the brecciated zones found at the site are included in the "Final Geologic Report on Brecciated Zones" (Reference 107). That report includes the results of the following:
- 1. Detailed geologic mapping at scales of 1 inch equals 1, 5 and 10 feet.
- 2. Excavation of backhoe test pits.
- 3. Brunton compass readings of nine structural elements.
- 4. Seismic refraction lines
- 5. Vertical and angle core borings.
- 6. Thin section study of samples of saprolite, rock and rock core.
- 7. X-ray diffraction analysis
- 8. Radiometric dating by uranium-lead and potassium-argon techniques.
- 9. A review of local seismicity.
- 10. A better understanding of the site geologic history through utilization of advances in concepts and techniques since submission of the PSAR.
The geology of large portions of the site was completely mapped in detail and there is a complete photographic record of all exposures. Mapped areas include the excavations for the Powerhouse (Reactor, Auxiliary, Service and Turbine Buildings) and the S.N.S.W. Pond Dam.
Results of those investigations are presented in the "Final Geologic Report on Brecciated Zones" and are summarized in subsequent portions of this document.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 2 (09 OCT 2016) 2.5.1.1 Regional Geology 2.5.1.1.1 Geologic History In late Precambrian and early Paleozoic time (see Table 2-80 for the Geologic Time Scale) the region of the present Appalachian Mountain System, in which the site is located, was a geosyncline in which thick deposits accumulated, mainly sedimentary in the northwest and volcanic in the southeast (Reference 1). In the Blue Ridge Province, sedimentary sequences locally more than 5 miles thick overlie billion-year-old basement (References 1, 2). The volcanic-sedimentary sequence in the Piedmont Province is about 6 miles thick in some places (Reference 3) and the basement on which it was deposited is unknown in most of the Piedmont.
By the late Paleozoic, sediments were being shed north westward from uplifts probably located in what is now the Piedmont (Reference 1).
During the Paleozoic, several episodes of intrusion and deformation, and at least one metamorphic episode affected rocks of the Blue Ridge and Piedmont (Reference 1). Early to mid-Paleozoic regional metamorphism as high as upper amphibolite facies occurred over large parts of both provinces. Deformation of sediments northwest of the Blue Ridge was restricted to folding and thrust faulting beginning in the mid-Paleozoic, without extensive metamorphism.
Toward the end of the Paleozoic, the whole region was uplifted and has not since been totally submerged. By Late Triassic, core areas of the Appalachians were deeply eroded, and non-marine sediments were deposited in a great rift-valley graben system (Reference 1). Diabase dikes and sills were formed during and after this Triassic sedimentation.
Several plate-tectonic models pertinent to the evolution of the Appalachian region have been presented in the last few years. Hatcher (Reference 4) proposed a northwest-dipping subduction zone, with continent-continent collision in the late Paleozoic. Odom and Fullagar (Reference 5) suggested an Ordovician collision of island arc and continent above a southeast-dipping subduction zone followed by late Paleozoic continent - continent collision related to a northwestdipping zone. In the interpretation of Rankin (Reference 6), the Inner Piedmont belt, Charlotte belt and Carolina slate belt are part of an African plate thrust onto the North American plate.
During the Mesozoic and Cenozoic Eras erosion of the present Appalachian Mountains supplied the sediments that comprise the Coastal Plain formations which bury remnants of the eroded Paleozoic mountains.
Pertinent geologic and seismologic information developed since first submittal of the FSAR, particularly information concerning hypotheses of Charleston, South Carolina seismicity (Section 2.5.2.3.3), are discussed in Section 2.6.
2.5.1.1.2 Physiographic, Lithologic, Stratigraphic and Structural Settings In this region areas of similar geologic structure and geomorphic history are associated in physiographic provinces. The physiographic provinces within approximately a 200-mile radius of the Catawba site include the Appalachian Plateaus, Valley and Ridge, Blue Ridge, Piedmont and Atlantic Coastal Plain, as shown on Figure 2-67. The site is located in the Piedmont Province. The physiography of the Piedmont as well as each of the surrounding provinces is discussed in subsequent sections.
The lithologic and stratigraphic relationships are shown on the Regional Geologic Map (Figure 2-68) and major tectonic features on the Regional Tectonic Map (Figure 2-69). These are schematically related on the Regional Geologic Cross Section (Figure 2-70). A compilation of the latest aeromagnetic data is presented on the Regional Aeromagnetic Map (Figure 2-71,
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 3 Sheet 1) and a similar compilation of gravimetric data on the Regional Bouguer Gravity Anomaly Map (Figure 2-72). Portions of a composite map of recent aeromagnetic data currently being assembled by the North Carolina Geological Survey are presented in Figure 2-71, Sheets 2 and 3, and are discussed in subsequent sections.
2.5.1.1.2.1 Piedmont Physiographic Province The Piedmont Province is a deeply eroded, plateau-like segment of the Appalachian Mountain System. The Piedmont in this region is about 80 to 120 miles wide. It is bounded on the northwest by the Blue Ridge Province and on the southeast by the Atlantic Coastal Plain Province. The plateau generally slopes southeastward from an elevation of about 1200 ft near the Blue Ridge to about 400 ft near the Coastal Plain.
The major streams flow in a generally southeastern direction across the Piedmont surface. The major drainage pattern is dendritic, modified by lithologic control where sufficiently thick units of quartzite, metarhyolite, strongly sheared rocks or marble are present. It is also modified by rock jointing so that short reaches of first, second and third order streams exhibit linearity and sets of parallel trends which are angulate.
The Piedmont Province at the latitude of the North Carolina-South Carolina state line has been divided into four northeast-trending geologic belts as shown on Figure 2-73: the Inner Piedmont, Kings Mountain, Charlotte, and Carolina slate belt (Reference 7). The older country rocks in these belts are sequences that were originally sedimentary and volcanic deposits. The site is located in the Charlotte belt.
Recent work suggests that the belts may be the result of different responses to regional metamorphism and tectonics by rocks that were originally similar (Reference 8). In some places the boundaries are indistinct and a sequence of rocks may extend from one belt into another (Reference 8).
Normal faulting traditionally assigned Triassic age is widespread in North Carolina. Sediments rapidly accumulated in the resulting basins. Faulting and deposition of sediments were followed by intrusion of diabase dikes into the sediments and the surrounding crystalline and metamorphic rocks. These are generally accepted as Triassic events. Some paleomagnetic data (Reference 9) have been interpreted to indicate middle Mesozoic age for some intrusions.
However, the analytical error in radiometric dating could place the dikes in the Lower Jurassic, approximately 170 million years ago.
Most geologic time scales, for example, Table 2-80, indicate that the Triassic Period spanned the interval from about 225 to 180 million years ago. Based on new K-Ar data plus evaluation of ages from the literature, Armstrong and Besancon (Reference 10) suggested that the boundaries may be 255 and 210 million years. The suggestion for a 210 million year boundary is based on K-Ar ages of 200 and 206 million years for post-Triassic plutons in Canada and Idaho. The suggestion that the Permian-Triassic boundary should be at 255 million years is rather arbitrary but is consistent with data from a plot of sediment thickness versus time. The boundaries suggested by Armstrong and Besancon (Reference 10) do allow two K-Ar ages (each approximately 250 million years) on so-called Triassic diabase dikes from Virginia and North Carolina to fit into the Triassic Period.
Occurrences of laumontite and related minerals (prehnite is a common associate) are widespread in the Piedmont and are reported from 34 locations in North Carolina and Virginia (Reference 11). Laumontite generally occurs as fracture fillings and in hydrothermally altered zones adjacent to the fractures. Near Durham, North Carolina, laumontite occurs in a fracture in Mesozoic diabase (Reference 12) and near Leesburg, Virginia, prehnite and zeolites occur in
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 4 (09 OCT 2016) fractures and vesicles in diabase (Reference 13). The association of zeolites and related minerals with diabase suggests that the hydrothermal mineralization is related to the early-to mid-Mesozoic development of fractures along which the diabase was introduced. Therefore, the zeolites associated with Mesozoic fracturing and intrusion are probably of equivalent age, about 150 million years. It is assumed that other occurrences of zeolites in the Piedmont have similar origin and age.
2.5.1.1.2.1.1 Inner Piedmont Belt The Inner Piedmont is the westernmost belt of the Piedmont located immediately southeast of the Brevard zone. Within North and South Carolina, it has a width of approximately 50 miles but thins both to the north and to the south. The predominant rocks in this belt are gneisses and schists. However, they are interspersed with granitoids and a few scattered bodies of mafic and ultramafic rocks. In summarizing radiometric dates from the Inner Piedmont, Butler (Reference
- 14) concludes that high rank regional metamorphism was at least 430 to 410 million years ago (Silurian), and that the peak of metamorphism had passed by 400 to 375 million years ago (Silurian-Devonian). Because Butler's suggested time span (430 to 410 million years) for the peak of metamorphism is a minimum age, this event could have occurred at the same time as his suggested 470 to 430 million year time for the Blue Ridge metamorphic peak (Section 2.5.1.1.2.3). An Ordovician age (approximately 435 to 500 million years) for the major regional metamorphism and intrusion was suggested for this belt by Rodgers (Reference 1). Regardless, all authorities agree that the significant metamorphism occurred at least 375 million years ago.
The general structure within this belt is characterized by irregular foliation of low dip and some broad folds transverse to the northeast regional geologic trend (References 1 and 7).
2.5.1.1.2.1.2 Kings Mountain Belt The Kings Mountain belt is southeast of the Inner Piedmont and extends at least from Iredell and probably Rowan County, North Carolina (Reference 15) to Cherokee County, South Carolina. It is lenticular with a maximum width of 20 miles. The belt may extend for a significant distance northeast and southwest, although the continuity is obscured by intrusive bodies and metamorphic alteration. The principal distinctions between the Kings Mountain belt and its adjacent belt to the northwest (Inner Piedmont belt) are the higher grade of metamorphism, the generally lower dip of foliation and the generally broader (more open) folds in the Inner Piedmont (Reference 1).
The Kings Mountain belt contains a complex series of deformed rocks consisting of felsic and mafic schists and gneisses, quartzites, conglomerates and marble, generally considered to be of Precambrian and early Paleozoic age. Economically important middle to late Paleozoic pegmatites cut metamorphic rocks near the western edge of the Kings Mountain belt (References 16, 17).
Recently published literature concerning the geology of the Kings Mountain belt include five zones of mylonitic deformation with superimposedsemibrittle-faulting (References 122, 123).
The closest of these to the Catawba site, the Boogertown shear zone (approximately 10 miles from the site), occurs discontinously along the Kings Mountain belt-Charlotte belt boundary. The minimum age of deformation in the shear zones is interpreted to be middle to late Devonian (370-350 million years before present) based on the age of spodumenepegmatites related to one of the zones (Reference 124). This shear zone does not change the geologic and seismic evaluation of the site.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 5 2.5.1.1.2.1.3 Charlotte Belt The Charlotte belt extends from near Winston-Salem, North Carolina into South Carolina (Reference 7). It has a width of 30 to 40 miles. Rocks in this belt consist of a complex series of intrusive rocks, with some schist, quartzite, gneiss and amphibolite probably derived from sedimentary and volcanic deposits (Reference 18). Metamorphic rocks are mainly in the amphibolite facies. The most common intrusive rocks range in composition from granite to gabbro and some of the granitic bodies are of batholithic dimensions (References 7 and 9). It is mainly the extensive complex of intrusive rocks which distinguishes the Charlotte belt from the adjacent belts.
Three principal episodes of plutonism are delineated for the Charlotte and Carolina slate belts (References 19, 20): (1) Pre-metamorphic plutons range in composition from granite to gabbro and have ages of 595 to 520 million years (Precambrian-Cambrian), (2) Granite plutons, some probably syn-metamorphic and some post-metamorphic, and a group of post-metamorphic gabbro-diorite-syenite plutons have ages of 415 to 385 million years (Silurian-Devonian), and (3) Post-metamorphic granitic intrusions are about 300 million years old (Pennsylvanian). North-to northwest-trending Mesozoic diabase dikes are widespread in the region.
Fullagar (Reference 20) and Bulter (Reference 14) concluded that the thermal peak of regional metamorphism in the Charlotte belt occurred 380 to 420 million years ago. Fullagar based this conclusion on age determinations of 415 to 385 million years on metamorphosed adamellite plutons near Salisbury, North Carolina and determination of about 407 million years on the unmetamorphosed Lowrys adamellite (18 miles southwest of the site) and about 413 million years on the unmetamorphosed Concord syenite (35 miles northeast of the site). The adamellite at the Catawba site is part of one of the pre-metamorphic plutons of 595 to 520 million year age.
K-Ar dates of about 300 million years on whole rocks and mineral separates at the Catawba site have been obtained and represent an episode of rapid uplift and cooling late in the metamorphic history of the site. This cooling event correlates with a major episode of intrusion of adamellite bodies elsewhere in the Charlotte and Carolina slate belts 300 million years ago (Reference 20).
With the exception of a few broad folds such as the anticline at Nanny Mountain, South Carolina (Reference 18) and the Davie County Triassic fault basin, the structure of the Charlotte belt is a function of plutonic contacts.
Results of a recent geophysical investigation in parts of southern Mecklenburg County, North Carolina and York and Lancaster Counties, South Carolina suggest a fault that terminates about 5 miles east of the Catawba site. The fault is indicated to trend east-west to east-northeast.
Mapping in this area by USGS personnel reportedly has indicated no field evidence of this fault.
Thus, the USGS states that no faults will be shown on the forthcoming geologic map of the Charlotte 2 degree sheet in this area of Mecklenburg County, North Carolina and York and Lancaster Counties, South Carolina (Reference 125). This postulated fault does not change the geologic and seismic evaluation of the site.
Details of lithology and structure of the Charlotte belt in York County, South Carolina, where Catawba Nuclear Station is located, are presented in Section 2.5.1.2.1.
2.5.1.1.2.1.4 Carolina Slate Belt The Carolina slate belt is the easternmost belt exposed in the Piedmont. It extends from Georgia to Virginia, with its greatest development in North Carolina where it has a width of approximately 50 miles. The term "slate belt" is not geologically correct but one which is retained because of tradition (References 3 and 8). Rocks comprising this belt consist of both metavolcanics and metasediments. The metavolcanics include tuffs, rhyolitic and andesitic flows
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 6 (09 OCT 2016) and breccias; the metasediments include slate and argillite. Considerable recent mapping has resulted in division of a portion of the slate belt into stratigraphic units (References 21 and 22). It is thought that the slate belt rocks of North and South Carolina can be extended along strike with both the Little River series of Georgia and with the Virgilina volcanic group of northern North Carolina and Virginia (Reference 3). Intrusions ranging in composition from granite to gabbro are also found within the Carolina slate belt.
It is generally thought that the metavolcanics and metasediments of the Carolina slate belt are late Precambrian to early Paleozoic in age. Paleontological evidence from Stanly County, North Carolina (Reference 23) and recent radio metric age determinations of approximately 520 millions years (Reference 20, 24) indicate that some of the metasediments and metavolcanics are of Early to Middle Cambrian age. Some of the granitic intrusions are Cambrian, 570 to 535 million years old, while others are Pennsylvanian in age, with dates of 325 to 299 million years (References 20 and 24). The youngest intrusions are the north west to north trending diabase dikes traditionally assigned Triassic age.
2.5.1.1.2.1.5 Tectonic Structures For this discussion, tectonic structure is considered to be a deformation of major scale (extent measured in miles) in the earth's crust. Tectonic structures that are prominent regional features in the southern Appalachians are shown on Figure 2-69. Some aspects of various major structures in the southern Piedmont are listed in Table 2-81. Those and other structural features of the southern Piedmont are shown on the Subregional Geologic Index, Figure 2-73. Tectonic structures that might be of significance to the Catawba site, either because of their proximity to the site or because of the recent re-evaluations of their origin or extent, are discussed in the following paragraphs.
The major structural feature which, in North Carolina, closely parallels the boundary between the Charlotte belt to the west and the Carolina slate belt to the east is the Gold Hill-Silver Hill fault complex. Intermittent and sub parallel quartz veins, stringers of ore, sheared rock and a major change in lithology exist within a zone trending northeast-southwest through the Gold Hill mining district of North Carolina. This zone, the Gold Hill-Silver Hill fault complex, extends southwest of Gold Hill for 25 to 30 miles becoming indistinct in Union County, North Carolina.
The closest point of the fault complex to the site is 17 miles. A 600-ft core taken from the Edgmoor Granite, about 18 miles south of the site, contains brecciated zones that have been interpreted to be the Gold Hill Fault (Reference 25). If thus projected into South Carolina, the fault complex could pass as close as 11 miles southeast of the site.
Based on evaluation of considerable data including shear zones mapped parallel to and thought to be contemporaneous with the Gold Hill Fault (References 26, 27) and a mapped Triassic diabase dike (Reference 26) which cuts one of the shear zones, Bulter (Reference 27) concluded the age of the faulting to be pre-Triassic. This is supported by K-Ar dates of 238 to 254 million years (Permian) obtained from an unsheared dike sample collected from within one of the shear zones of Bates and Bell (Table 3 of Brecciated Zones Report). Sundelius and Taylor (Reference 28) show an undeformed diabase dike cutting the Gold Hill Fault itself. This dike has been sampled and K-Ar dates of 228 and 232 million years obtained (Table 3 of Brecciated Zones Report) further substantiating the pre-Triassic age assigned to activity along the Gold Hill Fault.
The location of the Gold Hill-Silver Hill fault complex is indicated on the Regional Aeromagnetic Map (Figure 2-71, Sheet 2). An abrupt change in the trend and density of magnetic contours occurs there, confirming the presence of a discontinuity. A strong magnetic trend extends northeastward from the fault complex (beyond Greensboro on Figure 2-71, Sheet 2). This could
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 7 be an extension of the Gold Hill-Silver Hill fault complex, although no geologic data exist to substantiate that speculation. In any case, that northeasterly trend is crossed by north-trending lenticular features on the aeromagnetics, most likely diabase dikes.
The Eastern Piedmont fault system is defined as "a related series of linear polyphase-deformed cataclastic zones of varying thickness and dip associated with characteristic magnetic anomalies" (Reference 29). It is described as extending from Alabama to Virgina as shown on the Regional Tectonic Map (Figure 2-69) and includes the Towaliga and Goat Rock faults, previously thought to extend only across Georgia (Reference 30) and possibly into South Carolina (Reference 25). It also includes the Modoc fault that crosses the Georgia-South Carolina state line (Reference 25). The Eastern Piedmont fault system is proposed to have a geologic history similar to the Brevard zone: early mylonitization, perhaps related to ductile folding, followed by brittle deformation (Reference 29). Talwani and Howell (Reference 31) suggest that epicenters of historic and recent small earthquakes plot along the projected trend of the fault system. However, the Siloam granite of Georgia cuts the projected trend of the Goat Rock fault, and with a rubidium-strontium age of 269 million years (Reference 32), the Siloam granite would be younger than faulting. Also,as shown on the Geologic Map of Georgia, the fault system is cut by Mesozoic diabase dikes (Reference 30). Thus, there is surficial geologic evidence that contradicts associating this fault system with historic and recent small earthquakes.
The trace of the Eastern Piedmont fault system in North Carolina and South Carolina can be detected on the Regional Aeromagnetic Map (Figure 2-71, Sheet 3). It is crossed by the traces of north-trending lenticular magnetic features that are probably diabase dikes.
2.5.1.1.2.1.6 Linear Features Linear features in the region surrounding the site are shown on the Regional Linear Features Map, Figure 2-74. The area evaluated includes the Piedmont and parts of the Coastal Plain and excludes the area west of the Brevard zone. ERTS-1 (LANDSAT-1) images (Bands 5 and 7) at scales of 1:500,000 and 1:250,000 were utilized for the regional study.
The expression of linear features on imagery of the Piedmont is, for the most part, subdued and a subjective approach is required to delineate them. Tonal changes, principally related to land use patterns are a frequent basis for identification of linear features. Such variations in land use patterns may reflect differences in soil fertility which, in turn, may reflect differences in compositions of the underlying rock. Other bases for delineating linear features are alignment of linear or curvilinear river segments and abrupt changes in the shape of lake shores or orientation of drainageways.
Linear features within a 15-mile radius of the site are discussed in Section 2.5.6.10.
2.5.1.1.2.2 Brevard Zone The Brevard zone is not a physiographic feature, but rather a major structural geologic feature of the southern Appalachians (see Figure 2-69). It is located approximately between the Piedmont Province to the southeast and the Blue Ridge Province to the northwest and varies in width from one to four miles. This zone consists of rocks of diverse composition including gneisses and schists. Various theories explaining the zone include fault theories, fold theories, megatectonic theories or combinations thereof. Roper and Justus (Reference 33) suggest that the Brevard zone is "polygenetic" in origin and has been subjected to considerable tectonic development ending in the Permian; others assign a different age, but in no instance younger than Triassic.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 8 (09 OCT 2016) 2.5.1.1.2.3 Blue Ridge Physiographic Province The Blue Ridge Province is a mountainous zone to the west of the Piedmont. It is approximately 15 to 70 miles wide and 600 miles long, extending from Pennsylvania to Georgia. The mountains are subdued, but rugged ranges exist. Surface elevations generally vary from 1,500 to 5,000 ft; the highest elevation in the Blue Ridge is 6,684 ft at Mt. Mitchell. In general, drainage is to the west. However, slopes separating the Blue Ridge and the Piedmont Provinces are typically steep and provide initial run-off for some of the largest streams of the Piedmont, which drain to the east. The southeastern boundary is primarily a topographic change where the steep slopes of the Blue Ridge join the rolling hills of the Piedmont.
The northwestern boundary coincides with major faults, along which metamorphic rocks have been thrust to the northwest over younger, unmetamorphosed sedimentary rocks of the Valley and Ridge.
The Blue Ridge is highly deformed with an increase in metamorphic grade from west to east.
Rock units included in this Province consist of layered biotite and muscovite schist and gneiss, hornblende schist and gneiss, granite, pegmatite, quartzite, marble and dolomite. These units are Precambrian and early Paleozoic metamorphics, generally of amphibolite grade, with some intrusions of predominantly early Paleozoic age. The schists and gneisses are considered the oldest rocks in the region. Re-evaluation of existing radiometric dating places the peak of Paleozoic metamorphism in the Ordovician Period at a minimum of 430 million years and a maximum of 470 million years (Reference 14). Pegmatite is younger, with recorded age determinations of 380 million years (Devonian) in the Spruce Pine, North Carolina area (Reference 34).
2.5.1.1.2.4 Valley and Ridge Physiographic Province The Valley and Ridge Province is a series of fold mountains expressed as linear ridges and wide valleys. It is about 50 miles wide in northeastern Tennessee, extends northeastward into Pennsylvania and southwestward towards Georgia and Alabama. It is characterized by narrow, elongate ridges and intervening valleys that trend northeast, paralleling the regional orientation of the inclined strata of Paleozoic sedimentary rocks. The ridges tend to be evenly crested and are underlain by sandstone, siltstone and siliceous limestone and dolomite. Valleys are typically broader than the ridges and are underlain by shale and less siliceous, less dolomitic limestone.
Large scale folds dominate the structure of the Valley and Ridge Province in the northern portion, while middle to late Paleozoic thrust faults are dominant in the southern portion.
Metamorphism is confined to minor effects along thrust faults.
Drainage in the Valley and Ridge Province generally follows the northeast trending valleys.
Some major streams flow across this trend for short distances because of entrenchment of ancient stream courses. Karst topography is common in many areas of carbonate rocks.
2.5.1.1.2.5 Appalachian Plateaus Physiographic Province The Appalachian Plateaus Province borders the Valley and Ridge to the northwest. Its easternmost segment in Tennessee is known as the Cumberland Plateau. This province varies from 20 to 100 miles in width and extends from central Alabama to New York. Its southeastern margin, bordering the Valley and Ridge, is an abrupt topographic rise known as the Allegheny Front which in eastern Tennessee is called the Cumberland Escarpment. Surface elevations range from about 700 to 3,000 feet. The topography is gently sloping to undulatory with localized mountainous areas. Erosion by the streams which drain to the east and west has resulted in some steep-sided valleys.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 9 Rock units underlying the Appalachian Plateaus are composed of thick sequences of late Paleozoic sedimentary rocks, predominantly sandstone and shale with subordinate amounts of limestone and coal. Rock strata are generally nearly horizontal, but are gently folded into a broad syncline with a few small thrust faults and elongate anticlinal folds superimposed.
The Highland Rim (Reference 35) forms the western boundary of the Appalachian Plateaus and the eastern margin of the adjacent Interior Low Plateaus Physiographic Province.
2.5.1.1.2.6 Atlantic Coastal Plain Physiographic Province The Atlantic Coastal Plain Province consists of the Coastal Plain and the Continental Shelf (Reference 36). The Coastal Plain is bounded on the east by the Atlantic Ocean and on the west by the Piedmont Province. It has a range of width of approximately 90 to 125 miles with a westward rise of approximately two to four feet per mile. The elevation ranges from sea level at the coast to about 500 feet.
The Coastal Plain is composed of Cretaceous to Recent continental and continental margin sediments. The sediments range from poorly consolidated to well consolidated and include gravels, sands, clays and other materials; some limestones and sandstones also exist. The sediments were deposited on a seaward sloping peneplain traditionally considered to be composed of rock similar to the more inland provinces. However, recently acquired aeromagnetic data show marked contrasts in basement materials underlying the Piedmont and the Coastal Plain suggesting differences in rock types and/or structural trends (Reference 37).
Deposition in the Coastal Plain began at least 130 million years ago, making the lower units Cretaceous or older. These units are over lain by successively younger sediments forming a wedge-shaped block that increases in thickness from its contact with the Piedmont on the west to a thickness on the order of 7,000 to 10,000 feet on the coast of the Carolinas (References 38 and 39).
2.5.1.2 Site Geology 2.5.1.2.1 Geology of the Subregion The site is located in the Charlotte belt of the Piedmont. See Figure 2-73 for Subregional Geologic Index and Figure 2-75 for Subregional Geologic Map. The oldest and most abundant rock at the site is adamellite from the first principal episode of plutonism (595 to 520 million years). The adamellite is part of a major map unit that underlies nearly half of the Charlotte belt in York County and extends northward into North Carolina and southward into Chester County, South Carolina (Reference 18). The rock ranges from well foliated to massive. Evidence of foliation is faint but commonly present in exposures of rock on the site. The foliated adamellite at the site and at many other localities in York County is cut by mafic dikes which have been metamorphosed to the amphibolite facies (References 18 and 40). The strike of the dikes is commonly northeastward.
Mineral assemblages in the pre-metamorphic rocks in the Charlotte belt of York County indicate regional metamorphism to amphibolite facies (Reference 18). Because of generally inappropriate chemical composition, aluminosilicate index minerals are rarely developed in the Charlotte belt; however, a sillimanitemus covite-quartz rock occurs about nine miles south-southwest of the site confirming the imposition of high-rank metamorphism (Reference 18).
Foliated adamellite at the site is part of one of the metagranitic bodies described by Butler and Ragland (Reference 19). These bodies are characterized by foliation parallel to regional trends, microtextures indicating deformation, and cross-cutting metamorphosed mafic dikes. The bodies
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 10 (09 OCT 2016) are particularly extensive in the Charlotte belt and in the northern Carolina slate belt. Fullagar (Reference 20) obtained whole-rock Rb-Sr ages ranging from 595 to 520 million years on the pre-metamorphic intrusions. Early intrusions nearest the site on which radiometric ages have been obtained are the Edgmoor, 14 miles southsoutheast of the site, and the Great Falls, 34 miles south-southeast, which were dated at 545 +/- 30 and 554 +/- 115 million years, respectively.
2.5.1.2.1.1 Geologic History Geologic investigations in York County have been conducted by Tuomey (Reference 41 and 42), Overstreet and Bell (Reference 43), Butler (Reference 18) and Law Engineering (PSAR). A synthesis of geologic literature suggests the following sequence of geologic events in York County.
Accumulation of a thick sequence of sandstone, siltstone, tuff, volcanic flows and perhaps impure calcareous sedimentary rocks during late Precambrain or early Paleozoic time.
A series of igneous intrusions ranging in composition from granite to gabbro during the late Precambrian and early Paleozoic. Some preceded and some were accompanied by metamorphism to the amphibolite facies followed by later greenschist (or retrograde) metamorphism. Metamorphism was accompanied by large scale folding with axial planes trending northeast and a foliation was imparted to the pre-metamorphic rocks.
Intrusion of post-metamorphic granitic bodies during the late Paleozoic (Pennsylvanian).
Emplacement of north-to northwest-trending diabase dikes during early to middle Mesozoic.
2.5.1.2.1.2 Physiography The topography of the Piedmont in the subregion is characterized by low, rounded hills and gentle slopes. Surface elevations range from about 450 to 800 feet. The principal stream in this portion of the Piedmont is the Catawba River which cuts across the northeast-trending geologic structure and flows to the south and southeast.
The results of study of aerial photographs and topographic maps of the subregion are presented on Figure 2-76, Subregional Drainage Map with Linear Stream Segments. Lithologic control is evident in the area of the Kings Mountain belt where a strong northeast orientation of drainage occurs. Northeast-trending structural control is reflected in the drainage pattern at the south end of Nanny Mountain, but the main effect of the Nanny Mountain structure is a general absence of drainage features. There is a zone of structurally controlled drainage trending northwest-southeast approximately 10 miles northeast of the site. This zone probably reflects the Mesozoic dike swarm that extends from the Blue Ridge of North Carolina to the Coastal Plain of South Carolina. Geologic structure of the subregion is discussed in Section 2.5.1.2.1.4.
2.5.1.2.1.3 Lithology The major rock types present in the subregion and shown on Figure 2-75 (Subregional Geologic Map) are discussed in the following paragraphs.
Phyllite - Phyllonite: These rocks occur in the Nanny Mountain area and at scattered locations throughout the Charlotte belt. They are very fine grained, white to bluish gray and are composed principally of sercite, quartz and iron oxides.
Quartzite: Quartzite up to 30 ft thick occurs at Nanny Mountain and vicinity. It is typically light colored and stained to a yellowish or reddish color by weathering of pyrite. Where much sericite is present, the quartzite has a distinct cleavage.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 11 Amphibolite: Amphibolites crop out over large areas in the Charlotte belt. They are fine to medium grained rocks of highly variable mineral composition in which hornblende and plagioclase are the essential minerals. The rock is commonly foliated but may be massive over large areas. Amphibolites also occur as xenoliths and dikes in older plutonic rocks or interlayerd with phyllite, gneiss and schist.
Foliated Adamellite: Adamellite is extensive throughout the Charlotte belt and is the largest map unit in York County (Reference 18). It is fine to medium grained, generally is light colored, but contains dark flecks of biotite. The major minerals are potassium feldspar, quartz, plagioclase and biotite. The rock mass has been affected by regional metamorphism, often exhibits a faint to moderate foliation, but in many areas is relatively massive. Metamorphosed mafic dikes cut the foliated adamellite. Adamellite is the predominant rock type on the site.
Meta-gabbro: Meta-gabbro occurs at scattered locations in the eastern part of the Charlotte belt.
It is medium to fine grained and contains hornblende and actinolite (replacing pyroxenes),
plagioclase, biotite, epidote and chlorite. The meta-gabbro is the product of regional metamorphism of bodies that were originally gabbro intrusions. Some bodies of meta-gabbro are cut by tonalite dikes.
Diorite and Tonalite: These rocks underlie large portions of the Charlotte belt and are among the oldest intrusives in the area. Diorite and tonalite are usually massive or faintly foliated. Diorite along the western margin of the Charlotte belt is strongly sheared. Angular xenoliths of amphibolite are common and, in places, diorite and tonalite are cut by metamorphosed mafic dikes. The rocks are medium grained and contain plagioclase, hornblende, biotite and quartz.
Classification is based on the relative abundance of these minerals.
Gabbro: Gabbro occurs as large intrusive masses assumed to be steep-sided stocks. It is dark gray and generally medium grained and massive. The major minerals are plagioclase, olivine, augite, hypersthene, hornblende and biotite. Terrain underlain by gabbro is characteristically low-lying and flat.
Coarse Grained Adamellite: Adamellite occurs as several relatively young intrusions in York County. Similar intrusives occur in a belt extending through the Piedmont of the Carolinas. This adamellite is coarse grained and porphyritic. The major mineral constituents are microcline, plagioclase, quartz, biotite and muscovite. It intrudes most other rock types in the area and is not metamorphosed. It is cut by diabase dikes.
Diabase: Diabase occurs in numerous dikes throughout the Piedmont. They can be traced on the surface by the characteristic rounded boulders and brownish yellow soil and in the subsurface by their characteristic magnetic signatures. The dikes strike from north-northeast to northwest and dip steeply. The most abundant minerals in the diabase are plagioclase, augite and olivine. These dikes are commonly referred to as being Triassic in age, but their true ages may range from Permian to Jurassic.
2.5.1.2.1.4 Structure The results of geologic mapping within a 10-mile radius of the site are shown on the Subregional Geologic Map, Figure 2-75. Structural features noted during the field reconnaissance generally conform to the regional structural pattern of the Piedmont. The mean of 80 Brunton compass readings of schistosity and foliation in York County is a strike of N44°E and a dip of 72° south east (Reference 44). Foliation in adamellite at the site has a similar attitude (Section 2.5.1.2.2.5).
The Nanny Mountain anticline is located about 3 miles northwest of the site. It was mapped as part of geologic reconnaissance during the PSAR studies. The field work included examination
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 12 (09 OCT 2016) of road cuts and outcrops and traverses over Nanny Mountain. Samples were collected for study in thin section. The results of those petrographic analyses are presented in the PSAR.
The rock types exposed at Nanny Mountain, phyllite-phyllonite, amphibolite and foliated adamellite, are indicative of metamorphism at relatively high temperatures and pressures.
Although faulting is not associated with the Nanny Mountain structure, the nature of the rocks exposed there suggests that some shearing was associated with the folding that produced the anticline. Results of radiometric dating by the potassium-argon method reported in the PSAR indicate that the Nanny Mountain structure is at least 300 million years old.
The Nanny Mountain anticline has a curved axial surface trending from N15°E to N40°E. The axis of a larger scale fold that is shown on the Tectonic Map of the United States (Reference 45) in the vicinity of the site also has a north east trend. Evidence for the existence of this fold was noted during the PSAR field studies and interpretations of aerial photographs. Measurements of foliation within a 10-mile radius of the site suggest a broad syncline trending northeast with an axial trace passing through or very close to the site. A similar conclusion can be drawn from the drainage pattern in the area surrounding the site (Figure 2-76).
Between two and four miles northwest of the site is an alignment of hills trending in a northeast-southwest direction. This alignment has been observed in the field and on aerial photographs, where there is a marked stream lineation pattern. Occurring within this area are several quartzite layers and numerous quartz veins, also trending in a predominant northeast-southwest direction. In the area of the Nanny Mountain anticline, the quartzites and quartz veins tend to bend in a more northerly direction, conforming to the northerly plunge of the anticline. Previously worked gold mines are distributed along this trend and are shown on Figure 2-75.
The quartz veins are thought to be associated with or earlier than the last period of granitic intrusion during late Paleozoic time, 300 million years ago. Gold and other mineral deposits associated with the quartz veins were probably emplaced during a period of mineralization during or shortly after the quartz intrusions. The regional granitic intrusions, quartz veins and later mineralization are generally regarded as part of a continuous orogenic process that took place during the middle to late Paleozoic (Reference 46). Quartz veins have been traced in the field southeast of the Nanny Mountain area where they are truncated by a Triassic (?) diabase dike. Similar field relationships between diabase dikes and vein deposits have been found in Virginia (Reference 46) which indicates that the diabase dikes were intruded after the quartz veins and the associated ore deposits were formed.
The pattern of major drainage features within approximately 15 miles of the site which was presented in the PSAR was supplemented during the investigation for the Brecciated Zones report. The purpose of the additional study was to determine if the orientation of linear segments of streams in the subregion could be correlated with a particular aspect of geologic structure.
This subregional drainage pattern and 145 linear stream segments are shown on Figure 2-76.
The orientations of the linear stream segments are plotted as the rose diagram included on Figure 2-76. This diagram shows the predominance of the northwest and northeast oriented stream segments. The dominant orientations are between N15°W and N50°W, with approximately N45°W the most common orientation. Northeast orientations between N20°E and N40°E are also shown to be common stream alignments.
A great number of field measurements of joints have been made on the site. The most common joint strikes are between N35°W and N50°W and between N30°E and N45°E (Section 2.5.1.2.2.5). The greatest density of joints on the site is a northwest-striking system. A strong correlation exists between the joint strikes and the orientation of linear stream segments, particularly in the northwest oriented system.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 13 Thus, the development of the regional drainage pattern has been controlled to a great degree by the northeast-and northwest-striking joint systems occurring in the study area. In addition, the linear stream segment orientations correlate well with the general northeastward structural and lithologic characteristics of the Piedmont.
A zone of structurally controlled drainage trending northwest-southeast approximately 10 miles northeast of the site (at its closest point) probably represents part of a Triassic (?) dike swarm, that extends from the Blue Ridge of North Carolina to the Coastal Plain of South Carolina.
Although it is a significant geologic feature in the southern Piedmont, the dike trend is not evident on LANDSAT-1 imagery (Section 2.5.1.1.2.1.6). This dike swarm and a number of other north-and northwest-trending diabase dikes in the Piedmont area are associated with forces that produced fractures in the earth's crust during a time approximately 180 to 240 million years ago. This age range is based on potassium-argon dating of dike materials from various locations in the Piedmont (Reference 10).
Considered in the light of recently developed theories of plate-tectonics, the diabase dikes and dike swarms are associated with the most recent rifting of continental plates about 200 million years ago (Triassic). Prior to this rifting, southeastern North America, northeastern South America and western Africa were contiguous. Because Triassic (?) dikes exist on all three continents, it is suggested that the intrusion of the diabase occurred before or during continental break-up. The pattern of dikes, primarily oriented north and northwest in the southeastern United States, formed in response to the crustal stresses accompanying break-up. The diabase represents intrusion of oceanic crustal material into the edges of the continental masses (Reference 47).
2.5.1.2.1.5 Unrelieved Residual Stresses There are numerous references in the literature that unrelieved residual stresses, including horizontal stresses greater than those expected from overburden loading alone, exist within the earth's crust. Observations of construction excavations in rock and behavior of rock supported structures for both nuclear and fossil fuel power plants have been made in the Piedmont. There is no record of adverse effects of such unrelieved residual stresses.
Available in-situ stress data from overcoring tests are listed in Table 2-82 and presented graphically on Figure 2-77. Except for Mt. Airy, North Carolina and Douglasville, Georgia, the average direction of L:
Max H
is north to northeast and therefore approximately follows the average geologic trend of the Piedmont. However, when geologic conditions near the test locations are considered, the influence of local geologic features on Max H
is not well defined.
The available data generally show all possible conditions:
Max H
ranges in direction from about parallel to local structure to approximately perpendicular to local structure. However, there are variables potentially affecting these "average" relationships. Factors which can affect the magnitude of the stresses measured by overcoring include temperature (Reference 48), depth of measurement, topography (Reference 49), equipment (Reference 50), method of data reduction (Reference 51), and working (quarrying) conditions in the areas of measurement.
With one exception the stresses listed in Table 2-82 were measured by the overcoring technique at depths less than about 100 ft at sites in the Piedmont. The data for the Bad Creek Pilot Tunnel were taken at a depth of about 500 ft in the eastern edge of the Blue Ridge (Reference 52). When compensated for variations in depth of measurement, all available in-situ stress data discussed here have mutually consistent magnitudes.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 14 (09 OCT 2016)
There is a history of no adverse effects from in-situ stresses in Piedmont rock. None were noted at the Catawba site.
2.5.1.2.2 Geology of the Site The discussion that follows includes data gathered during the PSAR and Brecciated Zones Report studies plus information accumulated during detailed geologic mapping of the foundation excavations.
2.5.1.2.2.1 Geologic History The geologic history of the site and its relationship to the geologic history of the region are summarized in Table 2-83. This table was assembled using data gathered during the PSAR studies, published and unpublished data concerning foliated adamellite bodies in the Charlotte belt and data gathered during the studies of the brecciated zones and other areas of the excavations.
The geologic history of the site can be considered to consist of three major phases: an initial phase dominated by plutonism (more than 500 million years ago), and intermediate phase dominated by metamorphism (about 400 to 300 million years ago) and a final phase dominated by hydrothermal activity (ending about 150 million years ago).
Rock types present on the site (discussed in Section 2.5.1.2.2.3) are adamellite, amphibolite, diorite, prophyritic diorite, aplite and pegmatite. These rocks originated as felsic, mafic and intermediate intrusions which have been metamorphosed to amphibolite facies and deformed by several episodes of faulting and brecciation. Uranium-lead age of zircon from the adamellite country rock (Table 2-83, Event IA) is 532+/- 15 million years (Brecciated Zones Report, Table 3) and represents the age of crystallization of the adamellite and probably the intermediate and mafic dikes as well. Offset of the early pegmatite dikes (emplaced during Event IB) accompanied emplacement of the mafic dikes (Event IC). An early (late magmatic) episode of faulting is recorded as stretched breccia fragments along some mafic dike contacts and necking out of some mafic dikes into undeformed adamellite (Event ID). Regional metamorphism reached its peak at least 380 million years ago (Silurian-Ordovician) with development of foliation in the adamellite and the mafic dikes (Event IIA). The foliation is older than the nearby, unfoliated Lowrys pluton, dated at 407 +/- 11 million years (Reference 20). Subsequent movement deformed the foliation in the mafic dikes at their contacts. Additional necking out of mafic dikes occurred and pegmatite stringers developed at the mafic dike pinch-outs (Event IIA).
Minerals typical of greenschist facies metamorphism locally replace the earlier amphibolite-facies minerals, mainly in and adjacent to the sheared and brecciated zones (Event IIB).
Potassium-argon dates on separated minerals and whole-rock samples range from 300+/-5 to 359+/- 11 million years (Brecciated Zones Report, Table 3, Sheet 2). The three different types of materials analyzed (biotite, hornblende, and whole-rock samples) have significantly different closing temperatures to migration of argon. The observation that these three materials define isochrons of 300 million years indicates a rapid cooling rate prevailed at that time. This suggests a more rapid uplift in the site area than had previously been suspected (Reference 14). The rapid uplift may correlate with widespread intrusion of granitic plutons about 300 million years ago (Section 2.5.1.1.2.1).
Minor thrust faulting then affected the site (Event III) followed by renewed normal-oblique faulting along the pre-existing trends defined by the mafic dikes (Event IV). A complex hydrothermal event followed which resulted in fracture filling and partial replacement of rock in and adjacent to preexisting steep faults and joints (Events VA and VB). In all cases veins containing the hydrothermal mineral assemblages (prehnite-calcite or laumontitecalcite) occupy
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 15 dilation fractures that parallel, occur within and cut across earlier tectonic features including sheared and brecciated zones. In no instances are these mineral fillings observed to be broken by later movements. Thus, these unbroken hydrothermal minerals are younger than shearing and brecciation. A potassium-argon analysis of laumontite scraped from filled fractures in rock core yields an age of 86+/- 30 million years (Brecciated Zones Report, Section 5.10.2). This represents a minimum age for the hydrothermal event and is also a minimum age for all tectonic events preceding its formation.
2.5.1.2.2.2 Physiography The site is located on the west side of Lake Wylie between two major tributaries to the Catawba River and Lake Wylie, Beaver Dam Creek to the north and Big Allison Creek to the south.
Surface elevations range form about 570 (Lake Wylie) to 640 feet. Pre-construction contours are typical of the low, rounded hills of the Piedmont.
2.5.1.2.2.3 Lithology The predominant rock type underlying the site is classified as adamellite and is fairly uniform in composition across the site. Mafic dikes constitute a subordinate rock type and are discontinuous and irregular across the site. Discontinuous and irregular dikes of intermediate composition are observed at one location on the site.
Both the mafic dikes and the adamellite are metamorphosed. The mafic dikes exhibit a moderate foliation and the adamellite exhibits a faint foliation. Mineral assemblages of amphibolite rank are observed in both rock types. Mafic dikes consist essentially of fine-grained plagioclase, hornblende and biotite. Plagioclase, making up almost one half the rock, is andesinelabradorite in composition. Hornblende occurs as dark green, stubby, euhedral to subhedral grains. Biotite occurs as subhedral to euhedral grains in subparallel alignment.
Adamellite is medium grained and contains plagioclase (oligoclase), potash feldspar, quartz, biotite and hornblende. Epidote, sphene, apatite and some iron oxides occur as common accessories in both rock types. In these rocks the equilibrium mineral assemblage hornblende-medium plagioclase-epidote-biotite (-quartz) is characteristic of the amphibolite facies. Pressure and temperature conditions that produce amphibolite facies mineral assemblages are typical of regional metamorphism of moderate to high grade.
Superimposed on the amphibolite-facies mineral assemblages, mainly in and near shear and breccia zones, are minor amounts of later-formed minerals typical of the greenschist facies, produced under lower temperature-pressure conditions. Quartz-epidote veinlets cut across foliation of metamorphosed adamellite. Original minerals in adamellite are replaced to varying degrees by muscovite, chlorite, epidote and calcite, and portions of amphibolite are altered to chlorite-rich and epidote-rich assemblages. The assemblage calcite-epidotemuscovite-chlorite-quartz is typical of the greenschist facies (Reference 53). Chlorite and fine-grained iron oxide replace biotite and wall rock adjacent to fractures. Iron oxide, mostly hematite (with other iron-bearing minerals), occurs mainly in shear planes and is a major component of some of the small shear zones. Iron oxide-rich shear zones developed essentially contemporaneously with chlorite and early calcite and are in most places minutely contorted and deformed.
The youngest mineral assemblages (not produced by weathering) observed in rocks at the Catawba site occur in and near dilation-type fractures. Minerals formed during this time are predominantly prehnite, laumontite and calcite. This assemblage has been introduced into the fractures by hydrothermal solutions (Reference 54). Prehnite is slightly earlier than laumontite.
Both are accompanied by calcite.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 16 (09 OCT 2016) 2.5.1.2.2.4 Structure Subsurface profiles based on the borings are indexed on Figure 2-78 and Figure 2-79 and presented on Figure 2-80 through Figure 2-93. The locations of refraction seismic traverses are also shown on Figure 2-78 and Figure 2-79 and the results are presented on Figure 2-94 through Figure 2-97. Results of uphole seismic compressional and shear wave velocity tests in borings A-61 and A-63 are presented on Figure 2-98 and Figure 2-99, respectively.
Except for a few scattered rock outcrops, the rocks at the site are covered by saprolite (residual soils containing relict rock texture and structure) varying from sandy silt to silty sand. The lower saprolite boundary is gradational into partially weathered rock which is underlain by hard rock.
Contours on the top of "continuous" rock (rock core having an RQD of 75 or more) are derived from boring data and are shown on Figure 2-100. The top of "continuous" rock has a variation of approximately 100 feet in depth across the site. The irregularity of the rock surface is the result of a differential weathering process common in the Piedmont (Reference 55). Differential weathering is mainly controlled by composition and by the frequency and orientation of joints.
Where joints are closely spaced, more groundwater may enter and chemical weathering preferentially occurs. Differential weathering in the vicinity of the site is accentuated by the local presence of mafic dikes that are more weatherable than the surrounding adamellite. In areas where the mafic dikes are deeply weathered, increased weathering locally affects the adjacent adamellite.
2.5.1.2.2.5 Geology of Excavations The geology of the Powerhouse excavation and the S.N.S.W. Pond Dam excavation has been mapped in detail at scales of 1 inch equals 5 feet (Reactor-Auxiliary Building area, Figure 2-103) and 1 inch equals 10 feet (Turbine-Service Building area, Figure 2-104 and S.N.S.W. Pond Dam, Figure 2-105). The rocks encountered in these excavations are discussed in Section 2.5.1.2.2.3. The geologic events illustrated by features in the excavations are discussed in Section 2.5.1.2.2.1 and summarized in Table 2-83.
To obtain additional information on the structural and metamorphic history of the site, several structural elements listed below were measured in the excavations using a Brunton-type compass. The data are analyzed by plotting the poles on the lower hemisphere of an equal-area net (point plot) and contouring by computer (density plot) where a sufficient number of data points is present. These data are presented on Figure 2-101 for the Powerhouse excavation and on Figure 2-102 for the S.N.S.W. Pond Dam. In the listing of structural elements that follows, the numbers of observations are given in parentheses; first, those from the Powerhouse and second, those from the Dam.
- 1. Foliation (13-15)
- 2. Joints (1192-347)
- 3. Contacts of Pegmatite Dikes (63-15)
- 4. Contacts of Mafic Dikes (30-3)
- 5. Shear Planes (85-64)
The total number of structural readings listed above is 1827. The 639 structural readings made during an intense study of Zones 2 and 3 are not included here because of the bias that might be introduced by such a concentration of data. That study is described in detail in Section 5.3 of the Final Geologic Report on Brecciated Zones.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 17 Measurements of foliation in the Powerhouse excavation show a good deal of scatter (Figure 2-101, Sheet 1). As shown on Figure 2-102, Sheet 1, the maxima of readings of foliation in the S.N.S.W. Pond Dam excavation indicate a strike in the range of N30°E to N55°E and a dip that is very steep to vertical. This is consistent with the mean of 80 readings of schistosity and foliation reported from the Charlotte belt in York County: N44°E, 72°SE (Section 2.5.1.2.1.2 and Reference 44).
Examination of the joint diagrams (Figure 2-101, Sheet 2 and Figure 2-102, Sheet 2) indicate two main trends: N30°E to N45°E and N35 °W to N50°W. Both are steeply dipping: 65° to 85° southeast and 85° southwest to 85° northeast, respectively. These trends are similar to regional patterns of northeast-and northwest-striking joints which are developed in the older igneous and metamorphic rocks and younger Triassic sedimentary rocks and Mesozoic diabase (Reference 40).
The two prominent maxima of pegmatite orientations in the Powerhouse excavation and shown on Figure 2-101, Sheet 3 are N68°E, 85° northwest and N82°E, 65° north. The point plot shown for pegmatites in the S.N.S.W Pond Dam excavation (Figure 2-102, Sheet 3) indicates a more northwesterly trend. These data are consistent with those of the earlier, localized study (Brecciated Zones Report, Section 8, Figure 10, Sheet 4).
Readings on contacts of mafic dikes in the Powerhouse excavation (Figure 2-101, Sheet 4) indicate three dominant attitudes: N05°E to N15°W, 45° to 55°east; N20°W to N45°W, 55° to 70° northeast; N45°W to N50°W, 85°southwest to vertical. No definite trends of mafic dike contacts can be established in the readings for the S.N.S.W. Pond Dam excavation (Figure 2-102, Sheet 4) because of the scanty data.
The results of structural measurements of shear planes are shown on Figure 2-101, Sheet 5 (Powerhouse) and Figure 2-102, Sheet 5 (Dam). The two most prominent trends are common to both areas of the site: N-S to N15 °W, 80° west to vertical; N40°W to N45°W, 80° southwest to vertical. The fact that the northwest trend of shear planes in the Powerhouse excavation appears essentially parallel to one of the strong mafic dike trends (Figure 2-101, Sheet 4) is coincidental. The observation made during geologic mapping is that the long shear zones tend to follow the north to north-northeast trending mafic dikes (Brecciated Zones Report, Section 5.3).
Several general observations can be noted about the pattern of shear fractures in the Powerhouse excavation:
- 1. About 90% of the shear fractures have left-lateral strike-slip displacements (apparent offsets viewed in a horizontal plane). About 10% have right-lateral displacements.
- 2. Shears occur predominantly in four orientations: north-northeast, north, north-northwest and northwest.
- 3. The north and north-northeast-trending sets are distinctly longer than the north-northwest and northwest-trending sets.
- 4. The north, north-northwest and northwest trending sets contain more numerous individual shear planes than the north-northeast-trending set (thus, the distribution on Figure 2-101, Sheet 5).
Although the fracture pattern that occurs in the S.N.S.W. Pond Dam excavation is much less dense (perhaps because it is obscured by weathering), it has similar characteristics.
At Catawba Nuclear Station, the angular relationship between faults that could be considered to be primary and those that could be considered to be subordinate does not fit the pattern that would be theoretically expected for a system of wrench faults. Theories of the development of
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 18 (09 OCT 2016) structural patterns of strikeslip (or wrench) faults are based on the existence of brittle conditions.
That assumption is not valid for the Catawba site because the pattern of faulting was established during the waning stages of intrusive activity (Table 2-83, Event ID) when partly plastic conditions prevailed. The largest faults (by virtue of length and/or amount of apparent horizontal displacement) on the site parallel contacts of mafic dikes and include in their histories reactivation of faulting during later, brittle stages in the geologic history of the site (Table 2-83, Events II and IV).
As documented in the Final Geologic Report on Brecciated Zones, faulting along most shear-breccia zones ended during or shortly after the imposition of greenschist metamorphic conditions, about 300 million years ago. Some zones underwent later reactivation of faulting and subsequently were healed by hydrothermal minerals. Laumontite is part of a major mineral assemblage of the hydrothermal episode that postdates all faulting at the Catawba site.
Because of its association with Mesozoic diabase dikes (Section 2.5.1.1.2.1) the age of the laumontite is considered to be about 150 million years. A sample of laumontite obtained from rock core at the site was dated at 86+/-30 million years by the potassium-argon method (Section 5.10.2 of Brecciated Zones Report). This is a minimum age of all faulting at the Catawba site.
Shearing and brecciation along the faults at Catawba Nuclear Station occurred under geologic conditions that are different from those currently existing at the site. The last movement along the shear-breccia zones occurred at least 56 million years ago and probably closer to 150 million years ago. Therefore, these zones cannot be considered to be "capable faults" under the criteria of Appendix A to 10CFR Part 100.
2.5.2 Vibratory Ground Motion 2.5.2.1 Seismicity The epicenters of all the reported earthquakes within 200 miles of the site having Modified Mercalli (MM) intensities of IV-V or greater have been plotted on Figure 2-106 and are listed within Table 2-84. Also shown on the figure and listed in the table are all earthquakes with magnitudes of 3.0 or greater, regardless of their felt intensity. The table includes several microearthquakes (magnitudes less than 3.0); these events were either not felt or felt with intensity IV or less. There have been 13 recorded earthquake epicenters within a 50-mile radius of the site having an intensity of III or greater. There have been no felt earthquakes and no instrumented events of meaningful magnitude (large enough to be identified readily as natural, not man-made such as quarry blasts) within this 50-mile radius since March 7, 1975, the date of the last entry in Table 2-85. These are plotted on Figure 2-107 and listed in Table 2-85.
Epicenters of all earthquakes which were probably felt at the site are shown on Figure 2-108 and included in Table 2-86.
The largest earthquakes ever recorded in the southeastern United States are the New Madrid earthquake series of 1811-1812 (December 16, 1811, and January 23 and February 7, 1812) and the Charleston, South Carolina earthquake of August 31, 1886. New Madrid, Missouri, and Charleston, South Carolina, are located 500 and 160 miles from the site, respectively.
The epicentral intensity during the New Madrid events is estimated to have been XII MM, representing total destruction at the epicenter. These earthquakes shook an area of at least 2,000,000 square miles and caused topographic changes over an area of 30,000 to 50,000 square miles. Only a very small amount of damage was reported, probably due to the sparse population at that time. The duration of motion at distances of 100 miles and greater (which includes the Catawba site) may have been as much as 1 to 2 minutes (Reference 56). The intensity at the distance of the Catawba site is estimated to have been IV to VI, MM.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 19 The epicentral intensity of the Charleston earthquake is reported to have been X MM. The earthquake was reported to have been felt over an area of 2,000,000 square miles (Reference 57). The Charleston earthquake is estimated to have been felt in the vicinity of the Catawba site at intensities of VI-VII MM.
On May 31, 1897, a moderately large earthquake occurred in Giles County, Virginia, about 155 miles from the site and in the Valley and Ridge Geologic Province. This earthquake has a reported epicentral intensity of VII-VIII MM (Reference 58). The Giles County earthquake is estimated to have been felt at the site at about intensity IV MM.
The Union County, South Carolina earthquake of January 1, 1913, has a reported epicentral intensity of VII MM. The epicenter is located about 45 miles from the Catawba site and probably was felt at the site at about intensity V-VI MM (Reference 59).
2.5.2.2 Geologic Structure and Tectonic Activity A map on which the major tectonic structures in the region surrounding the site are shown is presented on Figure 2-73. Geologic structures within the area of the site (the Subregion) are shown on Figure 2-73. As indicated on those figures, the site is located on or near the axial trace of a northeasttrending syncline. Axes of other folds are located 3-1/2 miles northwest (Figure 2-75), 10 miles southeast and 19 miles northwest of the site (Figure 2-73). Folding ended at least 300 million years ago (Section 2.5.1.2.1.4).
The nearest regional fault structure is the Gold Hill-Silver Hill fault complex whose projection into South Carolina is 11 miles southeast of the site. The Gold Hill Fault is cut by and is, therefore, older than Triassic diabase dikes as discussed in Section 2.5.1.1.2.1.5.
There is no geologic evidence of surface faulting within the Piedmont or adjacent geologic regions that is even remotely related to earthquakes that have occurred in historic time. This conclusion is supported by a published review of historic surface faulting in the continental United States and adjacent parts of Mexico (Reference 60).
2.5.2.3 Correlation of Earthquake Activity With Geologic Structure No correlation of earthquake epicenters to geologic structure is known in this region. Only in the past few years has there been instrumentation in the eastern United States to determine scattered earthquake locations and focal depths.
An interesting aspect is the grouping of epicenters. It may be characteristic in the southeastern United States for seismic energy releases to occur in areal patterns that are in the form of somewhat isolated "clusters" (Reference 61). Examples of such clusters are the Giles County, Virginia area, the Tennessee-North Carolina border area paralleling the Blue Ridge, the Charleston area, and the New Madrid area.
Because epicentral locations cannot be correlated with tectonic structures in this region, earthquakes are identified with the tectonic province in which they are located. According to Appendix A to 10CFR 100, a tectonic province is characterized by the relative consistency of the geologic structural features within the province. Tectonic provinces for the eastern United States are individually discussed in the following sections.
2.5.2.3.1 Piedmont and Upper Coastal Plain The Piedmont and Upper Coastal Plain seismotectonic region extends from Alabama to New York and is a deeply eroded, plateau-like segment of the Appalachian mountain system. The eastern boundary includes projections of the easternmost exposed Piedmont rocks beneath
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 20 (09 OCT 2016)
Coastal Plain sediments. The western limit is the Brevard Zone which extends from Alabama to Virginia along the eastern edge of the Blue Ridge geologic province.
The Catawba site is located within this seismotectonic region. There have been five historic earthquakes with intensity VII experienced within the region. Of these earthquakes, only two occurred in the southern Piedmont. One was the Union County, South Carolina earthquake of January 1, 1913, located some 45 miles southwest of the site. The other was at Arvonia, Virginia on December 22, 1875, located 240 miles from the site. The Arvonia event was probably not felt at the site. Both Union and Arvonia are underlain by thick residual soils derived from the weathering of the underlaying crystalline rocks. At both towns, the minimum rock depth is about 20 feet; the maximum is about 70 feet. Thus, considering the amplifying effect of the residual soil on which the structures are founded, the bedrock accelerations must have been somewhat less than those at the surface, where the intensity observations were made.
2.5.2.3.2 Blue Ridge The Blue Ridge seismotectonic region encompasses the Blue Ridge physiographic province and extends from Georgia to Pennsylvania. The eastern boundary is the Brevard Zone and wester boundary generally coincides with the major thrust faults of the Valley and Ridge geologic province.
The February 22, 1874 McDowell County, North Carolina earthquake is reported to have had epicentral intensity VII, the historic maximum for the Blue Ridge. It was located some 75 miles northwest of Catawba and was probably not felt at the site's distance.
2.5.2.3.3 Charleston Epicentral Area High historical seismic activity consisting of over 400 events distinguishes the Charleston Epicentral Area from the remainder of the Coastal Plain. Based on seismicity data, the Charleston locality is unique.
The largest historic event is the August 31, 1886 Charleston earthquake, which exhibited epicentral intensity X, and was felt over an area of about 2 million square miles. The site, at a distance of approximately 163 miles, experienced ground motion of intensity VI-VII from this earthquake. This is the maximum earthquake intensity documented for the Catawba site. There is presently no reason to postulate a recurrence of an earthquake similar to the 1886 Charleston event outside the vicinity of Charleston.
The seismotectonic regions for the lower Coastal Plain are established on the basis of the spatial distribution of historic seismicity. The Charleston Epicentral Area seismotectonic region is defined to include the zones of concentrated historic seismicity in the lower Atlantic Coastal Plain.
The Charleston Epicentral Area includes two zones exhibiting frequent small earthquakes. The first epicentral zone is northwest of Charleston in a zone extending from Middleton Place to Summerville; the second zone is located farther to the north-northwest near Bowman, South Carolina. The epicentral zone near Charleston is about 163 miles from the Catawba site; the zone near Bowman is about 115 miles from the Catawba site.
At present, the cause of seismicity in the Charleston Epicentral Area seismotectonic region is uncertain. However, knowledge of the geologic setting is increasing. Until a definition explanation of the Charleston earthquake is agreed upon, the present definition of the seismotectonic region on the basis of past seismicity represents the state-of-the-art for delineating earthquake source areas in the region.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 21 2.5.2.3.4 Deformed Appalachian Highlands The Deformed Appalachian Highlands seismotectonic region consists of the Valley and Ridge geologic province and the eastern portion of the Appalachian Plateau. The Valley and Ridge is characterized by narrow, elongated ridges and intervening valleys which trend in a northeast-southwest direction. The exposed sequence of Lower Paleozoic sedimentary rocks is typically repeated by thrust faulting. The Appalachian Plateau is a region of underformed sedimentary rocks consisting primarily of Paleozoic clastics.
The largest earthquake is the May 31, 1897 Giles County, Virginia event, which exhibited epicentral intensity VII-VIII. This earthquake, located 155 miles north of Catawba, produced ground motion of Intensity IV at the site.
2.5.2.3.5 New Madrid Faulted Zone The New Madrid Faulted Zone is approximately coincident with the northern section of the Mississippi Embayment, which consists of Cretaceous and younger Coastal Plain sediments.
Some of the largest historic earthquakes in the United States has epicenters within the New Madrid Faulted Zone, and occurred during the winter of 1811-1812. The largest event of this series has been assigned an epicentral intensity of XII. The site experienced ground motion of intensity VI due to this earthquake.
2.5.2.3.6 Central Stable Region The large area bounded on the east and south by the Deformed Appalachian Highlands and the New Madrid Faulted Zone comprises the Central Stable region.
The largest historic event was on August 12, 1929 in Attica, New York, with epicentral intensity VIII. Because the maximum intensity is only VIII and would be at considerable distance from the site, the Central Stable region has no influence on the maximum seismic intensity at Catawba.
2.5.2.3.7 Florida Platform and Lower Coastal Plain This seismotectonic region consists of the southern Georgia Coastal Plain, the Florida Peninsula and the Florida continental shelf. The seismicity of this large area is low and the largest event only exhibited epicentral intensity VI. Earthquakes in this region would have no effect on the site.
2.5.2.4 Maximum Earthquake The historical records indicate that the greatest earthquake intensity experienced at the site was from the Charleston earthquake of August 31, 1886, with an estimated site surface intensity of VI-VII MM (Table 2-86).
The maximum earthquake intensity which has occurred in historic time within the southern Piedmont, the tectonic province in which the site is located, is VII MM. This maximum occurred at Union, South Carolina on January 1, 1913, and Arvonia, Virginia on December 22, 1875.
These two earthquake epicenters are about 45 miles and 240 miles from the site, respectively.
The Arvonia event was probably not felt at the site.
The regional geologic structure and ancient faults previously discussed have not been active since Triassic time. The historical record of earthquakes in the southeastern United States for the past 200 years indicates that the ancient geologic structure and faults are not related to the historic earth quakes. Earthquakes in the Piedmont, particularly within the vicinity of this site, have occurred at scattered locations. Therefore, since there is an absence of geologic structure
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 22 (09 OCT 2016) that can be related to earthquakes, it is conservative to assume that these observed epicentral intensities (maximum of VII MM) could occur anywhere within the region or even in the immediate vicinity of the site.
The largest intensity, VII MM, within the tectonic province containing the site (the Piedmont and Upper Coastal Plain) is not exceeded by assuming that the largest earthquake associated with each surrounding tectonic province occurs at the point of closest approach of that province to the Catawba site. This is indicated in the following tabulation.
Tectonic Province Approximate Closest Approach to Catawba Site Largest Historic Earthquake Expected Site Intensity New Madrid 500 mi.
XII, MM VI(2)
Deformed Appalachian Highlands 100 mi.
VII-VIII, MM IV-V(1)
Blue Ridge 70 mi.
VII, MM
<IV(1)
Piedmont Contains Site VII, MM VII(2)
Charleston 163 mi.
X, MM VI-VII(2)
Notes:
- 1. Assuming largest historic earthquake within tectonic province occurs at the closest approach of that province to the site, and using attenuation function of McGuire (Reference 62):
Is = Io + 2.44 - 3.08 LOG D (D>6 miles)
Is = Io (D6 miles)
Is = expected site intensity Io = epicentral intensity D = epicentral distance, miles
- 2. Historic Data Based on the above information, the set of conditions describing the largest vibratory ground motion at the site would be an earthquake occurring in the immediate vicinity of the site and producing the historic maximum intensity VII for the Piedmont tectonic province. Since there is an absence of geologic structure in the Piedmont that can be related to earthquakes, there is no available historic information relative to the nature of the faulting that would be associated with this earthquake. As described in Section 2.5.3 below, faulting on the site ended at least 56 million years ago and more likely 150 million years ago.
2.5.2.5 Seismic Wave Transmission Characteristics of the Site Static and dynamic engineering properties of the soil and rock materials that underlie the site are discussed in Section 2.5.4. Design response spectra that include considerations of the thickness and distribution of these materials are discussed in Section 2.5.2.8.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 23 2.5.2.6 Safe Shutdown Earthquake As stated in 10CFR Part 100, Appendix A, due to the limited historical data the most severe earthquake associated with the tectonic province (in which the site is located) is determined in a conservative manner and is usually larger than the maximum earthquake historically recorded.
There have been no reported earthquakes within historic times with an intensity of more than VII MM in the Piedmont. The Charleston earthquake of August 31, 1886, produced surface intensities of only VI-VII MM at the site. Therefore, the Safe Shutdown Earthquake for the site is based on an earthquake producing surface intensity of VII-VIII (7.5) occurring adjacent to the site. This is greater than the surface intensity of any earthquake within the Piedmont during historic time, and is greater than the surface intensity at the site from the Charleston earthquake of 1886.
The Richter relationship for intensity-acceleration is based on monitored seismograph records of earthquakes in California. The recordings are from instruments founded on both rock and non-indurated materials; however, the readings on non-indurated materials are adjusted to the values they would be on rock (References 63, 64). This relationship is for moderate intensity earth quakes at rock sites. The Richter acceleration-intensity relationship and the range of accelerations proposed by Coulter, Waldron and Devine (Reference 67) for rock sites are shown on Figure 2-109 and are the bases for choosing the SSE acceleration. The acceleration-intensity relationship proposed by Hershberger (Reference 65) and Brady (Reference 66) are also shown on Figure 2-109.
As stated in Section 2.5.4.8, all major Category I Powerhouse structures are supported on rock.
The criteria for defining rock at this site are also discussed in that section. At a few locations, the top of continuous rock is below the design bottom of the substructure mat of significant structures. At those locations, fill concrete is placed to extend from the top of continuous rock up to foundation grade.
The acceleration value for the above described Safe Shutdown Earthquake, chosen for foundations on closely jointed rock and slightly weathered rock, is 0.15g. This bedrock value relates very conservatively with the design surface intensity VII-VIII MM considering the maximum observed surface intensities of VII in the region and the overburden amplification that contributed to those maximum observed surface intensities.
The historic earthquake which is assumed to be the largest earthquake to occur in the Piedmont is VII MM. This is larger than the historic maximum surface intensity at the site, which is VI-VII MM from the August 31, 1886 Charleston earthquake.
The NRC position allows the maximum earthquake determined for the seismotectonic province to coincide with the Safe Shutdown Earthquake. On this basis, the acceptable Safe Shutdown Earthquake for Catawba would be VII MM. (The Safe Shutdown Earthquake actually assumed for Catawba is VII-VIII MM, or 1/2 intensity unit higher than the maximum historical epicentral intensity experienced in the site seismotectonic region). For intensity VII MM, the "Trend of the Mean" relationship of the data of Trifunac and Brady (Reference 66) yields 0.12g.
The design response spectra are discussed in Section 2.5.2.8.
2.5.2.7 Operating Basis Earthquake Appendix A of 10CFR 100 provides that the Operating Basis Earthquake (OBE) shall be specified by the Applicant and shall be defined by response spectra. The OBE acceleration value recommended for foundations on closely jointed and slightly weathered rock is 0.08g. This is a conservative value, and is slightly more than one-half the SSE.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 24 (09 OCT 2016) 2.5.2.8 Design Response Spectra Four synthetic earthquake records with maximum accelerations of 0.15g are developed to generate the response spectra for the Safe Shutdown Earthquake. In simulating the earthquakes, a maximum duration of 20 seconds is used in the model, of which 0 to 2 seconds is used for the rising period, 2 to 7 seconds for the constant acceleration period, and 7 to 20 seconds for the receding period. The shape of the response spectra of the simulated earthquakes for a single degree of freedom approximates the "Spectrum Curve" discussed by Newmark (Reference 68).
The numerical average of the response of the four earthquakes is used to generate the response spectra for 0.5, 1.0, 2.0 and 5.0 percent damping. The design response spectrum is a smoothed curve drawn through the averaged spectra. Figure 2-110 gives comparisons of the smoothed and the averaged spectra for 0.5, 1.0, 2.0 and 5.0 percent damping. Figure 2-111 and Figure 2-112 give the smoothed response spectra for 0.5, 1.0, 2.0 and 5.0 percent for 0.08g and 0.15g ground acceleration, respectively.
A composite ground response spectra was approved by the NRC staff for the Catawba specific snubber reduction program. This composite spectra was constructed using the above design response spectra and a site specific spectra provided by the staff. A plot of the composite input response spectra is shown in Figure 2-204.
Certain Category I tanks and other small structures are not founded on "continuous" rock as described in 2.5.4.10. These structures are designed using the design response spectra for 25 ft of soil (fill) above rock as shown on Figure 2-113. The spectra on Figure 2-113 are computed using a lumped mass model; the mathematics for the method are described in standard texts such as Chapter 2 of Newmark and Rosenblueth (Reference 111). The soil parameters used in the lumped mass model are shear modulus, Gused, equals 576 KSF, and soil damping, Dused at 10 percent of critical. These parameters are equivalent to strain-corrected soil properties (Gused and Dused) obtained when the program SHAKE (Reference 112) and the equivalent linear method are used to perform amplification analyses.
The subsurface conditions at structures listed in Table 2-90 and for which the amplified smoothed design response spectra on Figure 2-113 are used for design are summarized on Figure 2-114. As can be seen from Figure 2-114, the site conditions are such that only the Unit 2 Diesel Fuel Oil Tanks and the Unit 2 Above Ground Storage Tank actually justify consideration of a design response spectrum other than the one for rock on Figure 2-112.
However, all the structures named on Figure 2-114 were designed using the amplified response spectrum on Figure 2-113.
For the Unit 2 Diesel Fuel Oil Tanks and Unit 2 Above Ground Storage Tank, horizontal response spectra are calculated for the specific soil columns at the structure for comparison to the smoothed design spectrum actually used and shown on Figure 2-113. The program SHAKE is utilized for these calculations using the soil parameters shown on Figure 2-115. The strain dependency of shear modulus and damping of all the soil and sand in the soil columns is assumed to conform to the average curve for sand found in Seed and Idriss, 1970 (Reference 108). The strain dependency of the shear modulus of the compacted crushed stone backfill is calculated using information contained in Reference 113. The shear wave velocity (shear modulus) at low strain of the compacted crushed stone is computed from information contained in Reference 113. The damping in the crushed stone is assumed to be about 15 percent less than the average sand damping at the equivalent confining pressure and shear strain.
The numerical average of the response due to input of each of the four synthetic earthquakes at an exposure of the rock-like material is used to plot the response spectra for actual site conditions as shown on Figure 2-116. The smoothed response spectrum for which the
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 25 structures are designed is also shown on Figure 2-116 for comparison with the response spectra computed for actual site conditions.
The computed response spectra for the actual site conditions at the Unit 2 Diesel Fuel Oil Tanks is below the design response spectrum except for periods above 0.4 seconds. The fundamental period of these tanks is less than 0.4 seconds. The computed response spectrum for the Unit 2 Above Ground Storage Tank is below the design response spectrum for periods less than 0.7 seconds. The fundamental period of the Above Ground Storage Tank is less than 0.7 seconds.
Thus, the design spectrum exceeds the computed spectra for actual site conditions within the period range of both these facilities and the use of the design response spectrum (25 ft of overburden) is conservative for the design.
As stated in Section 2.5.4.8, the granular and earth backfill, partially weathered rock and rock are not susceptible to liquefaction resulting from dynamic loading of the SSE.
The smoothed spectrum of ground motions (not a response spectrum) which is used for analyzing the buried piping for seismic design adequacy (referenced in Section 3.7.3.12) is compatible with the SSE response spectra for rock on Figure 2-112 when spectral amplification factors are considered. The spectrum of ground motions if found on Figure 11 of Newmark, et al., 1973 (Reference 114) and is scaled to 0.15 g to represent the ground motions for the SSE on rock at Catawba. For the pipeline in areas underlain by residual soil or compacted fill, the ground motion spectrum for 0.15 g is scaled upward by an assumed conservative value of 0.40/0.15 (equals 2.66 times), at all periods, to represent the motion to which the pipeline is assumed to be subjected (see Section 3.7.3.12). The data for the Unit 2 Diesel Fuel Oil Tanks (maximum ground acceleration 0.25 g at the pipeline depth) on Figure 2-115 and the acceleration at low period (0.27 g at the ground surface) on the design spectrum for 25 ft of soil (Figure 2-113) show computed maximum ground accelerations considerably less than the 0.40 g assumed for pipeline design in these areas. The small thicknesses of residual soil/saprolite beneath the pipeline (refer to Figure 2-136) indicate little amplification of ground acceleration would occur in these locations. Thus, the ground motion spectrum used to analyze the buried pipelines for design adequacy was quite conservatively selected. Where partially weathered rock is the material through which the pipelines passes, the data for the Unit 2 Above Ground Storage Tank on Figure 2-115 indicate the computed maximum ground acceleration at the pipeline depth is 0.17 g. Therefore, the pipeline would be subjected to less motion in partially weathered rock areas than was assumed for the above computations in fill and residual soil.
Table 2-90 lists the foundation design condition and the response spectrum used for each Category I structure.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE UPDATED 2.5.3 Surface Faulting There is no geologic evidence of (capable) surface faulting in the Piedmont, the tectonic province in which the site is located. Therefore, a design basis for surface faulting is not applicable to this site.
2.5.3.1 Geologic Conditions of the Site The geologic conditions in the region surrounding the site and of the site itself are discussed in Section 2.5.1.1 and 2.5.1.2, respectively.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 26 (09 OCT 2016) 2.5.3.2 Evidence of Fault Offset Shear-breccia zones occur on the site and are discussed in detail in the Final Geologic Report on Brecciated Zones. As noted in that report, the maximum total offset (net slip) calculated is 23 ft ("Zone 2") and the maximum apparent offset measured is 17 ft ("Zone 7"). It is concluded from the evidence presented in the Brecciated Zones Report that faulting on the site ended at least 56 million years ago and more likely 150 million years ago.
2.5.3.3 Earthquakes Associated With Capable Faults There is no evidence for the existance of capable faults within 5 miles of the site. There is no correlation between epicenters of nearby earthquakes (Figure 2-107) and subregional tectonic structures (Figure 2-73).
2.5.3.4 Investigation of Capable Faults There are no capable faults within 5 miles of the site (Section 2.5.1.2.1.4 and Final Geologic Report on Brecciated Zones).
2.5.3.5 Correlation of Epicenters With Capable Faults There are no capable faults in the region surrounding the site (Section 2.5.1.1.2.1.5 and Section 2.5.1.2.1.4) and there is no correlation between the locations of earthquake epicenters and regional tectonic structures (Section 2.5.6.4.4).
2.5.3.6 Description of Capable Faults There are no capable faults within 5 miles of the site.
2.5.3.7 Zone Requiring Detailed Faulting Investigation There is no (capable) surface faulting in the region surrounding the site. Therefore, this criterion is not applicable.
2.5.3.8 Results of Faulting Investigation A specific zone of investigation is not required (Section 2.5.6.4.4). Results of the investigations of shear-breccia zones on the site are presented in the Final Geologic Report on Brecciated Zones.
2.5.4 Stability of Subsurface Materials and Foundations 2.5.4.1 Geologic Features The bedrock at the site consists primarily of adamellite which is a metamorphosed igneous rock of the Charlotte belt. The adamellite is a medium grained crystalline rock with faint foliation and uniform texture and mineralogy. The bedrock also includes a secondary rock type in the form of discontinuous and irregular mafic dikes within the adamellite. The mafic dikes are fine grained rocks consisting of predominantly dark colored minerals. Metamorphism has imparted a moderate foliation to the mafic rock. The soils overlying the bedrock are primarily residual soils formed by the in-place chemical weathering of the bedrock. Plant structures are founded on rock or residual soil as discussed in Section 2.5.4.10. Alluvial soils occur in the drainage swales at the plant site. Geologic features of the site are discussed in detail in Section 2.5.1.2. The subsurface conditions are discussed in Section 2.5.4.2.3.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 27 Conditions that might lead to uplift, subsidence or collapse do not exist on the site. Weathering has advanced with erosion since the last orogenic episode several hundred million years ago.
The site area was not glaciated during the Pleistocene. No subsurface mining or withdrawal of fluids (other than in shallow water wells) exists or has taken place in the site area. Groundwater withdrawal is limited to fractures in the crystalline bedrock. The distribution and characteristics of faults and joints on the site are discussed in Section 2.5.1.2.2 and in the Final Geologic Report on Brecciated Zones. Geologic conditions that could produce shearing or brecciation have not existed in the site area for at least 150 million years. There are no indications of unrelieved residual stresses in the bedrock on or near the site.
The crystalline bedrock at this site is not subject to long-term deterioration or solution activity.
The foundation rock for the nuclear safety related structures will not provide adverse response to seismic activity (see Section 2.5.2). Further, the residual soils and underlying crystalline bedrock are such that liquefaction is not a problem (see Section 2.5.4.8). Alluvial soils found in drainage swales at the site generally are unconsolidated. No Category I structure foundations bear on alluvium.
2.5.4.2 Properties of Subsurface Materials A detailed field investigation and comprehensive laboratory testing program are made to explore subsurface conditions and to determine the static and dynamic engineering properties of the materials underlying the site. The field investigation and laboratory testing program are presented in Section 2.5.4.2.1 and Section 2.5.4.2.2, respectively. The subsurface conditions and engineering properties of subsurface materials determined by the field investigation and laboratory testing program are discussed in Sections 2.5.4.2.3 and 2.5.4.2.4. Engineering properties which pertain specifically to foundation materials and embankment materials at the SNSW Pond Dam are discussed in Sections 2.5.6.2 and 2.5.6.4.
2.5.4.2.1 Field Investigation The field investigation, including methods of exploration and quantity, extent, and location of all borings and explorations, is described in the following paragraphs.
2.5.4.2.1.1 Test Borings and Sampling Approximately 160 test borings were drilled on the Catawba Site. The locations of borings at the site are shown on Figure 2-78. Figure 2-79 shows the detailed boring locations, together with original ground surface contours in the area of the powerhouse. Figure 2-144 is the boring location plan for the SNSW Pond Dam. Table 2-87 summarizes the elevations of the ground surface, top of continuous rock and groundwater level found at the individual boring locations.
Holes made by test borings were not plugged.
The test borings were advanced by rotary wash drilling and hollow-stem augers. Split-barrel sampling and penetration testing are performed in general accordance with ASTM D-1586.
The soils retained inside the split-barrel sampler are removed and visually examined and classified in the field by a soils engineer or engineering geologist. The samples are then stored in labeled, capped, glass jars. Some of these jar samples are transported to the laboratory for classification and moisture tests.
Undisturbed samples are obtained by the Shelby Tube Method in general accordance with ASTM D-1587. In hard soils and partially weathered rock, undisturbed samples are obtained with either a coring Pitcher barrel sampler or a Denison sampler. The Pitcher sampler resembles a double tube core barrel with a carbide tipped bit on the rotating outer barrel and 3-
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 28 (09 OCT 2016) inch O. D., 16-gauge, steel tubing as the replaceable inner barrel. The Denison sample is similar to the Pitcher sampler except that the non-rotating inner barrel is pushed into the ground by means of reaction against the drill rig, while the rotating outer barrel cores around the inner barrel.
Rock coring is performed in general accordance with ASTM D-2113. The rock core samples are identified by a geologist and the recovery measured. The recovery is the ratio of the sample length obtained to the depth of the drill run, expressed as a percent. A modified definition of recovery, the Rock Quality Designation (RQD), is also used. It is determined by summing only pieces of moderately hard or harder rock 4 inches or longer.
2.5.4.2.1.2 Test Pits and Trenches Visual examination of the near-surface soils is made and bulk samples are obtained for laboratory testing. The test pit locations are shown in Figure 2-78 and Figure 2-79. The soils in the borrow areas outlined in Figure 2-78 are explored specifically for use in construction of the SNSW Pond Dam (see Section 2.4.13.2.6). Summary results of the test pits are contained in Table 2-88.
Three deep test trenches are excavated in the powerhouse area as part of the geologic investigation for the PSAR and also are used to obtain bulk soil samples for laboratory testing.
The locations of these test trenches are shown in Figure 2-79.
2.5.4.2.1.3 Groundwater Measurements At selected borings, slotted PVC pipe is installed to prevent caving of the boring walls and to provide access to measure the groundwater level. The perimeter of the pipe is sealed at the ground surface to prevent entrance of surface water. All wash borings with PVC pipe are bailed to remove the drilling fluid and to permit the groundwater to stabilize at its natural level.
Groundwater measurements are made on a periodic basis after completion of each cased boring. An automatic water level recorder is used to continuously monitor groundwater levels at two locations (Borings A-33 and A-62). Groundwater data are presented in Table 2-87 and Figure 2-53, and contours of the preconstruction groundwater elevations are shown in Figure 2-
- 50. Locations of observation wells for monitoring of post-construction groundwater levels are shown in Figure 2-78 and Figure 2-79. A discussion of groundwater levels over the site is presented in Section 2.4.13.2.4.
2.5.4.2.1.4 In Situ Permeability Testing In-place permeability tests of soil and rock are performed in test borings in the plant area and at the SNSW Pond Dam. The tests are done according to recommendations for Field Permeability Tests in Boreholes contained in Reference 69. Figure 2-54 and Figure 2-55 show the schematic arrangement of field test equipment and present brief descriptions of the test procedures. Data reduction methods used are those found in References 69 and 70. The permeability data obtained in the plant area are recorded in Table 2-72 and Table 2-73. Data obtained at the SNSW Pond Dam are presented in Table 2-93 and Table 2-95. Permeability characteristics of the subsurface materials are discussed in Sections 2.4.13.2.5 and 2.5.6.6.1. Constant discharge pumping tests are also performed at two locations (near borings A-48 and A-85) and are discussed in Section 2.4.13.2.6.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 29 2.5.4.2.1.5 Geophysical Testing Geophysical techniques are used to determine wave velocities of both soil and rock and to supplement the boring program. The geophysical testing includes seismic refraction profiling and up-hole and cross-hole wave velocity measurements.
Prior to construction, approximately 9500 linear feet of seismic refraction traverses are made along the lines (Seismic Lines 1 through 4) shown in Figure 2-78. During construction, additional refraction profiling is done along the NSW pipe lines (Seismic Lines 5 through 9 in Figure 2-78),
in the foundation excavation for the Powerhouse (see Figure 2-117), and in the excavations for the NSW Pump Structure and NSW Intake Structure (see Figure 2-118).
Geophones are spaced at regular intervals along the seismic lines, and small explosive charges are detonated at selected locations to generate elastic waves which propagate through the subsurface media. A Dresser RS-4A 12 channel seismograph is used to record the compressional wave arrivals. Plots of the arrival times of the refracted waves from the higher velocity layers are made and propagational velocities and depths to these higher velocity layers are computed.
The interpreted results of the refraction profiling over the plant site (Seismic Lines 1 through 4) are shown on Figure 2-94 through Figure 2-97 Seismic profiles for NSW pipe lines (Seismic Lines 5 through 9) are shown on Figure 2-119, and seismic profiles for NSW Pump Structure and NSW Intake Structure are shown in Figure 2-118. Results of the seismic profiling in the power house area during construction are given in Figure 2-120.
Up-hole seismic velocity measurements are made prior to construction to determine the compression and shear wave velocities for the foundation materials in the plant area and at the SNSW Pond Dam. The test results from the plant area are given in Figure 2-98 and Figure 2-99.
A Dresser RS-4A 12 channel seismic recording unit is used for this investigation. Blasting caps are fired at different depths in each hole and compressional (P) and shear (S) waves are recorded by means of an array of both vertical and horizontal geophones located at the surface around the shot hole. Travel times of P and, where possible, S phases are plotted versus depth and the representative velocities for the materials encountered are taken from the inverse slopes of the lines connecting the points plotted. Test elevations are closely spaced (always equal to or less than ten feet).
Cross-hole testing for shear and compression wave velocities are made at two locations (near borings A-103 and A-105) on the axis of the SNSW Pond Dam, where the strata of primary interest for shear wave measurements are the residual soil and partially weathered rock which overlie the unweathered rock. Cross-hole results applicable to the SNSW Pond Dam are given in Figure 2-155 and Figure 2-156.
At each location selected for cross-hole velocity measurements, two or more borings are drilled.
Selected borings are used for generating energy and the others are used for monitoring the arrival of seismic waves. The results of verticality surveys of selected boreholes are used to accurately determine the distance between holes at depth.
Energy is input in rock and partially weathered rock by detonating zero delay seismographic caps in the energy holes. The timing break (starting time on the recording device) is produced by the blasting circuit. In soil the energy is introduced by using a hammer to drive a standard split spoon sampler attached to the end of steel drilling rod. The timing break is produced by a vertical velocity transducer attached to the steel drill rod above the surface of the ground just below the top of the rod. Measurements of the travel speed down the steel rod are made, and the total travel time corrected by the appropriate amount to yield the net travel time. The energy is received by a triaxially oriented set of transducers lowered into the adjacent receiving holes to
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 30 (09 OCT 2016) the same depth as the energy source. The triaxial device consists of one vertical exploration geophone and two horizontal geophones. A single component vertical transducer is used for the bottom 20 ft of depth at one test location at the dam site, due to inability to place the larger three component transducer to the proper depth in a rock core hole.
In most cases, two different systems are used simultaneously to record the output of the geophones. The first is the Dresser Model RS-4A 12 Channel portable seismograph. The second recording device is a Tektronix Model 434 dual trace storage oscilloscope. Two signals are input to the oscilloscope, one from the vertical geophone and the second from one of the horizontal geophones. The signals are recorded on the face of a storage cathode ray tube where they are photographed for a permanent record. The oscilloscope has a horizontal time sweep which has a range of from 0.0002 seconds per division to 5 seconds per division, accurate to three percent of the scale reading. This allows the accurate measurement of very short time intervals. At selected depths multiple tests are made to insure repeatability of the data.
The geophysical data for partially weathered rock foundation materials in the plant area are plotted in terms of the parameter K2 maximum on Figure 2-129; also shown on this Figure are the results of a dynamic triaxial laboratory test on partially weathered rock. The laboratory test on partially weathered rock forms a lower bound to the geophysical tests on the partially weathered rock. The K2 maximum values for partially weathered rock shown on Figure 2-129 exceed the typical published value of 70 for very dense sands (Dr=90 percent) tabulated in Seed and Idriss (1979), Reference 108.
The geophysical data for partially weathered rock and residual soil-saprolite foundation materials at borings A-103 and A-105 of the SNSW Pond Dam are plotted in terms of the parameter K2max on Figure 2-182. Two geophysical test data points on near-surface materials at boring A-103 which are removed in foundation preparation due to the standard penetration test value being less than 15 blows are indicated by the legend on the figure. Also shown on Figure 2-182 are the results of resonant column tests for a firm residual soil from FSAR Figure 2-157 (and laboratory tests for embankment soils).
In addition to the laboratory data from Figure 2-157, other laboratory tests on residual soil at the boring series A-103 are shown on Figure 2-183 in terms of Gmax versus confining pressure; Figure 2-183. also shows Gmax from the geophysical data at boring series A-103 and A-105 versus confining pressure. Figure 2-183 is prepared to more clearly show the relationship between the laboratory data, geophysical data, and design value used to represent the residual soil-saprolite in the foundation of the SNSW Pond Dam.
Figure 2-183 reveals the upper bound of the laboratory data is about at the lower bound of the geophysical tests when extrapolated to equivalent confining pressures. Figure 2-183 shows the average of the geophysical Gmax values exceed the average value of laboratory resonant column Gmax values by a factor of 70/30 = 2.3; this factor is compatible with general experience in the literature on the relationship between laboratory and geophysical Gmax values. The major reason for the laboratory Gmax values being less than the geophysical Gmax values in the foundation materials is probably the unavoidable sample disturbance that takes place in the process of obtaining undisturbed samples from borings.
Shown on Figure 2-183 is the expression relating Gmax and confining pressure used to compute the maximum shear modulus of residual soils-saprolite of the foundation with K2max = 70. The K2max = 70 value for the residual foundation soils-saprolites is equivalent to the published value for very dense sand as tabulated on Figure 2-182. This K2max = 70 relationship is the numerical average of the geophysical data points on Figure 2-183. Thus K2max = 70 is not a lower bound for the residual soil-saprolites of the foundation at borings A-103 and A-105 (excluding
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 31 geophysical tests on materials that are removed in the foundation preparation). The data on Figure 2-183 clearly justify the selection of the K2max value as an average, not a lower bound, to represent the firm residual soil-saprolites of the foundation of the SNSW Pond Dam.
2.5.4.2.1.6 Geologic Mapping Geologic mapping of the cleaned foundation excavations is done. The purpose of the mapping is to demonstrate that the brecciated zones and any other faults found to be present are non-capable as defined by Appendix A, 10CFR Part 100, and do not present a safety hazard to the plant. Mapping is at a scale of one inch equals ten feet in most areas (one inch equals 5 feet in the Reactor-Auxiliary Building area) and is supplemented by complete photographic coverage of the exposed foundation and by petrographic study of selected samples. The results of geologic mapping of the structure foundations and the SNSW Pond Dam foundation are discussed in Section 2.5.1.2.2.5 and in the Final Geologic Report on Brecciated Zones.
2.5.4.2.2 Laboratory Testing The various laboratory tests made to determine the engineering properties of soil and rock are discussed below.
2.5.4.2.2.1 Grain Size Tests Grain size tests are performed as described by ASTM D-421 and D-422. The results for soils from the plant area are presented in Figure 2-121. Grain size distributions for foundation soils and borrow soils applicable to the SNSW Pond Dam are shown in Figure 2-150, Figure 2-162, Figure 2-163, and Figure 2-164.
2.5.4.2.2.2 Soil Plasticity Tests - Atterberg Limits Plasticity tests are performed as described by ASTM D-423 and D-424. The tests are used to determine the Unified Soil Classification of the soils, the results of which are discussed in Section 2.5.4.2.4, Section 2.5.6.2.1, and Section 2.5.6.4.2.
2.5.4.2.2.3 Compaction Tests Standard Proctor (ASTM D-698) compaction tests are performed on representative bulk soil samples to determine their compaction characteristics, including their maximum dry density and optimum moisture content. The maximum dry density and corresponding optimum moisture content points obtained are summarized on Figure 2-122 for soils used for structural fill in the plant area. Test results for soils from borrow areas designated specifically for use in construction of the SNSW Pond Dam are presented on Figure 2-165.
2.5.4.2.2.4 Laboratory Permeability Tests Permeability tests are performed on selected undisturbed samples and on selected bulk samples of borrow soils for the SNSW Pond Dam that are remolded in a steel tube to 96 percent relative compaction, according to the standard Proctor test method (ASTM D-698).
Constant-head permeability tests are performed on some of these samples using equipment and methods similar to ASTM D-2434. Soil permeability results of foundation soils at the SNSW Pond Dam, as well as results for borrow soils, are presented in Table 2-102 and Table 2-103.
Laboratory variable-head permeability tests on undisturbed samples from the plant area are also performed and results of these tests are presented in Table 2-73.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 32 (09 OCT 2016) 2.5.4.2.2.5 Consolidation Test Consolidation tests are performed on selected sections of undisturbed samples and remolded samples. Testing is in general accordance with ASTM D-2435. The test results are presented in the form of pressure versus void ratio curves and consolidation rate data. A summary of representative curves is presented in Figure 2-123 and Figure 2-124 for compacted soils and in-place materials in the plant area and in Figure 2-154 and Figure 2-171 for soils at the SNSW Pond Dam.
A single section of each selected undisturbed sample and each selected remolded sample is trimmed into a disc, confined in a smooth stainless steel ring and sandwiched between porous plates in a fixed ring oedometer. It is then subjected to incrementally increasing vertical loads at a general load increment ratio of unity and the resulting deformations measured with a micrometer dial gauge sensitive to 10-4 inch. The test results are presented in the form of pressure versus void ratio curves.
2.5.4.2.2.6 Static Triaxial Shear Tests Each section of undisturbed sample selected for triaxial shear testing is measured, weighed, and portions of the soil removed for moisture and specific gravity determinations. From these data, the soil void ratios and weights per cubic foot are computed.
In addition to selected sections of undisturbed samples, selected bulk samples of potential borrow soils are remolded to the same geometry at 96 percent relative compaction. Triaxial shear testing is performed using equipment and procedures consistent with those described by Bishop and Henkel (Reference 71), Lambe (Reference 72) and Corps of Engineers (Reference 73). On an occasional undisturbed sample of limited length the "Dutch Method" or multistage method is used. The procedures of the multistage method are described by Bishop and Henkel (Reference 71).
If no drainage is allowed from the sample under the confining stress or the load to failure, the test is termed unconsolidated-undrained or quick (Q). Some samples are saturated under backpressure to simulate the pore water pressures that will exist. If drainage is allowed from the sample to equilibrium under the confining stress, but no drainage is allowed during the load to failure, the test is termed consolidated-undrained (R) and represents total stresses. If the pore water pressures developed during loading to failure are measured and the stresses then expressed as effective stresses, the test is termed (R) and is equivalent to a drained test (S).
Backpressure values in the range of 30 to 80 psi are used. The final degree of saturation is between 98 and 100 percent. The rate of axial loading in the CU triaxial tests is generally.01 to
.06 inches per minute.
The test results for triaxial shear tests are presented in the form of Mohr Diagrams. Three samples are used to define each Mohr envelope for remolded or compacted materials. Three samples are normally used to define each Mohr envelope for undisturbed foundation materials, however, only two samples are used in the exceptional case when three good quality undisturbed samples are not available for a test location. Summary test results for representative remolded and undisturbed samples applicable to the plant area are shown on Figure 2-125 and Figure 2-126. Figure 2-151, Figure 2-158, and Figure 2-167 through Figure 2-169 summarize the triaxial test results for foundation and embankment materials at the SNSW Pond Dam.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 33 2.5.4.2.2.7 Dynamic Triaxial Tests Several undisturbed specimens of soil, partially weathered rock materials and remolded samples are tested using specialized laboratory equipment to determine the behavior of these materials under dynamic loading.
Two types of dynamic triaxially confined compression tests are performed. Stress-controlled tests are run to evaluate dynamic failure potential of site soils, and other stress-controlled tests are run to evaluate dynamic soil modulus. Brief descriptions of the test procedures follow.
Failure-potential Tests: Triaxially-confined compression tests, incorporating cyclic dynamic loading under controlled stress conditions, are performed on saturated test specimens of both undisturbed and remolded samples. These tests are performed to evaluate the dynamic failure potential of the site materials.
Test specimens are either 1.4 or 2.9 inches in diameter and 6 inches in height. Remolded samples are compacted by static compaction to the desired 96 percent relative compaction at a molding moisture content wetter than the optimum. After placing the specimen in the triaxial chamber, a backpressure of 30 psi to 80 psi is applied to assure saturation prior to consolidation and subsequent shear testing. The final degree of saturation is, in general, between 98 percent and 100 percent. Time versus volume of water expelled measurements are recorded during consolidation under the designated confining pressure to assure satisfactory primary consolidation. In some tests, an axial deviator stress is also applied during consolidation, thus creating an anisotropic consolidation condition. Upon completion of primary consolidation, the drainage valves are closed and the predetermined, reversed cyclic, square-shaped, axial stress pulse is applied. Measurements of pore water pressure, strain, and axial load are simultaneously recorded throughout the tests.
Additional details on equipment and procedures for the tests of dynamic shear strength may be found in References 74 and 75. Data obtained from interpretation of these tests are presented in Table 2-98 and Figure 2-152 and Figure 2-153.
The equivalent number of uniform cycles of seismic loading was assumed based on information by Seed and Idriss as described in 1972 by Lee and Chan (Reference 94). Conclusions contained in Seed, et. al. (Reference 109) are that for earthquakes of the size of the SSE for Catawba, five equivalent uniform cycles of stress at 65 percent of the maximum stress provides an adequately conservative representation of such events (page 11 of Seed, et. al, 1975). Ten cycles loading were used in planning the cyclic triaxial tests and for determining the dynamic shear strength of the soil, which is thus conservative for the SSE.
Dynamic Modulus Tests: Stress-controlled, dynamic triaxially-confined compression tests are performed on undisturbed samples and remolded samples to obtain values of dynamic soil modulus.
Samples are saturated under backpressures and isotropically consolidated as described above for the stress controlled failure tests. Samples are then tested in an undrained condition under a cyclic axial deviator stress. Generally, the range of strains varies from about 0.01 percent to one percent. Tests are started at low loads. The cyclic load is maintained for 5 to 10 cycles. From these data the dynamic Young's modulus is then measured from the load-deformation. The shear modulus is then calculated from the following equation:
)
1(
2 E
G
where:
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 34 (09 OCT 2016)
G
= shear modulus (modulus of rigidity)
E
= Young's modulus
= Poisson's Ratio The value of Poisson's ratio used is chosen by engineering judgment. Normally, the value for soil ranges from.35 to.50 and is generally chosen as 0.50 for saturated samples. For this range of values, Poisson's ratio does not have a significant effect on the shear modulus.
The axial strains measured in the laboratory tests are converted to shear strains by the following relationship:
strain shear G
where is the shear stress, or one-half the cyclic axial deviator stress.
After these measurements are made, the sample is permitted to drain, and additional data are then obtained by repeating the test procedures at a higher deviator stress until strain levels on the order of one percent are reached. No damping values are computed from the cyclic triaxial tests. Damping values of soil are determined for the expected range of strains under seismic loading by use of the resonant column test as described below. The results of dynamic triaxial tests for modulus are presented in Figure 2-182, Figure 2-191, and Figure 2-192.
Resonant Column Tests: Resonant column tests are run on undisturbed and remolded solid cylindrical samples. Sample saturation procedures are similar to those for cyclic triaxial tests.
The results of resonant column tests provide dynamic shear modulus and damping values for soils at low strain levels. Generally, test results for strains on the order.001 to.01 percent are measured with the equipment used.
Further information on the equipment and procedures used in performing resonant column tests can be found in References 76 and 77. Results of resonant column tests are presented in Figure 2-157, Figure 2-182, Figure 2-190, Figure 2-191, Figure 2-192, and Figure 2-183.
The results for damping are shown in Figure 2-190. The relationship used in the analysis is shown on this figure and is the same as the published relationship for sands by Seed and Idriss, 1970. The test data on Figure 2-190 indicate this published relationship adequately represents the damping and its variation with strain for these soils.
2.5.4.2.2.8 Chemical Tests Chemical tests are performed on selected samples of soil and groundwater from the site. The chemical testing program and the results obtained are presented in Section 2.4.13.2.7.
2.5.4.2.2.9 Rock Unconfined Compression Tests Unconfined compression tests are performed on selected samples of NX size rock core from borings and also on 2-inch diameter samples cored from chunk samples of rock taken from the powerhouse excavation. The ends of the samples are cut perpendicular to the core axis with a masonry saw. The trimmed core ends are then lapped to insure smoothness or capped if lapping damages the specimen end. The specimens are then loaded as recommended in ASTM D-2938 in uniaxial unconfined compression to failure. Data for stress-strain readings are recorded during the axial loading of selected core samples. Summaries of rock test data are given on Figure 2-127 and Figure 2-128.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 35 2.5.4.2.2.10 Rock Unit Weight Unit weights are determined for most rock samples tested in unconfined compression. The samples are weighed, their volumes determined and the total unit weight is calculated. The results are presented on Figure 2-127 and Figure 2-128.
2.5.4.2.3 Generalized Subsurface Profile The general soil profile in the plant area is typical of residual soils that have weathered from crystalline bedrock. Underlying the organically stained topsoil is a thin soil stratum of fine grained red or tan sandy silts or clayey silts. This upper soil, formed by advanced weathering near the surface, quickly grades into saprolite, a residual soil that retains the relict micro-and macro-structure of the parent rock. The soil textures throughout the entire site consist of silty fine to medium sand weathered from the adamellite by far the predominant rock type-and the soils weathered from the basic dike rocks. The soil strength increases with depth, although the soil texture, silty fine to medium sand, remains essentially the same. Eventually, partially weathered rock, defined as having at least 100 blows per foot standard penetration resistance, is encountered. Still deeper, the degree of weathering is so small that the parent rock, adamellite, can be identified. This rock decreases in weathering with depth until moderately hard adamellite bedrock is encountered.
Some deposits of alluvial (water-deposited) soils of limited thickness and extent cover the in-place residual soils in the drainage swales at the plant site.
Figure 2-80 through Figure 2-93 illustrate subsurface conditions by way of penetration resistance and core boring results at the plant site. Subsurface conditions at the SNSW Pond Dam are discussed in Section 2.5.6.2.
2.5.4.2.4 Classification and Engineering Properties of Soil and Rock Classification and engineering properties of subsurface materials in the plant area are described below. Properties of foundation and embankment materials at the SNSW Pond Dam are presented in Section 2.5.6.
2.5.4.2.4.1 Alluvial Soils Alluvial silty sands are found in drainage swales leading to the lake. These sands have been washed into the swales from the surrounding higher ground. The alluvial soils are generally unconsolidated but are of limited thickness and extent. No Category I structures are constructed on alluvial soils.
2.5.4.2.4.2 Residual Soils There are two primary types of residual soils weathered from the parent adamellite bedrock.
The upper fine grained (more than 50 percent passing the number 200 sieve) soils consist of sandy silts and clayey silts which are stiff to very stiff in consistency, having standard penetration resistances of 10 to 30 blows per foot. These soils are generally found in the upper 1 to 5 ft zone and have a Unified Soil Classification of ML with an occasional clayey silt MH soil in the very upper layer. These fine grained soils generally have from 50 to 70 percent finer than the number 200 sieve with a liquid limit in the range of 35 to 44 and a plasticity index in the range of 8 to 16.
The deeper soils generally are a coarse grained (50 percent or less passing the number 200 sieve) silty fine to medium sand-saprolite. These materials have a Unified Soil Classification of
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 36 (09 OCT 2016)
SM, are generally of very low plasticity and have 10 to 30 percent passing the number 200 sieve. An enveloping plot of typical grain size distributions for several borings is shown on Figure 2-121. The saprolites were all weathered from the same parent adamellite bedrock formation.
Besides the soils derived by weathering of the parent adamellite bedrock there are some weathered seams of mafic rocks, in the form of numerous steeply dipping dikes, generally from 1 ft to 5 ft thick. Generally, the weathered dike material is a fine to medium sandy silt having a Unified Soil Classification of ML.
Permeabilities of the site residual soils are presented in Table 2-73 and discussed in Section 2.4.13.2.5.
In general, the residual soils described above have been excavated from beneath the Category I Structures during site preparation and have been used for backfill and for borrow in construction of the plant yard fills.
Duke's experience in the Piedmont region has been that residual soils of the type used for borrow material make excellent engineered fills when adequately compacted. Duke's normal practice is to compact such soils to a minimum of 95 percent of the standard Proctor maximum dry density for Group 1 fills, back fills, and dams. On most large earthwork jobs in the Piedmont the soils include major volumes of MH, ML, and SM classification groups. At the Catawba site, a compaction criterion of at least 96 percent of the standard Proctor was used due to the SM classification of the majority of the soil. Tests made on soil thus compacted showed excellent engineering properties and confirmed the adequacy of the 96 percent of minimum compaction.
The field tests performed and controls exercised to assure proper compaction of fills of the site soils are described in Sections 2.5.4.5.4.1, 2.5.6.4.5, and 2.5.6.4.9.
Bulk samples secured from test pits excavated in the residual soils in the plant structure area are compacted in accordance with the standard Proctor compaction test as described by ASTM D-698. Figure 2-122 summarizes the optimum moisture and corresponding maximum dry density of the borrow soils tested. The data in Figure 2-122 are divided into 3 groupings. This figure generally indicates that approximately 10 percent of the soils tested have a maximum dry density equal to or less than about 90 pcf. These light-weight soils are encountered in the upper fine grained soil material. A second grouping, consisting of about 30 percent of the samples tested (but constituting far less than 30 percent of the soil volume present at the site), has maximum dry densities in the range of 95 to 105 pcf. These soils generally consist of the weathered dike soils. The third group, comprising a majority of the soil samples tested, has maximum dry densities greater than 105 pcf. These materials are generally silty fine to medium sands, the predominant soil found at the site.
Figure 2-125 illustrates a summary plot of the Mohr diagrams for plant area soils compacted to 96 percent of the standard Proctor (ASTM D-698) maximum dry density and tested under triaxial conditions.
Figure 2-123 shows the consolidation characterisitics of several plant area soil samples compacted to 96 percent of the standard Proctor maximum dry density.
Permeabilities of compacted plant area soil samples are very low and are presented in Table 2-
- 89. Except for the buried Diesel Generator Fuel Oil Tanks and portions of the NSW pipe lines and SNSW Pond outlet works, no Category I structures are supported on structural fill (see Section 2.5.4.10).
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 37 Soils used in the SNSW Pond Dam embankment are excavated from borrow areas located on Figure 2-78. Results of tests on soils used in the SNSW Pond Embankment are presented in Section 2.5.6.4.
2.5.4.2.4.3 Partially Weathered Rock Partially weathered rock is a term arbitrarily applied to residual soilsaprolites having standard penetration resistances greater than 100 blows per foot. This material comprises a major weathering zone at the site.
The partially weathered rock material is similar in grain size distribution to the silty fine to medium sands (SM). The major difference is that the partially weathered rock is less weathered and harder. The depth to the top of the partially weathered rock stratum varies from several feet to 30+/- feet below the preconstruction surface.
Figure 2-124 summarizes the consolidation characteristics and Figure 2-126 illustrates a summary of Mohr envelopes for the partially weathered rock material. Dynamic moduli are presented in Figure 2-129. The results of falling head laboratory permeability tests are given in Table 2-73. The engineering properties shown in the figures and table noted above are based on testing of partially weathered rock in the plant area. Those materials, as they pertain to the foundation of the SNSW Pond Dam, are discussed further in 2.5.6.2. In general, the partially weathered rock has been excavated from the plant area and does not underlie any major Category I structures (see Section 2.5.4.10). The Refueling Water Storage Tanks and Diesel Generator Fuel Oil Tanks are underlain by partially weathered rock.
2.5.4.2.4.4 Bedrock The primary parent bedrock at the site is adamellite. The transition from partially weathered rock to the unweathered rock is somewhat gradational. The upper zones of the bedrock are variably weathered with many partially weathered rock zones between harder, less weathered rock layers. With increasing depth, the weathering decreases until moderately hard to hard continuous bedrock is encountered. For the purposes of this report, the elevation of the top of continuous rock is defined by a Rock Quality Designation (RQD) of at least 75 percent on moderately hard to hard rock in core, and is tabulated in Table 2-87. The top of continuous rock in the plant area is contoured from boring results on Figure 2-100.
Test results from packer permeability testing of the bedrock in the plant area are given in Table 2-72 and discussed in Section 2.4.13.2.5. There were no field tests conducted to determine the in-situ strength and deformation characteristics of rock.
Tests on rock cores in all states of weathering indicate a saturated surface dry density of 155 to 180 pcf. The results of unit weight, unconfined compression testing and stress-strain measurements of NX rock cores are shown in Figure 2-95.
The results of unit weight and unconfined compression testing of samples obtained from the rock exposed by the foundation excavations are presented on Figure 2-128. Locations from which these samples are obtained are shown on Figure 2-79. The unconfined compression test results on Figure 2-127 and Figure 2-128 are generally within a range of 4000 to 18,000 psi. In areal distribution, locations of measured rock strengths of 4000 psi to 10,000 psi are generally situated in an area of the Reactor - Auxiliary Building excavation that is relatively free of shearing and brecciation. It may be speculated that silica was preferentially mobilized and later fixed in and near shear breccia zones resulting in greater rock strengths (10,000 to 18,000 psi) in those areas.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 38 (09 OCT 2016)
Compression wave (P-wave) velocities measured by refraction profiling on the exposed foundation rock in the Units 1 and 2 excavations are shown on Figure 2-120. The measured P-wave velocities range from about 5400 fps to 18,100 fps. The locations of these seismic lines are shown on Figure 2-117. Figure 2-98 and Figure 2-99 show the uphole seismic compression and shear wave velocities, as well as computed values of shear modulus, from borings A61 and A-63 in the plant area.
In summary, analyses of the laboratory tests on the rock cores and the seismic data yield the following range of typical values for unit weight, unconfined compressive strength and wave velocity representative of the moderately hard to hard continuous rock in the powerhouse area.
UnconfinedCompressive Strength 4,000 psi to 18,000 psi Unit Weight 160 pcf to 170 pcf Dynamic Poisson's Ratio 0.23 to 0.32 Compression Wave Velocity 5,400 fps to 18,100 fps Shear Wave Velocity 3,200 fps to 10,700 fps The moderately hard to hard bedrock is the supporting material for the major Category I structures which include the Reactor and Auxiliary Buildings. The existence of joints and shear joints in the foundation rock mass was accounted for in evaluating the rock's ability to resist punching shear failure as described in Section 2.5.4.10.1.
2.5.4.3 Exploration The exploration of the Catawba site is discussed in detail under Field Investigation in Section 2.5.4.2.1. The purpose of all the explorations is to identify and characterize the subsurface soil, groundwater and geologic conditions, determine the properties of the subsurface materials by in-situ testing methods, and provide representative samples of the subsurface materials for laboratory determination of their properties.
2.5.4.4 Geophysical Surveys A detailed discussion of the geophysical surveys is presented in Section 2.5.4.2.1.5. The geophysical data are used to supplement the test boring data. Appropriate geophysical data are also used in combination with dynamic laboratory test data to evaluate the dynamic load response of foundation materials at the plant site (see Section 2.5.4.7).
2.5.4.5 Excavations and Backfill 2.5.4.5.1 Excavations Locations and limits of excavations, fills and backfills are shown on Figure 2-20. The general excavation for the Category I structures begins using conventional earth moving methods and equipment. With increased depth the degree of weathering decreases and the percentage of hard or dense soils increases. Ripping is required to remove these hard or dense materials. As the excavation extends deeper, ripping becomes less productive and blasting of the partially weathered rock and rock using holes drilled on a grid pattern is required. Figure 2-130 shows the general elevations in the plant area where drilling for blasting is started.
The extent of the Category I excavations is shown on Figure 2-132 through Figure 2-134. The substructure profiles are shown on Figure 2-80 through Figure 2-93.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 39 The SNSW Pond Dam is a Seismic Category I fill and is discussed in Section 2.5.6.
2.5.4.5.2 Dewatering Control of groundwater seepage and surface runoff into the foundation excavations in the plant area during construction are achieved by gravity drainage through ditches which lead to sumps where the accumulated water is removed by pumps. No other construction dewatering measures are required for the plant foundation excavations. Dewatering of the SNSW Pond Dam foundation is discussed in Section 2.5.6.3.
2.5.4.5.3 Foundation Preparation The major Category I structures are supported on continuous rock, except those discussed in Section 2.5.4.10. In areas where the continuous rock level is below the foundation bearing level, fill concrete is required to bring the area up to a suitable elevation. The exposed rock foundation surface is prepared for concrete placement in the following manner: (1) Loose rock and soil are removed from the bearing surface, (2) Unacceptable irregularities such as overhangs are removed, (3) The surface is cleaned with air, water, or brooms and (4) Fill concrete is placed as necessary to obtain the required elevation for placement of the structure foundation.
No special foundation preparation is required for the Category I foundations on partially weathered rock, other than protection from the weather.
Foundations for Group I fill are stripped to remove weak or unsuitable materials, leveled by filling depressions or cavities with Group I Backfill, and then scarified prior to the placement of fill materials.
Quality Control for establishing final foundation grades, the foundation preparation, and the embankment construction is in conformance with construction specifications and drawings. All work necessary for constructing a safetyrelated Category 1 foundation or embankment is controlled by the Duke Power Company Quality Assurance Program. This program is incorporated in the FSAR for the Catawba Station by reference in Section 17.0.1 to the Duke Power Company topical report, "Quality Assurance Program," Duke 1-A which has been accepted for use by the Nuclear Regulatory Commission. Written procedures and requirements followed during construction are covered in the Duke Power Company Construction Department Quality Assurance Program.
Discussion of foundation preparation for the SNSW Pond Dam is presented in Section 2.5.6.3.
2.5.4.5.4 Backfill Nuclear Safety Related Structures are backfilled utilizing three types of materials: Earth Backfill, Granular Backfill and Fill Concrete as shown on Figure 2-132 through Figure 2-134. Field tests performed and controls exercised to assure proper compaction of granular backfill materials are described in Section 2.5.4.5.4.2. A zone of Group I Earth Backfill was placed in the powerhouse yard as shown on Figure 2-66. Additionally, Earth Backfill was placed from yard grade to a depth of three feet. In other areas around the powerhouse either Earth or Granular Backfill was used.
In areas where the three types of backfill methods stated above cannot be utilized, flowable fill (controlled low strength materials) may be used as approved and directed by the responsible engineer on a case by case basis. Flowable fill is placed and tested per Chapter 7 of ACI 229R-94 Report, ASTM PS 28-95, 29-95, 30-95, 31-95 and as directed by the responsible engineer.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 40 (09 OCT 2016) 2.5.4.5.4.1 Group I Earth Backfill The source of earth backfill is the on-site residual soil obtained from general grading cuts, foundation excavations and borrow areas. Compaction characteristics and other engineering properties of this material are given in Section 2.5.4.2.4.2.
The distribution and gradation of Group I Earth Fill materials are controlled to assure that no lumps, pockets, or layers of material differing substantially in texture and gradation from surrounding materials occur. After dumping, the material is spread in a nine-inch horizontal layer over the fill area. Harrowing, if required to break up and blend fill materials or to obtain uniform moisture distribution, is performed after spreading.
Compaction is performed at moisture contents within plus or minus three percent of optimum, based on standard Proctor tests for the particular soil. Moisture checks are performed in accordance with ASTM D 2216 at a minimum frequency of two per day. Soil having a moisture content outside the specified limits is either wetted or dried to adjust moisture content prior to compaction, or removed.
Each layer is uniformly compacted to obtain densities not less than 96 percent of the Standard Proctor maximum dry density (ASTM D698). Field determination of compaction is performed in accordance with ASTM D2937, Shelby Tube Method, for each change in soil type or borrow source, and on a frequency of every 1000 cubic yards placed. Discussion of the fill placement for the SNSW Pond Dam is presented in Section 2.5.6.
Figure 2-136 and Figure 2-137 summarize the results of field moisture tests and field compaction tests, respectively for Group I backfill by means of relative frequency histograms.
(Draft SER Page II)
Figure 2-173 summarizes the results of field moisture tests for the SNSW Pond Dam by means of a relative frequency histogram. The distribution and result of field density tests are given in Figure 2-170 and Figure 2-176 for the SNSW Pond Dam embankment and blanket drain respectively.
2.5.4.5.4.2 Group I Granular Backfill Granular Backfill materials (crushed stone) are spread in 12-inch layers and compacted to a minimum relative density of 80 percent in accordance with ASTM D 2049. ASTM D 2049 has been withdrawn and replaced by ASTM D 4253 and ASTM D 4254. ASTM D 4253 and ASTM D 4254 may be used in lieu of ASTM D 2049. Field measurement of density is performed in accordance with ASTM D 1556, Sand Cone Method, on a frequency of once every 400 tons of material. Materials that fail the density test are recompacted or removed. Figure 2-138 summarizes the results of relative density tests for granular backfill by means of a relative frequency histogram.
2.5.4.5.4.3 Fill Concrete Fill concrete, where used, has a minimum compressive strength of 3000 psi after 28 days.
2.5.4.5.4.4 Flowable Fill Flowable fill, where conventional backfill methods cannot be utilized, may be used in areas for backfill as approved and directed by the responsible engineer on a case by case basis.
Flowable fill is placed and tested per Chapter 7 of ACI 229R-94 Report, ASTM PS 28-95, PS 29-95, PS 30-95, PS 31-95 and as directed by the responsible engineer.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 41 2.5.4.6 Groundwater Conditions Groundwater conditions at the plant site are discussed in Section 2.4.13.2.4. The deep foundation and basement excavations for the plant area structures extend below the groundwater table, and as a result there is groundwater seepage into the excavations. This water seepage is controlled during construction by methods discussed in Section 2.5.4.5.2.
With the exceptions of very low pits in the Reactor and Auxiliary Buildings, hydrostatic uplift on the plant area structures is relieved by a permanent groundwater drainage system consisting of foundation underdrains and continuous wall drains which maintain the groundwater levels at or near the base of foundation mats and basement walls. The permanent groundwater control system is discussed in detail in Section 2.4.13.5. The deep pits are designed to withstand any resultant uplift and hydrostatic loads. Stability of the structures against groundwater induced forces is thus achieved by removing the buoyant uplift and hydrostatic forces or by providing suitable reaction against them.
Effects of plant construction on the groundwater conditions and plans to monitor post-construction groundwater levels are discussed in Sections 2.4.13.2.2, 2.4.13.2.4, and 2.4.13.2.5. Records of field and laboratory permeability tests are given in Section 2.4.13.2.5.
2.5.4.7 Response of Soil and Rock to Dynamic Loading The testing performed to evaluate the dynamic properties and characteristics of the soil and rock at the site is discussed in Section 2.5.4.2. Evaluation of the liquefaction potential at the site is discussed in Section 2.5.4.8. The soil structure interaction analyses is discussed in Section 3.7.2.4. The dynamic response of buried pipe lines and earthworks is described in Section 3.7.3.12 and Section 2.5.6.5.
2.5.4.8 Liquefaction Potential All major nuclear safety related structures are founded on rock or partially weathered rock except for localized portions of the NSW pipe lines and the NSW conduit manholes, the SNSW Pond Outlet Works, the Diesel Fuel Oil Tanks as presented on Table 2-90. The rock and partially weathered rock will not be subject to liquefaction resulting from dynamic loading from the SSE.
As shown on Figure 2-134, a portion of the NSW pipeline near borings A-138, A-152, A-153, and A-154 is supported by compacted backfill which is underlain by residual soils, partially weathered rock and rock. Level ground conditions exist in this location on the pipeline, which is about 270 ft from the nearest slope. At boring A-138, a pocket of firm alluvial soil is present between the compacted backfill and partially weathered rock. Nearby borings A-152, A-153, and A-154 did not encounter the alluvial soil during continuous sampling, confirming the alluvium is of limited horizontal extent. The alluvium is a firm micaceous slightly silty fine to coarse sand having standard penetration resistance (N) equal to 13 blows per foot. The compacted backfill is a silty fine to coarse sand and micaceous fine sandy silt, having typical N-values of 15 to 34 blows per foot, with a few N-values of 10 to 14 bpf. The residual soil, which is present only in borings A-152 and A-154, is a micaceous silty fine to coarse sand, having typical N-values of 20 to 43, with one N-value as low as 9 in a thin zone of residual soil at boring A-152. The behavior of the earthfill, granular, residual and alluvial materials during earthquake loading is discussed in the following sections.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 42 (09 OCT 2016) 2.5.4.8.1 Earth Backfill The remolded fill soils tested for cyclic shear strength from the general borrow area of the structures and plant excavations, and similar soils tested from the SNSW Pond Dam borrow areas do not exhibit liquefaction behavior in the sense that the pore pressure does not increase to equal the mean confining pressure. Since liquefaction does not occur, a failure criterion is assigned to these soils based on 5 percent strain, as discussed in Section 2.5.6.5. Analyses made for the SNSW Pond Dam embankment under shaking by the SSE confirm that the dynamic stresses generated in the soils are less than those required to cause 5 percent strain (see Section 2.5.6.5). Thus, it is concluded that Group I fills of site soils compacted to 96 percent standard Proctor maximum dry density and founded on firm saprolite soils and/or partially weathered rock will not undergo liquefaction or excessive deformations during the SSE.
The basis for using 96 percent standard Proctor maximum dry density for compacting Group I fills of site soils is discussed in Section 2.5.4.2.4.2. Field tests performed and controls exercised to assure proper compaction of fills of the site soils are described in Section 2.5.4.5.4.1.
The low consistency (N=10) soil encountered in one sample of boring A-154 is a slightly clayey micaceous fine to medium sandy silt and is compacted fill, which is underlain by residual soils (micaceous silty fine to coarse sand). Degree of compaction, not standard penetration resistance, is the control criterion for placing the earth backfill. As discussed above, the compacted earth backfill will not undergo liquefaction or excessive deformations during the SSE. This conclusion is confirmed by the results of soil column response analyses presented on Figure 2-127, comparing the cyclic shear strength of the compacted fill to the cyclic shear stresses induced by the SSE. The minimum safety factor within the compacted fill is 1.91.
2.5.4.8.2 Granular Backfill The granular backfill material compacted to a minimum relative density of 80 percent will not be susceptible to liquefaction or excessive deformations during the SSE.
2.5.4.8.3 Residual Soil The low consistency (N=9) soil encountered in boring A-152 is a micaceous silty fine to coarse sand and is residuum. The N=9 condition occurs in a single SPT sample. (This boring is located 50 ft perpendicular from the axis of the pipeline). Because of its location, the low consistency residual soil in boring A-152 is not of concern to the NSW pipeline. The residual soil directly below the pipeline in borings A-138, A-153, and A-154 is partially weathered rock having N-values of 100 or more and silty fine to coarse sandsaprolite having N-values of 20 or more.
2.5.4.8.4 Alluvial Soil In order to evaluate the liquefaction potential of the pocket of alluvial soils encountered in boring A-138 (a single sample within the interval from 33 to 30 ft), a soil column analysis using the program SHAKE is performed to compute the cyclic shear stresses induced by the SSE (Figure 2-131). The shear wave velocity of the compacted fill is assumed the same as from the SNSW Pond Dam (Section 2.5.6). The shear wave velocity of the alluvium is estimated from the standard penetration test value, overburden pressure, and information contained in Anderson, et al., (1978) (Reference 115). The shear wave velocity of the partially weathered rock was obtained from in-situ measurements of shear wave velocity in similar material at the Catawba site. The cyclic shear strength of the compacted fill is obtained from tests on the compacted fill for the SNSW Pond Dam, Figure 2-153. The cyclic shear strength of the alluvial sand is estimated from Figure 24 of Seed (1979) (Reference 116). As can be seen from Figure 2-131, the cyclic shear strength of the compacted fill exceeds the induced cyclic shear stresses by a
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 43 safety factor of 1.91 (minimum). The cyclic shear stresses exceeds the cyclic shear strength in the pocket of alluvial sands; thus, this localized zone of alluvial soil has a potential for pore pressures to become equal to the confining pressure during the SSE (consisting of the synthetic time histories).
Borings with continuous sampling are drilled 20 ft each way along the axis of the pipeline and perpendicular to the pipeline (50 ft downstream in the pre-construction drainage feature) to explore the lateral extent of the alluvial soil encountered in one SPT sample of boring A-138.
These borings do not encounter any of this soil; therefore, the alluvial soil at boring A-138 occurs as a pocket of limited lateral extent (maximum of 40 ft along the pipeline) and is confined on all sides by soil (residuum and compacted earth fill) that does not liquefy or undergo excessive deformation during a seismic event.
If the local alluvial soil does experience pore pressures equal to the initial effective confining pressure in this material during the SSE, the relatively small volume and constrained nature of the alluvial zone would not cause a mass soil movement in the flat ground; rather the effect would be localized settlement. Lee and Albasia (Reference 1) measured the volumetric strain of laboratory specimens on reconsolidation after liquefaction. They reported considerable scatter in their results, but when averaged and plotted as Figure 7 in their paper, the data showed a "fairly well defined trend with volumetric strains decreasing as the relative density increases."
For 50 percent relative density, they showed 1.5 percent maximum average strain; for a 6 ft.
thick liquefied soil mass, these results indicate about 1 to 1-1/4 inches of settlement potential.
The settlement of the pipeline subgrade should be less than the potential compression of the sand pocket due to the 24 ft. of cover of compacted earthfill above the layer and below the pipelines. No estimate of the reduction in settlement is made due to the relatively insignificant amount of settlement potential.
The effect of the consolidation of the sand pocket is assumed analagous to the effect of tunnelling in soil, for which there is empirical evidence that the surface settlement trough takes the form approximated by a normal probability curve (Reference 2). To estimate the shape of the potential settlement trough that would result on the ground surface (and at the pipe subgrade in the absence of the pipe) due to reconsolidation of the sand pocket after liquefaction, a trough shaped like a normal distribution curve centred at boring A-138 is assumed. It is assumed that the 1-1/4 inch settlement potential is also the settlement of the ground surface at boring A-138, and thus forms the maximum ordinate for the normal curve. It is further assumed the normal curve passes through settlement values that are 61 percent of the maximum value at 20 ft. away, and 13.5 percent at 40 ft. away. Then, using the properties of the normal distribution curve, the maximum radius of curvature anywhere along the settled ground surface is calculated as 48,350 inches, occurrring at boring A-138 as a concave upwards shape.
This is a conservative estimate of the radius of curvature, considering that the sand pocket may feather out laterally rather than end abruptly as assumed for these calculations. The radius of curvature lengthens quickly to either side of the bottom of the settlement trough and is about 70,000 inches at 10 ft. away from the center. Because of their shallow depth of burial (and thus low soil confining pressure), the 42-inch diameter NSW pipes may not deform as much as the settled subgrade profile. However, for the sake of calculating the maximum for potential stress in the pipe caused by the deflection, the maximum radius of curvature computed and stated above is assumed to be imposed on the buried pipes, and the induced stresses from this curvature moment are computed. The stress resulting from the above curvature, the internal pressure, and an SSE seismic loading is calculated to be 16.36 ksi. This is less than the ASME allowable of 16.44 ksi for a combination of sustained and occasional loads. The material for the NSW pipe is SA-155, Class 2, Grade C55 with a minimum yield stress of 30 ksi, a minimum ultimate tensile stress of 55 ksi and a pipe wall thickness of.435 inches.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 44 (09 OCT 2016) 2.5.4.9 Earthquake Design Basis The earthquake design basis is discussed in Section 2.5.2.
2.5.4.10 Static Stability Major Category I structures are identified on Figure 2-78 and listed on Table 2-90. Major Category I structures are supported by mat foundations which bear on rock or fill concrete to rock. These rock bearing mats have maximum gross total static bearing pressures ranging from 10 to 20 ksf and average bearing pressures in the range of 3 to 10 ksf.
Category I structures founded on partially weathered rock (or coreable weathered rock) include the Refueling Water Storage Tanks, the NSW and SNSW Intake Structures and the SNSW Discharge Structures. These structures are also supported by mat foundations. As shown in Table 2-90, bearing pressures are low.
The NSW pipelines and the NSW Conduit Manholes bear on residual material, except at several isolated locations where the pipes are placed on Group I Earth Fill. The buried Diesel Fuel Oil Tanks and the NSW Pond Outlet Works bear on Group I Earth Fill. These buried structures apply very little net loading to the foundation soils.
Figure 2-134 shows subsurface profiles through the NSW pipelines at one location where the pipes bear on structural fill. A localized zone of alluvium underlies the structural fill at this location. Due to the limited extent of this zone, as determined by subsurface investigation, it will have no effect on the stability of the NSW pipe. At all other locations where the NSW pipes bear on structural fill, all alluvium and other unsuitable materials were removed prior to placing fill.
The sequence of foundation preparation for the NSW pipe backfill was as follows. The trench was excavated using common earthmoving equipment to the depth required for pipe installation.
After the pipe was installed and prior to placing backfill, the foundation was inspected by the field engineer, who verified its adequacy. If, in the judgement of the field engineer, an area contained loose or otherwise unsuitable material, that material was removed to a depth which ensured an adequate foundation. The plan and longitudinal profiles of the NSW pipes are shown with subsurface information on Figure 2-132 (Sheet 1-4).
The NSW pipes are designed for external pressure due to Cooper E-80 railroad loading in locations where the pipes pass under railroads and for AASHTO H50 truck loading along the remaining portion of the pipe. Additionally, the pipes are qualified for the loads from the Dry Cask Transporter at locations where the pipes pass under or near the Dry Cask Haul Road. A portion of the 42" diameter NSW supply and return lines are protected from the heavy loads from the Dry Cask Transporter by a grade level pile supported concrete bridge structure.
The relationship of the plant area foundation mats and basement walls to the preconstruction ground surface and weathering profile are shown on selected cross sections through the plant in Figure 2-132. Subsurface profiles at the Diesel Fuel Oil Tanks are shown in Figure 2-133.
Field and laboratory test results for the various foundation bearing materials are presented in Section 2.5.4.2.4. The static stability of Category I structure foundations is discussed in detail in the following paragraphs and summarized in Table 2-90. Stability of the SNSW Pond Dam is discussed in Section 2.5.6.5.
2.5.4.10.1 Category I Mat Foundations on Rock Category I structures with mat foundations bearing on rock are indicated on Table 2-90.
Generally, these mat foundations bear below the zone of major weathering on moderately hard to hard continuous rock having a rock quality designation (RQD) of 75 percent or greater as
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 45 determined from core borings. One exception is the foundation for the Unit 2 Diesel Generator Building which bears on rock above the top of continuous rock (TOCR) as defined above from core boring results. The foundation rock at this location, as shown by boring A-8, is moderately hard to hard and requires rock coring to penetrate, but contains some weathered joints and consequently has an RQD less than 75 percent.
The safe bearing capacity of mat foundations on the rock is many times the static bearing pressures imposed by the structures. Intact specimens of the rock are, on the average, stronger than the concrete in the mat foundations. Thus, the ultimate bearing capacity of mat foundations on the rock is such that very high factors of safety exist with respect to bearing capacity instability. This is shown by the following very conservative evaluation of the rock's ability to resist punching shear failure.
The results of the seismic wave velocity measurements made on the exposed foundation rock (Figure 2-120) show the joints are essentially tight and closed. However, for purposes of demonstrating the conservativeness of the static foundation design, the rock mass is assumed to be comprised of rock columns, created by relatively closely spaced vertical and near vertical and slightly open jointing surfaces. The ultimate bearing capacity of a foundation is then the sum of the compressive strengths of the rock columns under the bearing area, and is thus taken to be equivalent, on a unit load basis, to the average unconfined compressive strength of the rock. For this evaluation, 965 ksf (6700 psi) is considered to be the minimum ultimate unit pressure bearing capacity of the moderately hard to hard rock on an area the size of the loaded areas. This value is obtained from the average of these tested samples' unconfined compressive strength values less than 10,000 psi tabulated in Figure 2-128 excluding one low value (915 psi) that represents a very localized condition. As shown on Table 2-90, the minimum safety factors derived by this conservative evaluation of the rock bearing capacity range from 48 for the Reactor Buildings to 182 for the NSW Pump Structure.
Foundation settlements on rock are produced by elastic deflection of the rock, compression of any open horizontal joints and consolidation of any random zones of soil or soft rock. From close examination of rock cores, geologic mapping of the foundation excavations, and seismic refraction surveys, no significant open joints or soft zones are observed in the rock beneath the Category I mat foundations. Thus, settlement of these mat foundations is limited to elastic compression which occurs instantaneously during construction as the load is imposed. The magnitude of this elastic deflection is negligible for the range of bearing pressures exerted by the structures, as shown by example in the following evaluation of settlement of the Unit 2 Diesel Generator Building.
The Unit 2 Diesel Generator Building foundation is selected for quantitative settlement evaluation since a finite thickness of foundation rock at that location has an RQD less than 75 percent. The settlement is calculated using the following equation from the theory of elasticity (Reference 78).
E A
B
)
1(
P 2
/
z 2
1
Where:
= settlement at foundation level, P
= total load on foundation, A
= area of foundation,
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 46 (09 OCT 2016)
V
= Poisson's ratio E
= static Young's modulus for in-situ rock; Bz
= shape and rigidity factor This equation assumes that the foundation is placed on a linear elastic isotropic half space. It is conservatively assumed that the lower RQD rock beneath the Unit 2 Diesel Generator Building extends to infinite depth, although it actually has a finite thickness of 26 ft or less immediately below the foundation. Using RQD results from boring A-8, the weighted average RQD of the foundation rock is calculated to be about 37 percent. This RQD value is used to relate Young's modulus values determined by uniaxial unconfined compression tests on intact test specimens to the modulus of the rock mass, by way of a published correlation between RQD and modulus reduction factor (Reference 79). The static modulus values for the test specimens are tabulated in Figure 2-127. A representative modulus value of 12x106psi for the test samples is obtained by taking the average of the tabulated values that are less than 20x106psi. A modulus reduction factor of 0.1 is given for an RQD of 37 percent, and a Young's modulus of 1.2x 106psi is thus obtained for use in the above equation. The geometry and loading information given in Table 2-90, an appropriate shape and rigidity factor of 1.07, and a static Poisson's ratio of 0.23 are used. An elastic deflection less than 1/10 inch is calculated for the average bearing pressure. It is thus seen by this conservative evaluation that elastic deformation of the rock beneath the Unit 2 Diesel Generator Building is very small and insignificant. Settlement measurements made at four locations on the roof of the Unit 2 Diesel Generator Building show that settlement after one year has essentially stabilized at 0.02 ft. or 0.24 in.
2.5.4.10.2 Category I Mat Foundations On Partially Weathered Rock Bearing capacity of mat foundations on partially weathered rock is analyzed using the bearing capacity equation which was first derived by Terzaghi (Reference 80) and later improved by Meyerhof (Reference 81). The equation for ultimate bearing capacity of an infinitely long foundation is as follows:
c o
CN dNq 2
BN q
qo
=
ultimate bearing capacity, psf C
=
cohesion in psf, = soil unit weight, pcf (buoyancy included where appropriate) d
=
depth of footing, feet below finished grade B
=
footing width in feet Nc, Nq, N
=
Bearing capacity factors dependent upon the angle of internal friction. These values are based on Meyerhof. For foundations having a limited length, corrections to Nc and N are multiplied times the bearing capacity factors in the above general equation for infinitely long footing foundations. For foundations that are eccentrically loaded or subject to inclined loads, Meyerhof (Reference 82) has proposed the concepts of reduced width and corrections to the bearing capacity factors, respectively.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 47 An estimate of the minimum ultimate bearing capacity of the partially weathered rock is made using the above equation with appropriate correction factors and the following criteria;
- 1. Foundation geometry and embedment as indicated on Table 2-90
- 2. Shear strength parameters of partially weathered rock 2a) Cohesion (C) = 1 ksf 2b) Angle of internal friction () = 44°
- 3. Unit weight () = 140 pcf Submerged unit weight (1) = 78 pcf
- 4. Groundwater level at ground surface for purposes of this calculation only The calculated ultimate bearing capacities of the partially weathered rock supported mats are shown on Table 2-90. As indicated on the table, the ultimate bearing capacity of the partially weathered rock far exceeds the imposed static bearing pressures by factors of safety equal to or greater than 30.
Settlement of mat foundations bearing on the partially weathered rock is evaluated using the appropriate analytical procedures from the theory of elasticity and the physical properties of the partially weathered rock materials determined by laboratory tests.
The Westergaard analysis (Reference 83) for stress created beneath a loaded area is used for calculating stress increases in the partially weathered rock materials underlying the mat foundations. The settlement analysis consists of dividing the foundation materials into layers and computing the initial gravity stress and the stress increase (from the Westergaard analysis) in each layer. Foundation settlements due to compression of the various layers are then computed from the stress-void ratio (or stress-change in sample height) curve resulting from the laboratory consolidation tests. The initial void ratio (eo) corresponding to the initial overburden stress (with due consideration for the effects of any stress removal by excavation) and the change in void ratio (e) caused by the average stress increase for a given soil layer are found directly from the stress-void ratio curve. The settlement for the soil layer of thickness H is then computed from the following relation:
oe 1
e H
)
H
(
Settlement
The total compression settlement of the foundation is then calculated by summing the settlements for each layer beneath the foundation.
The laboratory consolidation data for the deeper partially weathered rock samples (see Figure 2-124) are used in this settlement evaluation. Maximum calculated settlements for the partially weathered rock supported mats are negligible as shown on Table 2-90. Differential settlements of the partially weathered rock supported structures are not of concern because of the small magnitude of total settlement and the use of mat foundations. Maximum differential settlements within any one of these structures could equal the calculated total settlement. However, the mat foundations minimize curvature induced in the structures by any differential settlements.
Settlement measurements are taken at four locations on the foundation of each Refueling Water Storage Tank and have essentially stabilized at 0.023 ft. or 0.276 in. for the Unit 1 Tank and at 0.018 ft. or 0.216 in. for the Unit 2 Tank.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 48 (09 OCT 2016) 2.5.4.10.3 Category I Structures on Soil The Category I structures bearing on soil include the buried Diesel Fuel Oil Tanks, the buried SNSW Pond Outlet Works, the buried NSW pipelines and the NSW Conduit Manholes. The tanks and pipes are bedded directly on the foundation soils.
The Diesel Fuel Oil Tanks bear on a relatively thin zone of Group I Earth Fill which overlies partially weathered rock. The NSW pipes and conduit manholes generally bear within residual materials. In several low areas (e.g., drainage swales) that were filled in to achieve yard grade elevation, the pipes and manholes are embedded in Group I fill. In general, the yard fills are constructed prior to making excavations for the pipelines, tanks and manholes.
The loading of the pipes, tanks, and manholes are less than the pre-excavation overburden pressure. Thus, settlement is small and due essentially to recompression, after rebound, under the weight of backfill and structure. Using typical results of consolidation tests made on compacted samples, shown on Figure 2-171, a conservative estimate of the recompression settlement under the Diesel Fuel Oil Tank is made and is given in Table 2-90. These Category I Structures are not monitored for settlement.
Because of the negative net loading of these buried facilities, bearing capacity is not a concern.
For illustration, however, ultimate bearing capacities of the foundation soils under the tanks and pipes are estimated and given in Table 2-90. These estimates are based on design parameters given in Table 2-94 for the various foundation materials.
2.5.4.10.4 Subsurface Static Lateral Loading The rigid concrete basement and substructure walls of the Category I facilities are designed to withstand lateral loads resulting from backfill, construction equipment surcharge and differential hydrostatic pressure. The hydrostatic pressures against the Auxiliary and Reactor Buildings are removed by a permanent Category I Groundwater Drainage System as discussed in Section 2.4.13.5. The walls are designed for at-rest earth pressures. An at-rest earth pressure coefficient of 0.5 and total soil unit weights are used to develop the static design earth pressure loading. The at-rest earth pressure coefficient of 0.5 for compacted (earth) backfill and static earth pressure loading is compatible with the material type - silty sand (SM) having very low plasticity and generally 10 to 30 percent passing the number 200 Sieve. The Ko calculated from the expression for normally consolidated soils (1-sin 1) is 0.44. This would be applicable at depth within the backfill where the weight of the overlying material dominates over the preconsolidating effect of the compaction equipment. The work of Wroth (1975), Reference 110, suggests Ko for such soil would be equal to 0.44 to 0.50 for overconsolidation ratios of 1 up to 5 and for a Poisson's ratio of 0.3. The adopted value of 0.5 is therefore considered adequately conservative for the walls. Detailed recommendations for the distribution and magnitude of the earth pressures are shown on Figure 2-135.
The lateral earth pressure effects during compaction of the backfill was accounted for in design by applying equivalent surcharge or line loads to the backfill surface to simulate the static and impact loads from the compaction equipment. Dynamic lateral soil pressures under seismic loading are discussed in Section 3.7.2.4.
2.5.4.11 Design Criteria Design Criteria are discussed in Sections 3.1 and 3.8.
2.5.4.12 Techniques to Improve Subsurface Conditions Foundation improvements for the SNSW Pond Dam are described in Section 2.5.6.3.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 49 HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.5.4.13 Subsurface Instrumentation Subsurface instrumentation for the surveillance of Nuclear Safety Related structure foundations consists of a groundwater monitoring program as described in Section 2.4.13.3. No other post-construction instrumentation for foundation surveillance is utilized except for the instrumentation defined in Section 2.5.6.8.1 for the SNSW Pond Dam.
The latest results of the groundwater monitoring program are given in Section 2.4.13.5.
2.5.4.14 Construction Notes There were no significant construction problems associated with Nuclear Safety Related structure foundations through the period of the end of construction of the station.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.5.5 Stability of Slopes The Standby Nuclear Service Water Pond Dam Stability Analysis is presented in Section 2.5.6.
There are no other natural or man-made slopes the failure of which would prevent safe shutdown of the plant or which pose a hazard to the plant. Failure of the yard slope above the LPSW intake structure will not affect any safety related structures or components.
2.5.6 Embankments and Dams 2.5.6.1 General 2.5.6.1.1 Standby Nuclear Service Water Pond Dam 2.5.6.1.1.1 Purpose The Standby Nuclear Service Water Pond Dam, as stated in Section 2.5.5, is the only Nuclear Safety Related embankment at the site. The purpose of the dam is to contain an adequate supply of water to dissipate waste heat rejected during a unit LOCA and/or a unit cooldown.
Normally, this supply of water is taken from Lake Wylie; but with a postulated loss of Lake Wylie, due to a catastrophic seismic event, the SNSW Pond Dam impounds water within a cove of the lake and provides the necessary volume of cooling water. Section 9.2.2 describes in detail the SNSW System.
Normally, the SNSW Pond is operated between full pond elevation 574 msl and maximum drawdown 571 msl. Lake Wylie is operated between full pond elevation 569.4 msl and maximum drawdown 559.4 msl. Thus, under normal operating conditions, the SNSW Pond Dam is subjected to differential hydrostatic heads of 1.6 to 14.6 feet. Should conditions occur such that Lake Wylie Dam fails, the Lake Wylie pool level drops to elevation 550 msl, and the dam is subjected to a 24 foot differential hydrostatic head.
Refer to Section 2.4.1 for a hydrologic description of the site.
2.5.6.1.1.2 Site Selection The location of the SNSW Pond Dam, 2800 feet north of the plant site, is chosen to take advantage of the natural topography of the area. The site was originally dry land with a stream
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 50 (09 OCT 2016) channel along its bottom. In 1925, when Lake Wylie was raised, the area was flooded and the now existing cove was formed.
The general north-south oriented dam axis crosses the cove, which is oriented in a general east-west direction. The south abutment of the dam, before clearing and earthwork construction began, was a wooded hillside. The lightly-wooded north abutment of the dam had been partially developed as a residential and recreational area. A profile of the original ground surface is shown on Figure 2-145.
2.5.6.1.1.3 General Design Features The SNSW Pond Dam is an impervious, rolled, homogeneous earthfill structure. The dam is founded on stripped virgin residual soil or partially weathered bedrock, and extends 1710 feet between abutments at crest elevation 595.
The dam has a crest width of 35 feet, and upstream and downstream faces sloped at three to one (horizontal to vertical) (See Figure 2-140). All surfaces are protected against wave action with stone riprap (See Figure 2-141).
The dam centerline runs from the south abutment North 22° East approximately 656 feet, then follows a circular curve for approximately 272 feet, and then runs North 39° West approximately 782 feet to the north abutment. A plan of the dam is presented in Figure 2-142.
A zoned blanket drain and/or toe drain extends the full length of the east face (Lake Wylie side) of the dam to control rapid downstream pore pressures on this slope, should the lake level be rapidly lowered. A cross section of the blanket drain is shown on Figure 2-141. For a detailed description of the blanket drain, refer to Section 2.5.6.4.7.
The outlet works for the dam consists of a 60-inch diameter steel pipe south of the embankment, with a headwall on the SNSW Pond side and an endwall on the Lake Wylie side.
The pipe is sloped at two percent down from an invert elevation of 571.0 at the headwall to an invert elevation of 559.0 at the endwall. The pipe is protected against missile impact by a reinforced concrete slab above the pipe, where less than five feet of earth cover is provided.
Concrete cutoff walls are provided at 100 feet intervals to prevent seepage along the pipe-soil interface. Removable trash racks are provided at the headwall. The outlet works are shown on Figure 2-143.
The dam is statically designed for stability under the following conditions: (1) end of construction condition; (2) steady state seepage through the dam, with the worst combination of water levels; (3) sudden drawdown of Lake Wylie water level from maximum flood level. The analyses show the dam to have ample safety factors under all the loading conditions. Refer to Section 2.5.6.5 for a detailed description of static stability analyses and results.
Under earthquake loading conditions, the dam is analyzed for stability during the SSE using the finite element technique for the condition of steady state seepage through the dam with water levels at full pond on both sides. The analysis shows the dam to have an adequate margin of safety during loading by the Safe Shutdown Earthquake. Refer to Section 2.5.6.5 for a detailed description of analyses performed and the results obtained.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE UPDATED
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 51 2.5.6.2 Exploration 2.5.6.2.1 SNSW Dam Foundation Materials As described in Section 2.5.6.1, the SNSW Dam is located 2800 feet north of the plant in an existing cove of Lake Wylie. The site is selected to take advantage of the natural topography of the area.
The subsurface conditions at the SNSW Dam are explored using soil test borings at locations shown on Figure 2-144. The borings are drilled using procedures described in Section 2.5.4.2.
The subsurface materials encountered at the dam site include alluvial soils, residual soils, partially weathered rock and rock.
The surface soils on the abutments and under the alluvium in the lake are residual materials formed in-place by weathering of the parent bedrock. The residuum increases in consistency with depth and grades into partially weathered rock. This partially weathered rock material is exposed on the eroded slope between elevation 570 and 575 on the south abutment.
Underlying the partially weathered rock material is the adamellite bedrock.
Figure 2-145 through Figure 2-149 illustrate the subsurface weathering conditions encountered by the borings. Descriptions of the testing procedures used in the foundation materials is presented in Section 2.5.4.2.
2.5.6.2.1.1 Alluvium The alluvium consists of very soft to stiff sandy silts (ML) and very loose to firm silty fine sands, (SM) ranging in thickness from 0 to 14 feet. The major thickness of the alluvial deposits is in the deepest part of the lake in the vicinity of boring A-102. Near the lake shore the alluvial deposit pinches out. This deposit is formed by erosion products from higher ground and is variable in material type and shear strength. Standard penetration resistances vary from 0 where the spoon advances under the weight of the drilling tools to 14 blows per foot.
Due to its variability, and generally poor engineering characteristics, the alluvium is unacceptable as foundation material and is completely removed from under the dam base.
2.5.6.2.1.2 Residual Foundation Materials The residual foundation soils-saprolites are weathered in-place from the underlying adamellite bedrock. The general degree of weathering is most advanced at the exposed surface, and decreases with depth.
Classification In the area of the north abutment at borings A-104, A-105, A-157, A-158 and A-160, there is a near-surface stratum of fine grained sandy silts having a Unified Soil Classification of ML. These soils are removed from beneath the dam in the foundation preparation Below these silts at the north abutment and below the alluvium at the lake bottom and on the south abutment is a zone of silty sand saprolite with an SM Unified Soil Classification. An envelope of grain size distribution curves for the majority of the residual soils in the SNSW Pond Dam foundation is plotted on Figure 2-150. This presentation of the grain size distribution range represents only the textural classification of the residual soils. The soils, though classified texturally as silty sand, do not behave as would a silty sand of sedimentary origin. Instead, these soils, which are formed by the incomplete weathering of the parent bedrock, still reflect many of the characteristics of the original rock. Some features of the micro structure of the rock still persist in the soil - the grains are still interlocking.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 52 (09 OCT 2016)
Permeability Field and laboratory permeability tests are made at locations shown on Figure 2-145, Figure 2-147, Figure 2-149. No drill water loss occurs in any of the borings at the dam site. The field test locations are selected to yield data from undisturbed residual materials occupying various zones in the weathering profile where the head of water on the foundation is the highest. The methods used to perform the in-place soil and partially weathered rock permeability tests are described in Section 2.5.4.2.1.4. The field permeability test results for soil and partially weathered rock materials are presented in Table 2-93. The representative values of permeability are presented in Section 2.5.6.6.
Shear Strength Mohr envelopes of triaxial shear strengths for the residual foundation soils are plotted on Figure 2-151. The design values are shown on this figure and are tabulated in Table 2-94. Cyclic triaxial shear tests are performed on undisturbed foundation samples to evaluate the dynamic strength of these soils as shown on Table 2-98 and Figure 2-152.
Consolidation Figure 2-154 summarizes the consolidation characteristics for the residual soils. Consolidation of the residual foundation soils under the embankment loading is rapid with a high percentage of consolidation occurring very shortly after application of the embankment load (Reference 84).
The rapid consolidation of the embankment and foundation soils above the water table occurs because the soils are not fully saturated. Under loading the unsaturated soil voids are rapidly compressed resulting in quick consolidation. Below the water table the hydrodynamic consolidation occurs rapidly because of the small amounts of compression, and the rapid pore pressure dissipation afforded by the relict cracks in the in-situ saprolite.
Foundation settlements are calculated assuming the full embankment weight and the reservoir empty. A profile near boring A-103 is selected for this calculation. This profile includes 55 ft of embankment material overlying 20 ft of residual saprolites (N<100) and 10 ft of partially weathered rock. The calculation indicates total foundation settlements (excluding the embankment) of only 5-1/2 inches. Experience with construction on similar soils, plus calculations for time-rate of settlement for the above soil profile using the consolidation test data, indicate that most of the settlement occurs rapidly as the load is applied during embankment construction. The amount of foundation compression remaining after completion of embankment construction is calculated to be about 2/3 inch for the above described soil profile.
Variations in foundation conditions under the dam are not abrupt, but occur over relatively large horizontal distances, and the abutments are relatively flat-sloping. Furthermore, the foundation conditions are good, and settlements will be relatively small. Some differential settlement occurs, over considerable horizontal distances, and is not excessive for the embankment.
Therefore, cracking of the embankment is of no concern in the design.
In-Situ Wave Velocity Measurements Two test locations are selected on the axis of the SNSW Pond Dam. The primary location, near boring A-103, is in Lake Wylie at a water depth of approximately 7 ft. with the water surface at elevation 566. The second location, near boring A-105, is on shore but close to the shore line.
The zones of primary interest for shear wave measurements are the residual soil and partially weathered rock which overlie the unweathered rock. A description of the test procedures is presented in Section 2.5.4.2.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 53 Figure 2-155 and Figure 2-156 show the cross hole wave velocity data for the residual foundation materials. These values are all field-measured. No laboratory measurements of wave velocity are considered necessary. However, resonant column tests are done on undisturbed samples to evaluate the modulus damping-strain relationships. The results are shown on Figure 2-157.
The shear and compression wave velocities in Figure 2-155 show a reduction at a depth between 20 and 30 ft, then a rapid increase with depth. This wave velocity increase with depth reflects the decrease in the degree of weathering with depth, which is typical of the weathering profile in the Piedmont. The soil conditions disclosed by the boring data and static laboratory tests do not reveal a reason for the reduction in shear and compression wave velocity in Figure 2-155. The prepared foundation line at the location of boring A-103 actually is below the zone of reduced shear wave velocity (see Figure 2-145) thus removing the material in this zone.
2.5.6.2.1.3 Partially Weathered Rock The partially weathered rock materials have standard penetration resistances in excess of 100 blows per foot.
Classification When disturbed by split-spoon sampling, the individual grains of the partially weathered rock often disintegrate into a silty sand. The grain size distribution of this material lies within the envelope plotted on Figure 2-150, the residual foundation soil - saprolites.
Permeability A discussion of permeability is presented in Section 2.5.6.6.
Shear Strength The Mohr envelopes of triaxial shear strength are plotted on Figure 2-158. The 3-in. diameter undisturbed partially weathered rock samples for this testing are secured by using a Pitcher Sampler and drilling mud. Design values of shear strength are shown on the figure and are tabulated in Table 2-94. The partially weathered rock strength characteristics are such that it will not be a significant consideration in the dam analysis.
Consolidation Consolidation of the partially weathered rock materials will be rapid and minimal, and will not be a significant factor in the overall embankment settlement. Foundation settlements are calculated for the partially weathered rock at A-101 and A-102. These calculations, based on the full embankment weight with the reservoir empty, indicate a total foundation settlement of only 3/4 of an inch.
2.5.6.2.1.4 Bedrock Classification The bedrock underlying the SNSW Pond Dam is adamellite, identical to that underlying the entire plant to the south. During the drilling program no drill water losses are encountered in any of the soil or rock borings.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 54 (09 OCT 2016)
Permeability Field Packer Tests in the adamellite bedrock are made at boring A-101P8, A-103, A-103S2, and A-117. The test locations are selected after the foundation profile is defined to provide data along the dam axis and beneath the slope. The test methods are described in Section 2.5.4.2.1.4. Test results are preseneted in Table 2-95.
Shear Strength Rock shear strengths are based on laboratory unconfined compressive tests on NX rock core from the plant area. The shear strength and consolidation characteristics of the rock are such that they may be excluded in the dam analysis. The geologic features in the foundation rock of the SNSW pond dam are described in Section 2.5.1.2.2.5 and indicate steeply dipping to vertical foliation, jointing and shear zones which thus form no planes of weakness of concern to the evaluation of stability of the dam.
Dike Rocks The weathered and unweathered adamellite also contain steeply dipping dikes, generally of the composition of amphibolite, and occasionally a diorite. The dikes generally trend north to north-northeast, and thus parallel the dam axis in the deepest part of the structure. There are some dikes trending northwest, across the dam in an upstream-downstream direction.
The dikes have two characteristics of interest from the engineering viewpoint with respect to the dam foundation. First, they weather to a generally fine-grained soil (sandy silt), and are thus low permeability zones. Secondly, they are generally slightly softer than the surrounding partially weathered rock from adamellite. The first characteristic is potentially beneficial from the viewpoint of seepage through the foundation, though seepage is of no concern because of the low head differential to which the dam will be subjected. The second characteristic is potentially detrimental if the dikes represent zones of potential weakness along with foundation failure surfaces might propogate. Because of the persistently steep dip angles of the dikes (generally more than 65 degrees from horizontal), however, it is concluded that the dikes do not represent potential planes of weakness for foundation failures, and thus are of no concern for the integrity of the SNSW Pond Dam.
Shear-breccia zones that are present in the dam foundation are discussed in the Final Geologic Report on Brecciated Zones (Reference 107).
2.5.6.2.2 Embankment Material The areas investigated and designated for use as SNSW Pond Dam borrow areas are as follows: Cooling Tower Yard, North Abutment of SNSW Pond Dam and Borrow Area N-1. The locations and limits of the borrow areas are shown on Figure 2-78. Test pits, as described in Section 2.5.4.2, are used to explore these borrow areas. Samples are obtained from the test pits and tested in the laboratory using procedures described in Section 2.5.4.2. The physical proper ties of the embankment materials actually used in the dam are discussed in Section 2.5.6.4.
2.5.6.2.3 Geologic Features of the Foundation Materials The geology of the SNSW Pond Dam Foundation is discussed in Section 2.5.1.2.2. Engineering significance of the geologic features and the weathering profile are discussed in Section 2.5.6.2.1.4.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 55 2.5.6.3 Foundation and Abutments 2.5.6.3.1 General Treatment Cofferdams are constructed in Lake Wylie to dewater the dam construction area. Groundwater is controlled during construction by shaping and sloping the excavation bottom, and by ditches and sumps with pumps where necessary, so as not to degrade the foundation material. A discussion of groundwater control is given in Section 2.5.6.4.8.
The prepared foundation surface is examined for springs. The criterion for defining springs is a groundwater flow from observed, localized relict joints within the foundation material. The locations of the springs encountered in the foundation area are shown on Figure 2-159. The groundwater flow is controlled to allow embankment material to be placed and compacted in the dry by use of a granular drain in shallow trenches and vertical drainage pipe for pumping.
Following clearing, grubbing, and stripping topsoil, all the alluvial soils are removed from beneath the dam. Also, all other soils having shear strengths less than the design shear strength are removed. The limits of removal of unsuitable foundation soil under the centerline of the dam are shown on Figure 2-145 and Figure 2-147 through Figure 2-149. Dynamic penetrometers calibrated for the site and Standard Penetration Tests (ASTM D 1586) are used to verify the shear strength of materials that are left in-place. Foundation material having a Standard Penetration Test resistance (or equivalent dynamic penetrometer resistance) of less than 15 blows per foot is rejected and removed. Distribution of foundation field tests is shown on Figure 2-159.
Visual inspection and operation of the construction equipment shows the south abutment from Station 0+00 to Station 2+00 to be an area of the foundation consisting of moderately weathered to unweathered adamellite bedrock (See Figure 2-160). Therefore, foundation verification by penetration or penetrometer tests is not required in this area.
Irregularities in the partially weathered rock and rock are exposed, cleaned and filled with concrete prior to placement of the main embankment material. Dental concrete is placed on the south abutment as shown on Figure 2-159. Depth of dental concrete varies from three to twelve inches.
To further reduce seepage, the prepared foundation on partially weathered rock and rock is slush grouted prior to placement of the first embankment materials to fill any minor surficial irregularities and provide a bond between the prepared foundation and embankment materials.
The area slush grouted is shown on Figure 2-159. Depth of slush grout averages approximately two to three inches.
The effectiveness of dental concrete and slush grouting is evident by the seepage test results presented in Section 2.5.6.10.
The south abutment is somewhat steeper than other areas of the SNSW Pond Dam foundation.
Abrupt changes in this slope, such as horizontal benches, haul roads, and any pronounced surface irregularities are removed to reduce the potential for any cracking and leakage through this abutment.
As discussed in Section 2.5.6.1.1.1, the dam is subjected to relatively small differential heads during either normal operations or postulated loss of Lake Wylie. Thus, a grout curtain in the foundation materials is unnecessary.
2.5.6.3.2 Construction Specification Pertinent specification requirements relating to the dam foundation preparation are noted below:
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 56 (09 OCT 2016) 41.2.1 Foundation Preparation Groundwater will be controlled during construction by shaping and sloping the excavation bottom, and by ditches and sumps with pumps where necessary. In addition to the previously described stripping criteria, any materials in the dam foundation that deflect excessively or rut under the excavating equipment during the stripping operations will be removed from the foundation area. Also, the flat abutment slopes will be maintained to avoid any abrupt changes in grade.
The precise limits of unsuitable foundation material requiring removal will be verified by inspection and field testing by a soils engineer during foundation preparation. In addition to visual inspection of material behavior under the construction equipment, dynamic penetrometers calibrated for the site will be used to verify the in-place shear strength of materials that will be left in-place. All such work shall be done in the dry behind cofferdams.
Irregularities in soft soil or decomposed rock are to be exposed by dental excavation. Holes, irregularities, test pits, etc., are to be filled with Group 1 fill or lean concrete prior to placement of main embankment.
Immediately prior to placement of first lift of embankment material, any rock surfaces or any surfaces corrected during dental treatment shall be slush grouted. This shall consist of placement of a relatively fluid mixture of sand, cement and water in surfaces to be treated.
This material shall be spread by brooms or other mechanical means to provide complete coverage as well as optimum penetration into joints, crevices and exposed minor corners.
For horizontal surfaces, where heavy equipment is used for placement and compaction of fill material, a total of three feet of Group 1 fill material placed according to Subsection 26.2.6 shall be placed on the slush grouted surface and compacted. No additional fill or passage of equipment will be allowed on this fill until the grout has reached its final set.
For placement of fill against inclined slush grouted surfaces and/or areas where hand operated power tampers are used for compaction, the depth of Group 1 fill and the time to compact it shall be as directed by the Engineer.
2.5.6.4 Embankment 2.5.6.4.1 Embankment Features The Standby Nuclear Service Water Pond Dam, at its maximum section, is approximately 75 feet in height above prepared foundation, which was approximately 25 feet below the bottom of Lake Wylie. At this point, the distance between upstream and downstream limits of the embankment is approximately 470 feet. Each face of the dam is sloped at three to one (horizontal to vertical) from the 35-foot wide crest down to the toe. Embankment details are shown on Figure 2-140 through Figure 2-142.
The dam contains 536,000 cubic yards of compacted earthfill. Earth embankment materials for the dam are obtained from one minor and two major borrow areas. Locations of the borrow areas are shown on Figure 2-78. A minor quantity of material, 13,000 cubic yards, is obtained from the north abutment area of the dam. A major borrow area located north of the Standby Nuclear Service Water Pond provides approximately 351,000 cubic yards of fill material.
Material obtained from the second major borrow area, the Cooling Tower Yard area, provides 172,000 cubic yards. The distribution within the dam of the materials from each borrow source is shown on Figure 2-161.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 57 2.5.6.4.2 Embankment Material Properties The embankment soils are obtained from the borrow areas located in Figure 2-78.
The physical and engineering properties of the PSAR investigated borrow area soils, compared with the physical and engineering properties of the actually used designated borrow soils, show that the design assumptions used in the PSAR investigation are appropriate and valid for the borrow soils used.
Classification The soil profile to be excavated begins with a few inches to 1 ft. of organic topsoil, then a thin (0 to 3 ft.) near-surface horizon of silt or sandy silt having a Unified Soil Classification of ML (occasionally MH) and with maximum dry densities generally less than 90 pcf. These materials are not used in the SNSW Pond Dam embankment. Instead, the deeper soils in the borrow areas are used.
The deeper soils include a minor soil grouping of material that has Unified Soil Classification of ML, and maximum dry densities in the range of 95 to 105 pcf. These materials are predominantly the soils weathered from the dikes. These dike soils are an insignificant portion of the embankment material. Because of their limited occurrence, they become mixed with and dispersed within the surrounding sandy saprolites during the excavating, hauling and spreading process. The main volume of the embankment material used from all the borrow areas is a saprolite having a Unified Soil Classfication of SM, with maximum dry densities in excess of 105 pcf.
The grain size distribution band of the silty sand saprolites from the area of plant and structures, sampled during PSAR studies, is shown on Figure 2-162. The grain size distribution band of the soils sampled during studies of the borrow areas used for the dam construction is shown on Figure 2-163. The band occupied by tests on samples of soil taken from the materials after placement in the embankment is shown on Figure 2-164. This figure shows that the placed embankment soils are similar in terms of grain size distribution to those soils tested for the PSAR investigation.
The test data for optimum moisture and maximum dry density for the soils tested for the PSAR are shown on Figure 2-165. The optimum moisture/maximum dry density data for the soils from the three borrow areas used, determined by testing samples secured from test pits, are shown in Figure 2-166. This figure, compared to Figure 2-165, shows the materials placed in the dam are similar in terms of maximum dry density and corresponding optimum moisture content to those soils tested for the PSAR investigation.
Permability The permability of the embankment soils is described in Section 2.5.6.6.
Shear Strength Mohr envelopes of static triaxial shear strengths of compacted soils obtained during the PSAR investigation are plotted on Figure 2-167. Mohr envelopes of static triaxial shear tests on compacted soils from the borrow areas used to construct the SNSW Pond Dam are shown on Figure 2-168. Design values are shown on these envelopes and are tabulated in Table 2-94.
These design shear strengths are based on the deeper soils being compacted to 96 percent of the standard Proctor maximum dry density. The test procedures are presented in Section 2.5.4.2.2.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 58 (09 OCT 2016)
The Mohr envelopes of static triaxial shear strengths of sampled in-place embankment materials are plotted on Figure 2-169. The locations of these samples are shown on Figure 2-170. The consolidation-undrained corrected for pore pressure (R) test results shows that the strengths of a few samples are below the design curve at normal pressures of less than only 2.5 KSF. An analysis, done to investigate the effect of assuming this lower bound of reduced shear strength of the stability of the SNSW Pond Dam, concludes that the embankment satisfies the design stability criteria in Section 2.5.6.5.
The dynamic triaxial shear test results are discussed in Section 2.5.6.5.
Consolidation Figure 2-171 summarizes the range of consolidation characteristics determined during PSAR studies and from samples obtained from the borrow areas used in construction of the SNSW Pond Dam. These samples were compacted to 96 percent of the standard Proctor maximum dry density. Comparison of this data shows the soils actually placed to build the embankment have similar consolidation characteristics to those tested for the PSAR. Settlement analysis of the dam and design consideration thereof are discussed in Section 2.5.6.4.4.
The basis for using 96 percent standard Proctor maximum dry density for compacting Group I fills of site soils is discussed in Section 2.5.4.2.4.2. Field tests performed and controls exercised to assure proper compaction of fills of the site soils are described in Sections 2.5.6.4.5 and 2.5.6.4.9.
2.5.6.4.3 Slope Protection Slope protection is provided on the crest, upstream and downstream faces of the SNSW Pond Dam as shown on Figure 2-140 and Figure 2-141. The riprap extends from abutment to abutment and is underlain with a one-foot thick filter to provide protection against wave action.
The ability of the dam's protective riprap to resist the effects of wave forces is related to the quantity, layer thickness, and individual weight of riprap stones placed on the dam. Based on a maximum wave height of 4.5 feet and using design methods recommended by the Corps of Engineers in the "Shore Protection Manual" (Reference 85), a median stone weight of 760 pounds and a minimum layer thickness of three feet is determined to be adequate. Grain size distribution test data for the one-foot thick filter beneath the riprap layer is shown on Figure 2-172. Riprap stone consists of hard, adamellite bedrock material. Adherence to riprap (large stone) design gradation requirements is ensured by visual inspection and the weighing of representative stones. A listing of the equipment used to place the slope protection is provided on Table 2-96.
Riprap Class 1A (two-foot thickness of stone ranging in size from 3 inches to 24 inches in diameter with 50 percent greater than 18 inches) underlain with a one-foot thick filter is provided around the Nuclear Service Water Intake Structure, Standby Nuclear Service Water Intake and Discharge Structures, and the Standby Nuclear Service Water Pond Outlet Works Intake Structure for protection against wave action. Riprap in the area of the SNSW Pond Outlet Works Discharge Structure is identical to that placed on the SNSW Pond Dam. The extent of the riprap around all service water intake and discharge structures is a minimum of 10 feet beyond the concrete limits in all directions. The NSW Intake Structure is protected by riprap and a concrete apron to El. 575.0. The SNSW Intake Structure is protected by the riprap around the structure and at pond level by the riprap around the SNSW Pond Outlet Works.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 59 2.5.6.4.4 Settlement and Overbuild Compression of embankment material is calculated for the maximum height of such material under its own weight with the reservoir empty. A profile near boring A 101 is selected for this calculation. This profile includes 75 ft of embankment (overlying 10 ft of partially weathered rock). This calculation, using conservative interpretations of the consolidation tests, indicates a total embankment compression (excluding foundation compression) of 16 inches. This calculated settlement is about two percent of the dam height, a value that is often taken as a guide to estimate anticipated and acceptable compression for earthfill dams constructed of these types of soils. The calculated post-construction crest settlement attributable to compression of 75 ft of embankment soil is estimated at about 2-1/3 inches, based on the consolidation test data. Comparisons between measured and calculated settlements of the embankment crest at a time beginning after construction are given in FSAR Section 2.5.6.10.
To compensate for settlement and to maintain design freeboard, the crest of the dam is conservatively cambered two feet above the finished design elevation at the maximum dam section. The camber used in other areas is proportional to the height of the fill. Settlement monitoring after construction is presented in Section 2.5.6.8.1.
2.5.6.4.5 Fill Placement A listing of the equipment used to construct the embankment is provided in Table 2-96.
Throughout the placement of compacted earthfill, the distribution and gradation of materials are controlled to assure that no lumps, pockets, or layers of material differing substantially in texture and gradation from surrounding materials occur. After dumping, the material is spread in a nine-inch thick horizontal or nearly horizontal layer over the fill area. Harrowing, if required to break up and blend fill material or to obtain uniform moisture distribution is performed after spreading.
Compaction is performed at moisture contents within plus three to minus one percent of optimum, based on Standard Proctor Tests for the particular soil. Moisture checks are performed in accordance with ASTM D 2216 at a frequency of approximately four per day. Soil having a moisture content outside the specified limits is either wetted or dried to adjust moisture content prior to compaction, or removed. Results of moisture checks are presented in Figure 2-174.
After determining that moisture content and other conditions in a layer are satisfactory, the layer is compacted to attain not less than 96 percent of the Standard Proctor Density (ASTM D 698).
Determination of compaction is performed in accordance with ASTM D 2937, Shelby Tube Method, for each change in soil type or borrow source, and on a frequency of every 2500 cubic yards placed. Locations and results of field density tests are shown on Figure 2-170 and Figure 2-175. In addition, random undisturbed samples of the field compacted fill are obtained and tested in laboratory triaxial shear to verify that the as-constructed embankment meets or exceeds design shear strength requirements. Figure 2-170 shows spatial distribution of samples taken for Triaxial Shear Tests. With each Standard Proctor Test (performed at a frequency of approximately two per day and at each minor change in soil type or borrow area source), tests are performed to determine grain size distribution (ASTM D 422), liquid limit (ASTM D 423), and plastic limit (ASTM D 424) of the soil material. Refer to Section 2.5.6.4.2 for results of the triaxial shear and classification tests.
2.5.6.4.6 Protection Required of Fill Surfaces and Stockpiles During Construction During construction, the central portion of the dam is maintained slightly higher than the sides so that the top of the fill drains freely toward the side slopes.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 60 (09 OCT 2016) 2.5.6.4.7 Blanket Drain A zoned blanket drain is located in the downstream side of the dam below original ground elevation 570, and a toe drain above elevation 570. The blanket drain extends one quarter of the dam width into the dam, as shown on Figure 2-140 and Figure 2-176. The fine and coarse filter materials are placed and compacted to a minimum relative density of 80 percent in accordance with ASTM D 2049. ASTM D 2049 has been withdrawn and replaced by ASTM D 4253 and ASTM D 4254. ASTM D 4253 and ASTM D 4254 may be used in lieu of ASTM D 2049. Field measurements of gradation and density are performed at a minimum frequency of once a day or every 400 tons of material placed. Materials that fail the gradation test are removed. Materials that fail the density test are recompacted or removed.
Details of the drain are shown on Figure 2-141. Filter material gradation requirements and gradation test results are shown on Figure 2-178 through Figure 2-180.
Field density test results are shown on Figure 2-176. A discussion of the filter gradation is presented in 2.5.6.6.3.
2.5.6.4.8 Special Fill Placement Activities Foundation preparation requires controlling groundwater in several areas. Temporary foundation drains are placed in such areas as shown on Figure 2-159. The drains consist of trench excavations filled with and having at the low point vertical pipe arrangements for pumping. The granular materials consist of No. 67 coarse filter gravel overlaid by an approximately one-foot thick fine filter of concrete sand. The vertical pipe is constructed of open-ended barrels placed end to end as the fill progresses. The maximum vertical height of such barrels is approximately 35 feet. Periodic pumping maintained low water levels in the drain during foundation preparation and initial fill placement.
The locations of the foundation drains are such that they do not affect the stability of the dam.
However, for conservatism, grouting programs were initiated after the embankment was placed to an elevation such that the drains were no longer required. Grout take was minimal. The vertical barrels were filled with concrete.
2.5.6.4.9 Construction Specifications Pertinent specification requirements relating to embankment construction are noted below:
2.5.6.4.9.1 Embankment Materials The materials for structural fill to construct the SNSW Dam will come from on-site areas designated by the Engineer.
After completion of foundation preparation as given in Section 2.5.6.4.8, placement and compaction of material shall proceed as specified for Group 1 fill. When structural fill has risen to a level above the original lake bottom, the wedge between the fill at the dam toe and existing alluvium material in the lake bottom shall be filled with soil.
To compensate for settlement the dam crest shall be cambered two feet above the finished design elevation in areas of deeper fill. Less camber will be used in other areas, proportional to the height of fill.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 61 2.5.6.4.9.2 Construction Control Quality control monitoring for construction of the SNSW Pond Dam will include inspection and field testing of the foundation preparation to verify the appropriate foundation preparation has been performed as described in Section 2.5.6.4.8. The compaction of the structural fill will be monitored for every borrow area, each change in soil type, and on a frequency of every 2,500 cubic yards placed. In addition, random undisturbed samples of the field-compacted fill will be tested in triaxial shear to verify that the embankment meets or exceeds design shear strength parameters.
2.5.6.4.9.3 Compacted Earth Fill - Group 1: Compacted Earth Fill - Group 1 - Backfill For Structures 2.5.6.4.9.3.1 Lines, Grades, and Cross Sections The earth fills shall be constructed to the elevations, lines, grades, and cross sections indicated on the contract drawings.
Increased heights and widths may be required to compensate for later shrinkage or settlement, but in no case will increases exceed five percent of the height indicated on the drawings.
The Owner reserves the right to vary the foundation widths and the fill slopes, and to make other changes in the sections of the compacted fills, when conditions indicate that they are necessary for the construction of safe and permanent structures. Section changes which result in changed material quantities will not be cause for revision of unit prices.
2.5.6.4.9.3.2 Origin of Fill Materials An earth movement plan and earth movement location drawing are included in this Specification for use in determining origin of materials and its placement. The Engineer will direct the Contractor of any changes he may require.
Material for compacted earth fills - Group 1 shall be obtained, insofar as possible from areas of required excavation on the project site. Should either the quality or quantity of such excavated material become unsuitable or inadequate for use in constructing fills, the Engineer will direct the Contractor to open borrow pits at specified locations adjacent to the project site. Materials from these borrow pits will then be utilized in the completion of fills.
2.5.6.4.9.3.3 Materials Excluded From Group 1 Fills The following materials shall not be placed in the fill included in this item:
Rocks larger than 12" in maximum dimension, topsoil, excessively wet soil, logs, brush, roots larger than 1" in diameter, sod, perishable materials, snow, ice, frozen earth, or any other materials which are disapproved by the Engineer.
2.5.6.4.9.3.4 Removal of Unsuitable Materials The Contractor shall excavate and remove from the fills, after its placement therein, any material which the Engineer considers unsuitable. Such unsuitable material shall be wasted in spoil areas shown on the drawings. If the presence of unsuitable material in the fills is through no fault of the Contractor, payment for the quantity removed will be made at the contract unit price for general site grading excavation, and the replacement quantity will be paid for at the contract unit price for compacted earth fill.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 62 (09 OCT 2016) 2.5.6.4.9.3.5 Preparation of Fill Foundations After stripping of foundations and removal of weak or unsuitable materials has been completed, and before the start of material placement, Group 1 Fill shall be used to fill all test pits, stump holes, minor excavations and depressions or cavities inside the earth fill limits. These fillings shall be placed in layers over areas of practical size, and shall be moistened or dried to obtain a moisture content within plus or minus three percent of optimum and then compacted by roller equipment in accordance with the applicable provisions of Subsections 2.5.6.4.9.3.7 and 2.5.6.4.9.3.8. In smaller areas where roller equipment cannot be employed, material shall be spread in layers and compacted with power tampers to the degree specified by the Engineer.
After the filling of depressions and prior to the placement of compacted fill, the entire earth fill foundation shall be proof rolled to detect any soft surface zones. Proofrolling shall be done with a 50 ton rubber tired or roller or a fully loaded 20 yard pan. A total of four passes of the roller or pan shall be made over each spot of the foundation, two in mutually perpendicular directions.
Areas that continue to deflect more than 1 to 2 inches after two or three repeated passes of the roller or pan shall be undercut and replaced with fill compacted as specified in Subsection 2.5.6.4.9.3.8.
A cone penetrometer, whose calibration applies to foundation area being tested, may be used in lieu of proofrolling to determine acceptance of the foundation.
After proofrolling or acceptance of the foundation, the entire foundation shall be thoroughly loosened to a depth of 3 to 6 inches by scarifying, plowing, disc harrowing or other approved method. After scarifying, the loosened area shall be covered by a layer of fill material having proper moisture content and being placed to a depth of not more that 6" loose measurement.
The scarified and deposited soils shall then be thoroughly mixed with a disc harrow or other approved equipment. The area so treated shall then be compacted, as specified in Subsection 2.5.6.4.9.3.8.
2.5.6.4.9.3.6 Placement of Material Fills After the foundation has been prepared, and the first bonding layer of fill has been placed, all as specified in Subsection 2.5.6.4.9.3.5, the Contractor shall proceed to place the compacted fill.
Throughout compacted earth fills, the distribution and gradation of materials shall be such that there will be no lumps, pockets or layers of material differing substantially in texture and gradation from surrounding materials. After dumping, the materials shall be spread by bulldozers, motor graders, scrapers, pans or other approved means in approximately horizontal layers over the entire areas to be filled. Unless otherwise directed, the thickness of the layers, prior to compaction with tamping type rollers, shall be not more than 9". After a layer of fill material has been dumped and spread, it shall be harrowed, if required, to break up and blend the fill materials, or to obtain uniform moisture distribution. Each successive layer shall be compacted in accordance with Subsection 2.5.6.4.9.3.5.
During the dumping and spreading process, small stones and gravel shall not be allowed to accumulate at any point so that they form a pervious lens or pocket in the fill.
As soon as practical after starting construction of a fill, the central portion thereof shall be built and maintained slightly higher than the sides so that the top of the fill will drain freely toward the side slopes. The transverse grades on top of the fills shall not exceed three percent. The fill surfaces shall be maintained in a free draining condition throughout construction.
Wherever the fills join hardened or dried surfaces of an original abutment, or the compacted surface of any layer of fill material is determined to be too smooth to bond properly with the
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 63 succeeding layer, the hardened abutment or fill shall be thoroughly loosened by scarifying, harrowing, plowing or other approved method, to a depth of 3 to 6 inches.
Equipment shall not travel continuously over the same routes. All traveling shall be along the length of the fills insofar as possible, except for turning movement. Ruts shall be continuously broken up to facilitate proper bonding and to eliminate possible seepage paths. When the fills have been otherwise completed, the Contractor shall dress the slope surfaces to the required limits of fill as directed by the Engineer.
2.5.6.4.9.3.7 Moisture Control Compaction shall be done at moisture contents within plus or minus three percent of the optimum for all Group 1 compacted earth fills and earth backfills except for the standby nuclear service water pond dam which shall be held to within plus three to minus one percent of optimum. Regardless of source, materials that are too wet to permit proper compaction shall not be spread on the fill until the moisture content is satisfactorily reduced. When material is too dry, each layer of the fill shall be sprinkled prior to its compaction. This water shall be worked into the material, by harrowing or otherwise, until a uniform distribution of moisture has been obtained. The amount of water so applied shall be sufficient to dampen the material to the required moisture content, and its application shall be so controlled that no free water will appear on the surface during the rolling process or afterwards. Should too much water be added to any part of the embankment so that the material is too wet to obtain the desired compaction, the rolling on that section of the embankment shall be delayed until the moisture content of the material is reduced to within the specified limits. Water jets shall not be directed at the embankment with enough force to cause separation of materials. If it is found impractical to add sufficient moisture to materials on the fill, the materials shall be prewetted at the source of excavation. The Contractor shall have an adequate supply of water available at all times.
2.5.6.4.9.3.8 Compaction of Fill Material When the moisture content and other conditions in any layer are satisfactory, that layer shall be compacted by the necessary number of passes of approved rollers required to attain not less than 96 percent or up to 100 percent of the Standard Proctor Density (ASTM D698) as designated in selected locations on the drawing. Quality Control Laboratory Density Test results shall be the basis for determining acceptance of Group 1 Fill.
Sheepsfoot or alternate types of rollers, subject to the approval of the Engineer shall be used to obtain compaction of Group 1 Fill. The entire areas of each layer shall be thoroughly covered by the number of passes necessary to produce a uniform compaction. When there is sufficient area, separate dumping, spreading, sprinkling, and compacting operations may be carried on at the same time at different places on the fills.
2.5.6.4.9.3.9 Slides in Earth Fills In the event of slides in any part of the fills during construction, or after completion, but prior to final acceptance of work, the Contractor shall remove the volume specified by the Engineer, and shall replace it in accordance with these specifications. If the slide was caused by fault of the Contractor, its repair shall be made without cost to the Owner; otherwise, the excavation and removal shall be done at the contract unit price for earth excavation and the refill shall be done at the contract unit price for compacted earth fill - Group 1.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 64 (09 OCT 2016) 2.5.6.4.9.3.10 Protection for Earth Fills Responsibility for the protection of earth fill slopes against erosion shall rest with the Contractor until such fills are accepted as complete by the Engineer.
2.5.6.4.10 Significant or Unusual Construction Activities There are no unusual construction procedures employed during construction of the SNSW Pond Dam.
2.5.6.5 Slope Stability 2.5.6.5.1 Stability Criteria The dam section is analyzed and designed for slope stability under the following limiting conditions:
Slope Loading Condition Minimum Safety Factor Required End of Construction 1A Upstream Slope 1.25 1B Downstream Slope 1.25 Steady Seepage Cooling Pond Elevation=571 Lake Wylie Elevation=550 (Loss of Lake Wylie Dam) 2A Upstream Slope 1.50 2B Downstream Slope 1.50 Sudden Drawdown of Lake Wylie Elevation 591.8 (Maximum Flood Level) to Elevation 550 (Loss of Lake Wylie Dam)
Downstream Slope 1.25 Sudden Drawdown of Cooling Pond Elevation 587.0 to Elevation 567.0 Upstream Slope 1.25 Steady Seepage (Seismic-Finite Element Analysis) with SSE Superimposed, Normal Full Pond Elevation 571 and Lake Wylie Level 569.4.
1.05 2.5.6.5.2 Design Methods The SNSW Pond Dam is a homogeneous earthfill structure which is designed using the stability criteria presented in Section 2.5.6.5.1 and design shear strength parameters for foundation and
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 65 embankment materials as presented in Sections 2.5.6.2.1.2 and 2.5.6.4.2 and on Table 2-94.
Figure 2-181 shows the representative dam cross-section selected for analysis. It is representative of the as-built dam cross-section for greatest height above partially weathered rock, which occurs in the areas between Stations 1+75 and 6+50. The selected analysis cross-section includes a foundation zone of firm to dense/very dense coarse grained saprolite between the embankment and partially weathered rock. As shown on Figure 2-140, this zone of material is removed during the foundation preparation activities up to about Station 4+00, and replaced with compacted fill, bearing directly on the partially weathered rock.
The static stability analyses results described in the following sections are representations for the cross-sections between the Stations 1+75 and 4+00. These values are conservative because the static shear strengths assigned to the firm to dense/very dense coarse-grained saprolite soils for analysis purposes are slightly lower than the values assigned to compacted structural fill. (See Table 2-94)
The dynamic stability analyses are likewise conservative for the cross-sections between about Stations 1+75 and 4+00 because the shear modulus of the coarse grained residual soils is higher than that of the compacted fill (Figure 2-182). As stated in Section 2.5.6.5.4, the use of a higher shear modulus gives more conservative results for induced cyclic shear stresses in the embankment. Hence, the stability analyses results for the cross-section stratigraphy shown on Figure 2-181, including the foundation saprolite zone, are conservative from both the static and dynamic viewpoint.
2.5.6.5.3 Static Stability Evaluation Static stability analyses of the dam are performed for: (1) end of construction condition; (2) steady state seepage through the dam with the worst combination of water levels; (3) sudden drawdown of the Lake Wylie water level from maximum flood level; (4) sudden drawdown of the Standby Nuclear Service Water Pond.
For the end of construction condition, prior to pond filling, the unsaturated unconsolidated-undrained shear strengths (Q) are used. The shear strength is expressed as total stresses and stability analysis is done in terms of total stresses. This analysis contains the implication that the field pore pressures will not exceed those experienced in laboratory tests.
The critical steady state seepage condition occurs with the pond level stabilized at the maximum storage level which can be maintained for a period sufficient to produce seepage throughout the embankment. The saturated-consolidated-undrained, corrected for pore pressure shear strength
(
S R
) is used in this stability analysis. These tests are conducted on saturated samples, which simulates the saturated condition of the soil in the embankment. The analysis is done in terms of effective stresses, using hydrostatic pressures as a conservative assumption for the pore pressures present. The analysis ignores capillary action, which adds conservatism to the pore pressure estimates by not attempting to account for increased effective stress (and greater soil strength resulting from capillary tension) (Reference 86).
The sudden drawdown condition applies where the water level is lowered faster than excess moisture can escape from the embankment. The saturated-consolidation-undrained, not corrected for pore pressure, or total (R) shear strengths are used in this stability analysis. The downstream and upstream face are assumed to be subject to drawdown and require analysis.
The zone analyzed is conservatively considered saturated from the maximum flood level to the maximum drawdown level.
In the analyses, a rotational shear displacement of a segment of the embankment and/or its foundation is presumed and the safety factor defined as the ratio of the resisting moments
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 66 (09 OCT 2016) provided by the soil shear strength and weight to the sum of the moments tending to produce motion. A computer program is used to allow many different assumed circular arc failure surfaces to be checked. In addition, hand calculations of some trial failure surfaces are performed. All calculations are performed using the circular arc and method of slices (Reference 87).
The results of the static stability analyses are presented in Table 2-97 and Figure 2-184. These results show the dam exceeds the design criteria for factor of safety given in Section 2.5.6.5.1.
Evaluation of wedge failure modes is not considered necessary due to the relatively small differences in shear strength assigned to the saprolite soil and embankment material.
Foundation preparation removed all saprolite foundation soils having standard penetration test blow counts less than about 27 blows from the deep area of the dam embankment.
2.5.6.5.4 Dynamic Stability Evaluation The analyses of the seismic loading effects are accomplished with finite element procedures.
The following studies are required:
- 1. Determination of the embankment stresses under static conditions before any seismic loading for normal pond and full lake level.
- 2. Evaluation of the dynamic moduli of the embankment and foundation materials by appropriate laboratory resonant column tests, as described in Section 2.5.4.2.2.7.
- 3. Evaluation of the dynamic properties of the foundation materials in-situ by field measurement of shear wave and compression wave velocities at various depths, as described in Section 2.5.6.2.
- 4. Evaluation of the available cyclic shear strength of the embankment material by appropriate laboratory cyclic triaxial shear testing, as described in Section 2.5.4.2.2.7.
- 5. Use of the synthetic time histories of bedrock accelerations produced by the SSE; and
- 6. Determination of the induced dynamic shear stresses throughout the embankment, comparison of these shear stresses to the available cyclic shear strengths, and evaluation of the embankment safety.
The reservoir elevation-time relationships from flooding as computed for the original full pond elevation of 571 feet for the Standby Nuclear Service Water Pond are given below. This information is valid for the original dam design and analysis.
Flood Condition Reservoir Elevations at Hours Shown After Storm Begins 0
5 hrs 11 hrs 16 hrs 26 hrs 25 Yr. Flood 571.0 572.5 (Peak) 571.7 571.2 571.0 0
6 hrs 11 hrs 16 hrs 26 hrs 31 hrs 40 hrs Standard Project Flood 571.0 576.2 (Peak) 574.7 573.4 571.3 571.2 571.0 The above information shows that the maximum height of reservoir rise above normal pond is only 1.5 ft for the 25 year flood and only 5.2 ft for the Standard Project Flood.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 67 Note that a stability evaluation was done for the modified full pond elevation of 574 feet as changed by a modification in 1997. This evaluation used similar methods to those described below. The later stability evaluation took into account more conservative (higher) elevation for flood water and found the dam to be stable for the revised more conservative elevations.
The following information was provided for the original design, analysis and filling of the dam. It is still valid. The modification done in 1997 raised the full pond elevation slightly, but this higher elevation is insignificant for the analysis. Information on the original analysis is left as is for historical purposes.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE UPDATED Due to the permeability characteristics of the embankment soil, there is a lag time between reservoir elevation changes and the increase of pore pressure (rise in phreatic surface) inside the embankment. Approximate theoretical calculations indicate it would take tens of days per foot of reservoir rise for the phreatic surface beneath the upstream slope to adjust. This is also demonstrated in the piezometric readings plotted on Figure 2-201. The SNSW pond filled to its normal elevation 571.0 on August 27, 1978 (see notes on Figure 2-202); piezometer P-1 and P-5 (see Figure 2-140) in the embankment material beneath the upstream slope did not begin reading water (above their respective tip elevations) until the end of February, 1979, or 6+
months later and 3 months after the downstream slopes of the dam had been flooded by breaching the cofferdam to Lake Wylie. After the end of February, 1979, the phreatic level in P-1 and P-5 continued to rise slowly in response to the pond and lake against the upstream and downstream slopes, respectively.
The reservoir level after flooding recedes to within less than 0.5 ft above normal pond within less than a day after reaching its peak. Due to the presence of lag time, as demonstrated by the piezometric readings discussed above, the embankment would experience little or no increase in pore pressure during this 1 day storm surcharge. Thus, a loading condition combining the flood levels with earthquake loading is considered unnecessary, as the pore pressures in the embankment would be essentially unchanged from those used in the analyses discussed later in this section.
The following discussion will demonstrate that the embankment is conservatively analyzed and adequately satisfies the design criteria for loading by the SSE. Due to its lower acceleration, the OBE would induce significantly lower cyclic shear stresses in the embankment than would the SSE. Analysis of the embankment for earthquake loading by the OBE is thus considered unnecessary because of the less severe loading compared to the SSE.
The embankment stresses under static conditions are analyzed using the non-linear stress-strain properties and finite element techniques and procedures outlined in Reference 88. The finite element representation of the embankment is shown on Figure 2-185. The static nonlinear soil parameters are shown in Table 2-94. The results of the static finite element analysis are shown on Figure 2-186, Figure 2-187 and Figure 2-188.
The cyclic laboratory testing to determine the dynamic shear strength consists of tests on remolded, saturated isotropically and anisotropically consolidated samples of embankment soil, with the results shown in Table 2-98 and Figure 2-152. The testing procedures and test data interpretation follow recommendations given by Seed and others (References 89, 90, 91, and
- 92) and Gordon, et. al., (Reference 93). The results of typical tests are presented on Figure 2-189. The failure criterion used is the stress required to cause 5 percent axial strain (from original height at the beginning of the cyclic test) in 10 cycles in the laboratory tests. The equivalent number of uniform cycles of seismic loading was assumed based on information by Seed and Idriss as described in 1972 by Lee and Chan (Reference 94). Conclusions contained in Seed,
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 68 (09 OCT 2016) et. al., (Reference 109 that for earthquakes of the size of the SSE for Catawba, five equivalent uniform cycles of stress at 65 percent of the maximum stress provdies an adequately conservative representation of such events (page 11 of Seed, et. al., 1975). Ten cycles loading were used in planning the cyclic triaxial tests and for determining the dynamic shear strength of the soil, which is thus conservative for the SSE. The results of the cyclic loading tests, expressed as field strength, 10 f )
(
, are summarized in Figure 2-153.
The vertical component of earthquake motion does not appreciably affect the shear stresses calculated on horizontal planes in an earth dam (Reference 117). Hence, only horizontal motion is employed to calculate the shear stresses induced in the embankment. For the SNSW pond embankment, however, a vertical motion equal to two thirds the horizontal motion was included.
For multidirectional horizontal shaking, level ground in the "free field" is affected by both horizontal components of earthquake motion. A dam, however, may respond much less in the longitudinal direction (along the length) then in the transverse direction; such evidence for a preferred upstream-downstream direction of motion appears in the seismoscope record of the Lower San Fernando Dam in the 1971 California earthquake. For the analyses of shear stresses induced in the SNSW Pond Dam, only upstream-down-stream horizontal shaking is considered appropriate. This is consistent with the usual practice in analyzing earth dams (Serf, Seed, Makdisit, Chang, 1976).
The seismic response and dynamic shear stresses induced in the embankment are analyzed with the computer program QUAD-4, developed at the University of California at Berkeley (Reference 95). This program allows, as input, both horizontal and vertical earthquake motions, and modulus and damping values that are strain dependent for each finite element of the mesh.
The static finite element analysis provides values of initial effective confining stress. These data are then used in conjunction with the relations of Figure 2-182 to assign the initial modulus values to each finite element. Full Lake Wylie level is used in the analysis since this produces lower normal stresses within the embankment than would be obtained with a lower Lake Wylie level and thus produces lower safety factors under cyclic loading. Normal SNSW Pond level is used. Higher pond or lake levels corresponding to flood conditions are considered inappropriate for anlyses of cyclic loading by the SSE, since this would involve combination of extreme events (flood plus SSE). Analyses of earth structures involving the OBE are considered unnecessary due to less severe cyclic loading (compared to SSE).
Figure 2-190 shows the relationship determined to represent the "most probable" damping values at various strain levels for the embankment and foundation material while Figure 2-191 shows the "most probable" values of shear modulus parameter (K2max) for the materials. Figure 2-157 presents the variation of shear modulus with strain. Computations are done to evaluate what effect some variation in damping and modulus has in the computed cyclic shear stresses.
Holding the damping to the "most probable" values on Figure 2-190, the most probable K2max values are increased by 25 percent and the cyclic shear stresses computed. The result is that cyclic shear stresses increase up to 30 percent in the saprolite and up to 28 percent in the embankment. The elements in the center of the embankment, where the local safety factor (see below) is lowest experience 22 to 24 percent increase in cyclic shear stress for the 25 percent increase in K2max. Next, holding the K2max values to the most probable values, the damping is decreased to the lower bound condition shown on Figure 2-190. The result is that elements in the saprolite experience up to 30+ percent higher cyclic shear stress, and those in the embankment experience up to 25 percent higher cyclic shear stress. Thus, the conclusion is that the most conservative combination of modulus and damping is the highest modulus at the same time with lowest damping. The lowest damping used is ultraconservatively taken 50 percent below the most probable values within the range of computed shear strains experienced during the earthquake.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 69 The calculations indicate a first natural period of the embankment foundation section analyzed, when subjected to the earthquake loading, of 0.4 to 0.5 seconds for high modulus, low damping, and 0.5 to 0.6 seconds for "most probable" values of modulus and damping. Reference to the response spectra for SSE conditions on Figure 2-110 shows that, for a range of periods less than 0.5 to 0.6 seconds, the spectral acceleration are at about their maximum values. Thus, the high modulus-low damping condition placed the embankment in the region of highest spectral accelerations. Hence, this combination of modulus and damping properties for all three materials simultaneously is judged to be adequately conservative and additional combinations of modulus and damping for the embankment, saprolite and partially weathered rock materials judged to be unnecessary.
The combination of 25 percent high modulus-50 percent low damping is used to compute the cyclic shear stresses for the finite element evaluation of the embankment and foundation.
Output from the QUAD-4 program includes the maximum values of horizontal and vertical accelerations at each node point, along with their time of occurrence, the time history of cyclic shear stress in each element, and the time history of cyclic shear strain in each element. The peak cyclic shear stresses induced by applying each of the four synthetic time histories shown in Figure 2-193, Figure 2-194, Figure 2-195, and Figure 2-196 are computed. The cyclic shear stresses 10 xy)
(
used in the subsequent evaluation of embankment stability are the average of four values of peak shear stress thus computed for each element under consideration, multiplied by a reduction factor of 0.65 as recommended by Seed to represent an equivalent uniform cyclic shear stress. The static finite element analysis results for initial shear stress and effective normal stress are used in conjunction with the test results on Figure 2-153 to determine the uniform cyclic shear stress 10 f )
(
required to cause 5 percent strain potential at any location within the dam below the line of saturation.
The dynamic shear strength 10 xy)
(
is not added to the static shear stresses in calculating the factor of safety of each element.
The safety of the dam is then evaluated by comparing the shear stresses induced during the earthquake motion 10 xy)
(
with the shear stresses required to cause 5 percent strain 10 f )
(
. If the latter stresses everywhere exceed the induced stresses by an appropriate margin (Section 2.5.6.5.1), the dam is expected to have adequate safety during the SSE.
The results of the evaluations of cyclic shear stress response and the available cyclic shear strength are presented in Figure 2-197. The evaluations show that the available cyclic shear strength 10 f )
(
exceeds the induced shear stresses
)
(
f throughout the embankment. The factory of safety, defined as 10 xy 10 f
)
(
)
(
has a minimum local value of 1.06 at the embankment centerline of surface 2 and has minimum values of 1.14 and 1.13 on surfaces 1 and 3, respectively. This exceeds the minimum required safety factor criterion of 1.05 (Section 2.5.6.5.1).
For SSE loading, the local factor of safety for each finite element is tabulated on Figure 2-200.
As described above, the lowest local factor of safety for all locations in the dam is 1.06, which occured at the embankment center line of surface 2 (Figure 2-197).
The lowest row of elements on Figure 2-200 represents the partially weathered rock. Due to its nature, is unnecessary to include this material in the cyclic strength determination. The upper rows of elements on Figure 2-200 represent embankment material above the line of saturation, which condition makes it inappropriate to include these elements in the cyclic strength determination.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 70 (09 OCT 2016)
The Newmark method was applied to the embankment for estimating any permanent deformations due to the seismic shaking. The original concept as proposed by Newmark (1965)
(Reference 118) was broadened to consider the amplifying characteristics of the embankment rather than assuming simply rigid body behavior. The method involves the following steps:
- 1. A yield acceleration (i.e., an acceleration at which a potential sliding surface would develop a safety factor of unity) is determined.
- 2. Earthquake induced accelerations in the embankment are determined using dynamic response analyses, and average accelerations (seismic coefficient) for various potential sliding masses are determined.
- 3. For a given potential sliding mass, when the reduced acceleration exceed the calculated yield acceleration, movements are assumed to occur along the direction of the failure plane and the magnitude of the displacement would be evaluated by double acceleration procedures. For conservatism, the potential sliding mass was accelerated in a direction perpendicular to the critical circle radius at its intersection with the mass center of gravity.
Newmark (1965) (Reference 118) and Franklin and Chang (1977) (Reference 119) have made such displacement calculations for dams of various heights and their results may be utilized to estimate the range of displacements for a given dam being analyzed.
The seismic coefficients were calculated using the shear beam method developed by Seed and Martin (1966) (Reference 120). The seismic coefficients for various elevations within the embankments are as follows:
Height Horizontal Seismic Coefficient for Horizontal Bedrock Acceleration of 0.15g (SSE)
Crest
.404 3/4
.344 1/2
.284 1/4
.211 The seismic coefficients vary with elevation in a manner that is shown from the tabulation above plotted graphically on Figure 2-198. Figure 2-198 is taken from the work of Makdisi and Seed (1977) (Reference 121). Also, shown at the top of Figure 2-198 is the maximum crest acceleration (U +max) of 0.386g computed from the finite element analyses using high modulus and low damping as described previously. Figure 2-198 shows the seismic coefficients used in the Newmark analyses are conservative with respect to seismic coefficients that would be derived based on the peak crest acceleration from the finite elements analyses and using the average line from Makdisi and Seed (1977) (Reference 121).
Franklin and Chang (1977) (Reference 119) show the variation of standarized permanent displacement with the ratio of yield acceleration (resistance coefficient) to maximum earthquake acceleration (seismic coefficient) for 179 baseline-corrected accelerograms. Their work shows, for yield accelerations corresponding to 0.8 to 0.9 times the maximum acceleration (seismic coefficient), the computed displacement is 1 inch or less. This finding essentially agrees with Newmark's work.
The analyses of the SNSW Pond Dam yielded a Newmark's safety factor of minimum 1.12 (Table 2-101) or a yield acceleration to maximum earthquake acceleration ratio greater than 1.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 71 This leads to the conclusion that the computed permanent displacement of the embankment as a result of shaking by the SSE is less than an inch.
Conservatism is present in the arbitrary adoption of 5 percent axial strain in the laboratory to represent failure of the test samples during cyclic triaxial testing. The failure axial strain in the conventional static triaxial compression tests is found to range from 6 to 20 percent for the embankment material. The shear strain at failure is approximately on the same order of magnitude as the axial strain at failure in laboratory triaxial tests, after correcting for plane strain conditions. The maximum shear strains computed for elements along surfaces 1, 2, and 3 are all less than 0.3 percent, with the largest strains occurring near the center of the dam. These strains are considerably less than the 5 percent axial strain value assumed for interpreting the cyclic triaxial tests.
2.5.6.5.5 Conclusions It is concluded that the SNSW Pond Dam meets the stability criteria from Section 2.5.6.5.1.
2.5.6.6 Seepage Control 2.5.6.6.1 Permeability of Foundation and Embankment Soils No drill water loss occurs in any of the preconstruction borings at the SNSW Pond Dam site.
The permeability of the foundation materials for the SNSW Pond was determined as discussed below.
Dam embankment is further investigated by in-situ permeability tests and by laboratory tests on relatively undisturbed samples of the foundation materials as described in Section 2.5.4.2. The locations of the permeability tests are shown on Figure 2-145, Figure 2-147, and Figure 2-149.
Permeability of the in-situ sandy silts is estimated to be in the range of 3 to 50 ft per year, with a representative value of 40 ft per year or 4 x 10-5 cm/sec. The sandy silts will only form a minor portion of the dam foundation, in the northern end.
Measured in-situ and laboratory permeability of the residual coarse grained (silty sand) saprolites and the partially weathered rock is in the range of 150 to 900 ft per year, with a representative value of 700 ft per year or 7 x 10-4 cm/sec. (Table 2-93 and Table 2-102). This representative value is weighted toward the higher field values which reflect the localized effects of relict joints in the saprolite. The value is conservative for the mass, and is well above the laboratory permeability values.
The measured in-place values for rock permeability at the dam site, which include the effects of rock jointing, range from 0 ft to 470 ft per year. A representative value for an upper limit of rock mass permeability is 6 x 10-4 cm per second, or 600 ft per year. This conservative upper limit is exceeded by none of the tests beneath the dam, and is exceeded significantly by only one of the approximately 60 tests performed at the nearby plant site.
Excavating and compacting the on-site soils destroys the relict joints and cracks present in the in-place materials, thus reducing the permeability. The representative permeability value for the compacted sandy silts is 6 x 10-7 cm per second, or about 0.6 ft/year. For the compacted silty sandy saprolites, the major soil type in the dam, a representative permeability is 2 x 10-6 cm per second, or about 2 ft/year. The range of test values is 0.1 ft/year to 10 ft/year. This permeability information is based on laboratory tests performed on compacted samples and shown in Table 2-103.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 72 (09 OCT 2016) 2.5.6.6.2 Seepage Analysis Section 2.5.6.1 describes the SNSW Pond Dam geometry and operating levels. The SNSW Pond operation maintains a water level between full pond elevation 574 and maximum drawdown elevation 567. Lake Wylie is operated between full pond elevation 569.4 and maximum drawdown elevation 559.4. Thus, under normal operating conditions, the SNSW Pond Dam will be subjected to differential hydrostatic heads of only 4.6 to 14.6 feet. Should conditions occur such that Lake Wylie Dam fails, the Lake Wylie pool level drops to elevation 550 and the dam is subject to only a 24 ft differential hydrostatic head.
Because of the relatively small differential hydrostatic heads across the dam, both during normal operations and during a postulated loss of Lake Wylie, a seepage loss analysis is deemed not necessary for this embankment and engineering judgment is sufficient to warrant the conclusion that piping and seepage do not pose any problem. Since the maximum head differential to which the dam conceivably could be subjected is 24 ft (elevation 574 - elevation 550), no concern exists that potential problems relating to seepage at the embankment-foundation interface can develop. The other factors which add to the assurance that no problems will develop due to seepage at the embankment-foundation interface are described below.
The prepared foundation at the downstream toe of the dam in the south abutment area, approximately station 0+50 to about Station 1+50, was excavated a few feet below the original ground surface at the toe. Between approximately Stations 1+50 and 4+00, the prepared foundation was excavated at least 15 ft below the elevation of the original lake bottom (Figure 2-159). From about Station 4+00 to beyond Station 11+00 on the north abutment the prepared foundation was excavated 15 to 25 ft below the ground surface at the downstream toe. After construction and seepage monitoring, the excavated areas outside the slope limits of the embankment were backfilled with soil (Figure 2-140). Thus, from about Station 1+00 to beyond 11+00 the prepared foundation at the downstream toe of the dam is buried 15 to 25 ft below the ground surface at and beyond the toe, and no potential for erosion and piping due to seepage at the embankment-foundation interface exists since the seepage would not have an exit, and the maximum head differential is low.
The south abutment from Station 0+00 to Station 2+00 is an area of the foundation consisting of weathered to unweathered bedrock. Irregularities were exposed, cleaned and filled with dental concrete as shown on Figure 2-159. Then, the prepared foundation was slush grouted as shown on Figure 2-159. This treatment precludes the possibility of any problem with erosion or piping at the embankment soil by seepage at the embankment-foundation interface.
2.5.6.6.3 Embankment Drainage An internal drainage system is not essential for adequate performance of this embankment because of the small differential heads present in the embankment, even under rapid drawdown. However, to provide added conservatism, a blanket drain is installed in the embankment, as described in Section 2.5.6.4.7, to primarily control any rapid drawdown pore pressures on the east slope of the dam facing Lake Wylie, should the lake level be rapidly lowered. A zoned blanket drain, consisting of a 6-in. or more thick layer of free draining material sandwiched between 6-in. or more thick fine filter layers is placed below elevation 570 as described in Section 2.5.6.4.7.
For the blanket drain, Figure 2-178, Figure 2-179, and Figure 2-180 show the gradation limits of the fine and coarse filter material derived from field gradation tests. Shown on Figure 2-164 is the fill gradation. The gradations shown and materials used satisfy the following criteria:
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 73 For:
Particle Retention
- 1.
5
)
Base
(
85 D
)
Filter
(
15 D
For:
Particle Retention
- 2.
25
)
Base
(
50 D
)
Filter
(
50 D
For:
Free Drainage
- 3.
20
)
Soil
(
K
)
Filter
(
K or 4
)
Base
(
15 D
)
Filter
(
15 D
The following values were used in computing the filter criteria:
Maximum D15 (Coarse) = 6.5 mm (Figure 2-178)
Minimum D15 (Coarse) = 2.6 mm (Figure 2-178)
Minimum K of Coarse = 14,300 ft/year (Table 2-104)
Average D50 (Coarse) = 9 mm (Figure 2-178)
Minimum D85 (Coarse) = 13 mm (Figure 2-178)
Average D15 (Fine) = 0.14 mm (Figure 2-179, Figure 2-180)
Log Average K of Fine = 360 ft/year (Calculated from Table 2-104)
Average D50 (Fine) = 1 to 1.5 mm (Figure 2-179, Figure 2-180)
Average D85 (Fine) = 3 mm (Figure 2-179, Figure 2-180)
Minimum D85 Fill = 0.40 mm (Figure 2-164)
Average D50 Fill = 0.20 mm (Figure 2-164)
Average D15 Fill = 0.013 mm (Figure 2-164)
Log Average K of Fill = 2 ft/year Section 2.5.6.6.1 The following calculations verify the filter criteria are satisfied for the blanket drain:
Fill and Fine Filter 35
.0 40
.0 14
.0 Fill 85 D
Fine 15 D
5.7 to 5
20
.0 5.1 to 1
Fill 50 D
Fine 50 D
180 2
360 Fill K
Fine K
and 8.
10 013
.0 14
.0 Fill 15 D
Fine 15 D
Fine Filter and Coarse Filter 2.2 3
5.6 Fine 85 D
Coarse 15 D
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 74 (09 OCT 2016) 9 to 6
5.1 to 1
9 Fine 50 D
Coarse 50 D
40 360 300 14 Fine K
Coarse K
and 5.
18 14
.0 6.2 Fine 15 D
Coarse 15 D
The toe drain shown on Figure 2-141 lies completely above elevation 570.0 and therefore does not perform as a filter. The small quantity of material used for the toe drain is the same material that was used for the blanket drain coarse and fine filters. Two gradation tests were used to control classification of the fine layer of the toe drain. The gradation curves for those tests fall within the limits of test results for the blanket drain fine filter shown on Figure 2-172.
Three gradation tests were used to control classification of the coarse layer of the toe drain. The gradation curves for those tests fall within the limits of test results for the blanket drain coarse filter shown on Figure 2-178.
For the upstream and downstream slope protection filter layer, the gradation limits are shown on Figure 2-172 as follows:
Maximum D15 = 0.64 mm Minimum D15 = 0.15 mm Average D50 = 5 mm The following calculation verify that the criteria relating to retention of soil particles are met by the upstream and downstream filter layer:
6.1 40
.0 64
.0 Fill 85 D
Filter 15 D
25 20
.0 5
Fill 50 D
Filter 50 D
As discussed in Section 2.5.6.6.2, the differential hydrostatic head across the SNSW Pond Dam is small, and therefore the filter material on the slope serves only as a uniform, coarse bedding material for the riprap. Details of the riprap are provided in Section 2.5.6.4.3.
2.5.6.7 Diversion and Closure The SNSW Pond Dam is located on a natural arm of Lake Wylie. Prior to the start of construction, cofferdams were constructed upstream and downstream of the SNSW Pond Dam.
After construction of the downstream cofferdam, the area west of the cofferdam was unwatered by pumping and the upstream cofferdam is constructed to prevent upstream runoff from entering the SNSW Pond Dam foundation area. Surface water is routed around the dam by pumping.
After completion of the SNSW Pond Dam, the upstream cofferdam was removed and the pond was filled by pumping from Lake Wylie. The area between the SNSW Pond Dam and the downstream cofferdam was kept unwatered for seepage monitoring as discussed in Section 2.5.6.8.1. Upon completion of the monitoring program, the downstream cofferdam was breeched as shown on Figure 2-20.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 75 2.5.6.8 Performance Monitoring 2.5.6.8.1 Instrumentation Instrumentation is installed in the SNSW Pond Dam as shown and detailed on Figure 2-140, Figure 2-142 and Figure 2-199 and consists of the following:
- 1. Two representative cross-sections of instrumentation, each consisting of three piezometers used to monitor the phreatic surface within the embankment.
- 2. Two piezometers placed into the foundation to monitor foundation pore pressures.
- 3. Twelve surface monuments placed on the crest along the full length of the dam to measure post construction settlement.
- 4. A temporary drainage ditch with V-notched weir along the downstream toe of the dam used to monitor seepage. This was used and in place during initial pond filling only.
The SNSW Pond was filled and the eastern slope of the dam was monitored for a 60 day period. During this period, the area between the SNSW Pond Dam and the downstream cofferdam was kept unwatered in order to accurately measure seepage.
Section 2.5.6.10 presents data obtained from the SNSW Pond Dam instrumentation.
2.5.6.8.2 Inspection Periodic inspections of the SNSW Dam and discharge facilities are performed by station civil engineers. These inspections conforms to the requirements of Regulatory Guide 1.127, except as discussed in Section 1.7, and include observations to detect erosion, riprap disturbances, slumping, gullying, excessive seepage, or any abnormal hazardous condition. The inspections are conducted under the direction of qualified engineers experienced in the investigation, design, construction, and operation of water control facilities.
The inspection team usually meets prior to the on-site inspection to review design criteria, instrumentation data, and previous inspection reports. A detailed inspection report is prepared after each inspection to document results of the inspection and to recommend any remedial measures which may be required.
As a result of the fourth inspection, measures were taken to 1) remove localized vegetation from the stone slope protection, 2) repair localized erosion areas around the SNSW Pond, and 3) remove debris (limbs, leaves, stones) from around the intake for the SNSW Outlet Works. The localized erosion represents three small areas described as follows:
Area No. 1: This area is located on the pond bank opposite the short-leg SNSW Discharge Structure and is the site of the excavation required to install the long-leg SNSW Discharge Pipe.
Construction in this area is complete, the area has been grassed, and the area is not currently eroding.
Area No. 2: This area is located below the concrete batch plant. The soils in this area are very erosion resistant as a result of near surface boulders and gravel. Although erosion at this site was extremely insignificant, the area was grassed as a result of recommendations in the fourth inspection report.
Area No. 3: This area is located below the landfill site. It is sufficiently far from the pond and buffered from the pond with trees and vegetation to make erosion of no consequence. The area was grassed as recommended in the fourth inspection report.
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 76 (09 OCT 2016)
The three areas described above represents a combined acreage of 0.3 acres. This is less than 0.1 percent of the SNSW Pond drainage area.
There has not been any abnormal hazardous conditions observed during any of the inspections.
Copies of the inspection reports are kept at the site for reference purposes and review.
Inspections will be made throughout the life of the plant and will be in accordance with Regulatory Guide 1.127 except as noted in Section 1.7.
HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.5.6.9 Construction Notes No significant construction problems occurred during construction of the SNSW Pond Dam. As shown on Figure 2-161, continuous progress was made in placing fill during construction seasons.
2.5.6.10 Operational Notes No significant problems have occurred since completion of the SNSW Pond Dam. Embankment performance is satisfactory as evidenced by the following data obtained from the instrumentation defined in Section 2.5.6.8.1:
- 1. At the time the downstream cofferdam was breached on about the first of November 1978, Piezometers Nos. 1, 2, 3, 5, 6, and 7 were dry.
- 2. At the time the downstream cofferdam was breached on about the first of November 1978, the water levels in Piezometer Nos. 4 and 8 are El. 539.0 and El. 545.8, respectively.
The piezometers were read thereafter with the results shown on Figure 2-201. Shown on Figure 2-201 (Sheet 5), are the levels of Lake Wylie. During the time shown on Figure 2-201, the average level of the SNSW Pond was about at normal pond elevation 571.0, with minor fluctuation. For each piezometer location, the estimated phreatic elevation for static conditions (wherein the SNSW Pond is at normal pond elevation 571 and Lake Wylie is at full pond 569.4) is shown as a dotted line. Comparison of the estimated elevations with the measured data on Figure 2-201 indicates the estimated phreatic elevations were appropriate.
- 3. The dam was completed in late 1977 as shown on Figure 2-161. Post construction settlement monitoring began on March 3, 1978, and is continuing. The interpreted results of the monitoring program through September 4, 1981 are listed below for selected dates:
Settlement, Ft., Since March 1978 Station Monument Nominal Depth Fill, Ft.
Sept.
1978 Sept.
1979 Sept.
1980 Sept.
1981 0+50 SM-1 35
.03
.07
.07
.07 1+00 SM-2 55
.045
.08
.08
.08 1+50 SM-3 65
.04
.09
.09
.09 2+00 SM-4 75
.035
.075
.075
.075 2+50 SM-5 75
.035
.07
.07
.07 3+50 SM-6 75
.035
.075
.075
.075
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 77 Settlement, Ft., Since March 1978 4+50 SM-7 70
.025
.065
.065
.065 6+50 SM-8 50
.015
.04
.04
.04 8+50 SM-9 45
.015
.04
.04
.04 10+50 SM-10 35
.01
.025
.025
.025 13+50 SM-11 15
.01
.03
.03
.03 16+50 SM-12 5
.01
.03
.03
.03 Essentially no settlement has been measured since mid-1979, less than two years after the end of construction. This is compatible with Duke's experience with rate of settlement on similar embankment dams of similar soils.
The settlement monitoring shows a crest settlement (obtained by subtracting measured elevations from the elevation on March 1978) of only 0.07 to 0.09 ft in the areas of maximum embankment height above prepared foundation. The measured elevations of the settlement monuments are plotted versus time on Figure 2-202 (12 sheets). The measured settlements are compatible with those that can be estimated for the same time period using laboratory consolidation tests and considering the embankment construction sequence. For stations on the south abutment, the following tabulation provides measured total and differential settlements (computed from the measured point values) since March 1978; also tabulated are calculated total and differential settlements for the same stations. The tabulation shows that measured differential settlements are less than the calculated values for the same time period. However, neither the measured nor the calculated differential settlements are of sufficient magnitude to be of engineering concern for the integrity or performance of the dam.
Monument Station Measured Settlement 3/78-9/79 Measured Differential 3/78-9/79 Calculated Settlement 3/78-9/79 Calculated Differential Settlement 3/78-9/79 Ft Ft Ft Ft SM-1 0+50
.07
.035
.010
.060 SM-2 1+00
.08
.095
.010
.022 SM-3 1+50
.09
.117
.015
.021 SM-4 2+00
.075
.138 Figure 2-203 shows the crest elevations at stations along the length of the SNSW Pond Dam. In May 1982, the crest elevation was well above the design crest (elevation 595) at all stations of the dam (including the areas of maximum embankment height) except for locations near the northern end of the dam, beyond about station 14+00. In this latter area,
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 78 (09 OCT 2016) the crest is not yet constructed to the design crest elevation. This area will be constructed to the design crest elevation.
- 4. Total measured flow, groundwater plus seepage, for the 60 day monitoring period ranges from 68 to 76 gpm. A base flow of 48 gpm was measured before pond filling began, therefore, seepage flow is estimated to be 20 to 28 gpm. Seepage of 20 to 28 gpm was obtained by subtracting the base flow recorded on May 24, 1978 (48 gpm) before pond filling from the range of flow measured after pond filling commenced (68 gpm on 9/15/78 and 76 gpm on 8/29/78). The flow measurements were obtained by recording the time in seconds required to fill a five gallon bucket. The measurements represent overflow from basins filled by water routed through ditches downstream and parallel to the toe of the dam.
As stated in Section 2.5.6.10 of the FSAR, the head difference during the seepage monitoring period represents a situation that will not exist as an operating condition. Flow measurements made before and during the monitoring period are presented below.
Date Flow, gpm SNSW Pond Elevation, ft.
Date Flow, gpm 05-24-78 48 540.0 09-12-78 70 05-31-78 48 540.0 09-13-78 72 08-03-78 64 557.5 09-14-78 70 08-16-78 70 565.3 09-15-78 68 08-17-78 70 566.8 09-18-78 71 08-18-78 70 566.3 09-20-78 70 08-21-78 69 568.0 09-21-78 70 08-22-78 69 568.5 09-26-78 72 08-23-78 69 569.0 09-27-78 70 08-24-78 69 569.5 09-28-78 70 08-25-78 69 570.0 09-29-78 70 08-28-78 74 571.0(1) 10-03-78 69 08-29-78 76 10-05-78 71 08-30-78 73 10-06-78 69 08-31-78 71 10-10-78 69 09-08-78 71 10-12-78 69 09-11-78 70 10-13-78 69 Note:
- 1. Full Pond During the seepage monitoring period, the area between the cofferdam and the toe of the SNSW Pond Dam was kept unwatered to elevation 527 by pumping, and the average level of the SNSW Pond is at about normal pond elevation 571 with minor fluctuations. Thus, during the seepage monitoring period, the SNSW Pond Dam embankment is subjected to a head
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 79 difference of 44 feet, a situation that does not exist as an operating condition; even should Lake Wylie Dam fail, the Lake Wylie pool level drops to elevation 550 and the SNSW Pond Dam is subject to a differential head of only 24 feet as explained in Section 2.5.6.6.2.
As explained in Section 2.5.6.6.2, because of the relatively small differential heads across the dam, both during normal operations and during a postulated loss of Lake Wylie, a seepage loss analysis is deemed unnecessary and no flownet is constructed for this purpose.
2.5.7 References
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- 37. Zietz, I., Popenoe, P., and Higgins, B., Regional Structure of the Southeastern United States as Interpreted From New Aeromagnetic Maps of Part of the Coastal Plain of North Carolina, South Carolina, Georgia and Alabama, Abs., Ann. Mtg., Geol. Soc. Amer., v. 8, No. 2, 1976, pp. 307.
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Geological Survey Prof. Paper 213, 1948, 156 pp.
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- 51. Hooker, V. E. and Johnson, C. F., Near-Surface Horizontal Stresses Including the Effects of Rock Anisotropy, U. S. Bureau of Mines, R. I. 7224, 1969, 29 pp.
- 52. Duke Power Company, Report of Bad Creek Pilot Tunnel, 1978.
- 53. Turner, F. J., Metamorphic Petrology: Mineralogical and Field Aspects, McGraw-Hill Book Co., 1968.
- 54. Thompson, A. B., "Pco2 In Low-Grade Metaphorphism: Zeolite, Carbonate, Clay Mineral, Prehnite Relations In the System, CaO-A12O3-SiO2-CO2-H2O", Contr. Mineral and Petrol. 33, pp. 145-161 (1971).
UFSAR Chapter 2 Catawba Nuclear Station 2.5 - 82 (09 OCT 2016)
- 55. Sowers, G. F., "Soil Problems in the Southern Piedmont Region", Proceedings, ASCE Soil Mechanics and Foundations Division 80, 17 pp. (March, 1954).
- 56. Nuttli, O. W., State-of-the-Art for Assessing Earthquake Hazards in the United States, Report 1, Design Earthquakes for the Central United States, U. S. Army Waterways Experiment Station, Soils and Pavement Laboratory, Vicksburg, Miss., 1973.
- 57. U. S. Coast and Geodetic Survey, Earthquake History of the United States, U. S.
Department of Commerce, Publication 41-1, 1970.
- 58. Law Engineering Testing Company, Report of Evaluation of Intensity of Giles County Virginia Earthquake of May 31, 1897, Report Prepared in Conjunction with Burns and Roe, November, 1975.
- 59. Law Engineering Testing Company, Earthquake Data Files.
- 60. Bonilla, M. G., Historic Surface Faulting in Continental United States and Adjacent Parts of Mexico, U. S. Geological Survey and U. S. Atomic Energy Commission, TID-24124, 1967.
- 61. Bollinger, G. A., "Seismicity of the Central Appalachian States of Virginia, West Virginia and Maryland - 1758-1968", Bulletin of Seismological Society of America 59, No. 5, pp. 2103-2111 (1969).
- 62. McGuire, R. K., "Effects of Uncertainty in Seismicity on Estimates of Seismic Hazard for the East Coast of the United States", Bulletin of Seismological Society of America 67, No. 3, pp.
827-848, (1977).
- 63. Gutenburg, B. and Richter, C. F., "Earthquake Magnitude, Intensity, Energy and Acceleration", Bulletin of Seismological Society of America 46, pp. 105-143 (1956).
- 64. Gutenburg, B. and Richter, C. F., "Earthquake Magnitude, Intensity, Energy and Acceleration", Bulletin of Seismological Society of America 32, No. 3, pp. 163-191 (1974).
- 65. Hershberger, J., "A Comparison of Earthquake Accelerations with Intensity Ratings",
Bulletin of Seismological Society of America 65 (1956).
- 66. Trifunac, M. D. and Brady, A. G., "On the Correlation of Seismic Intensity Scales with Peaks of Recorded Strong Ground Motion", Bulletin of Seismological Society of America 65, No. 1, pp.
139-162 (1975).
- 67. Coulter, H. W., Waldron, H. H., and Devine, J. F., Seismic and Geologic Siting Considerations for Nuclear Facilities, Fifth World Conference on Earthquake Engineering, Rome, 1973.
- 68. Newmark, N. M., Design Criteria for Nuclear Reactors Subject to Earthquake Hazards, Urbana, Illinois, May 25, 1967.
- 69. U. S. Department of the Interior, Bureau of Reclamation, Earth Manual, U. S. Government Printing Office, Washington, D. C., Designation E-18.
- 70. Cedergren, H. R., Seepage, Drainage and Flow Nets, John Wiley & Sons, Inc., New York, N. Y., 1967, pp. 87-89.
- 71. Bishop, A. W., and Henkel, D. J., The Measurement of Soil Properties in the Triaxial Test, Edward Arnold Publishers, Ltd., 1964.
- 72. Lambe, T. W., Soil Testing for Engineers, John Wiley and Sons, 1951.
- 73. U. S. Army Corps of Engineers, Laboratory Soils Testing, 1970.
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- 74. Taylor, P., and Hughes, J., Dynamic Properties of Foundation Subsoils as Determined from Laboratory Tests, Proceedings, 3rd World Conference on Earthquake Engineering, Vol. 1, 1965, pp. 196-211.
- 75. Taylor, R. M., and Bacchus, D. R., Dynamic Cyclic Strain Test on a Clay, Proceedings, 7th International Conference on Soil Mechanics and Foundation Engineering, Vol. 1, 1969.
- 76. Hardin, B. O., and Drnevich, V.P., "Shear Modulus and Damping in Soils", Vols. 1 and 2, Dept. of Civil Engineering, University of Kentucky, June and July, 1970.
- 77. Hardin, B. O., "Suggested Methods of Test for Shear Modulus and Damping of Soils by the Resonant Column", ASTM Special Technical Publication No. 479, 1970.
- 78. Timoshenko, S. P., and Goodier, J. N., Theory of Elasticity, Second Edition, McGraw-Hill Book Comapny, Inc., New York, N. Y., 1951, pp. 366-372.
- 79. Woodward, R. J., Gardner, W. S., and Greer, D. M., Drilled Pier Foundations, McGraw-Hill, Inc., 1972, pp. 51.
- 80. Terzaghi, K., Theoretical Soil Mechanics, John Wiley & Sons, Inc., New York, N. Y., 1943.
- 81. Meyerhof, G. G., "The Influence of Roughness of Base and Ground Water on the Ultimate Bearing Capacity of Foundations", Geotechnique 5, September, 1955, p. 227.
- 82. Meyerhof, G. G., "The Bearing Capacity of Footings Under Eccentric and Inclined Loads",
Proceedings of the Third International Conference on Soil Mechanics and Foundation Engineering 1, Zurich, 1953.
- 83. Westergaard, H. M., "A Problem of Elasticity Suggested by a Problem of Soil Material Reinforced by Numerous Strong Horizontal Sheets", Contributions to Mechanics of Solids, The MacMillan Company, New York, N. Y., 1938.
- 84. Sowers, G. F., "Engineering Properties of Residual Soils Derived from Igneous and Metamorphic Rocks" Proceedings, Second PanAmerican Conference on Soil Mechanics and Foundation Engineering, Brazil, 1963.
- 85. U. S. Department of Defense, Department of the Army, Coastal Research Center, Shore Protection Manual, 2nd Edition, U. S. Government Printing Office, Washington, 1975.
- 86. Sherard, J. L., Woodward, Gizienski, and Clevenger, (1963), Earth and Earth-Rock Dams, John Wiley & Sons, New York.
- 87. Sowers, G. F., Earth and Rockfill Dam Engineering, Asia Publishing House, Bombay, 1961.
- 88. Kulhway, F. H., Duncan, J. M., and Seed, H. B., Finite Element Analyses of Stresses and Movements in Embankments During Construction, Report No. TE-69-4 to U. S. Army Waterways Experiment Station, University of California, Berkeley, November, 1969.
- 89. Seed, H. B., and Lee K. L., "Liquefaction of Saturated Sands During Cyclic Loading",
Journal of the Soil Mechanics and Foundations Division 92, No. SM6, ASCE, pp. 105-134 (November, 1966).
- 90. Lee, K. L. and Seed, H. B., "Cyclic Stresses Causing Liquefaction of Sand", Journal of the Soil Mechanics and Foundations Division 93, No. SM1, ASCE (January, 1967).
- 91. Seed, H. B., Lee, K. L., and Idriss, I. M., "Analysis of Sheffield Dam Failure", Journal of the Soil Mechanics and Foundations Division 95, No. SM6, ASCE (November, 1969).
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- 92. Seed, H. B., and Peacock, W. H., Applicability of Laboratory Test Procedures for Measuring Soil Liquefaction Characteristics Under Cyclic Loading, Earthquake Engineering Research Center Report No. EERC 70-8, University of California, Berkeley, November, 1970.
- 93. Gordon, B. B., Dayton, D. J., and Sadigh, K., Seismic Stability of Upper San Leandro Dam, Meeting Preprint 2025, ASCE National Structural Engineering Meeting, San Francisco, April 9-13, 1973.
- 94. Lee, K. L., and Chan, K., Number of Equivalent Cycles in Strong Motion Earthquakes, University of California at Los Angeles, 1972.
- 95. Idriss, I. M., et al, Computer Programs for Evaluating Seismic Response of Soil Structures By Variable Damping Finite Element Procedures, Report No. EERC 73-16, University of California, Berkeley, July, 1973.
- 96. Butler, J. R. and Dunn, D. E., Geology of the Sauratown Mountains Anticlinorium and Vicinity, North Carolina, Southeastern Geology Special Pub. No. 1, 1968, pp 19-47.
- 97. Conley, J. F., Geology of the Albemarle Quadrangle, North Carolina, North Carolina Dept.
of Cons. and Dev., Bull. 75, 1962, 26 pp.
- 98. Dalrymple, G. B. and Lanphere, M. A., Potassium-Argon Dating, W. H. Freeman and Co.,
San Francisco, 1969, pp 225.
- 99. Dames & Moore, Supplemental Geologic Investigation Report, Virgil C. Summer Nuclear Station - Unit 1, South Carolina Electric & Gas Company, 1974.
100.
Espenshade, G. H. and Potter, D. B., Kyanite, Sillimanite and Andalusite Deposits of the Southeastern States, U. S. Geol. Survey Prof. Paper 336, 1960, pp 1-5, 64-85, 94-95, 107-108.
101.
Fisher, G. W., et al., Studies of Appalachian Geology: Central and Southern, Interscience Publishers, 1970, Tectonic Map.
102.
Hunt, C. B., Physiography of the United States, W. H. Freeman and Co., San Francisco, 1967, pp 137-204, Figure 1.1.
103.
Keith, A. and Sterrett, D. B., Description of the Gaffney and Kings Mountain Quadrangles, U. S. Geol. Survey Geologic Atlas of the U. S., Folio 222, 1931.
104.
Law Engineering Testing Company, Geology of Cherokee Nuclear Station, Duke Power Company, 1974.
105.
Stromquist, A. A., Choquette, P. W. and Sundelius, H. W., Geologic Map of the Denton Quadrangle, Central North Carolina, U. S. G. S. Map GQ-872, 1971.
106.
Stuckey, J. L. and Conrad, S. G., Explanatory Text for Geologic Map of North Carolina, North Carolina Dept. of Cons. and Dev., Bull. 71, 1958, 51 pp.
Note: References 96 through 106 are not cited in the text.
107.
Duke Power Company, Final Geologic Report on Brecciated Zones. March 1, 1976.
108.
Seed, H. B. and Idriss, I. M., Soil Moduli and Damping Factors for Dynamic Response Analysis, Earthquake Engineering Research Center, Report No. EERC-70-10, December 1970.
109.
Seed, H. B., Idriss, I. M., Makdisi, F., and Banerjee, N., Representation of Irregular Stress Time Histories by Equivalent Uniform Stress Series in Liquefaction Analyses, Report No.
EERC-75-29, College of Engineering, University of California, Berkeley, October 1975.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.5 - 85 110.
Wroth, C. P., In Situ Measurement of Initial Stress and Deformation Characteristics, Session IV, ASCE Geotechnical Engineering Division Special Conference on In Situ Measurement of Soil Properties, Raleigh, North Carolina, June, pp. 1-14.
111.
Newmark, N. M. and Rosenblueth, E., Fundamentals of Earthquake Engineering, Civil Engineering and Engineering Mechanics Series, Prentice-Hall, Inc. Englewood Cliffs, New Jersey, 1971.
112.
Schnabel, P. B.; Lysmer, J.; and Seed, H. B., "SHAKE, A Computer Program for Earthquake Response Analysis of Horizontally Layered Sites" Report No. EERC-72-12, December 1972, College of Engineering, Univ. of California, Berkeley, California.
113.
Hardin, B. O., "Shear Modulus of Gravels," UKY TR74-73-CE19, Soil Mechanics Series No. 16, University of Kentucky, College of Engineering Dept. of Civil Engineering, September, 1973.
114.
Newmark, N. M., Blume, J. A. and Kapur, K. K. (1973), "Seismic Design Spectra for Nuclear Power Plant" Proceedings of Power Division of ASCE, Vol. 99, No. PO2, November 1973, pp. 287-303.
115.
Anderson, D. G., Espana, C. and McLamore, V. R., (1978) "Estimating In-Situ Shear Moduli at Competent Sites," Proceedings of the ASCE Geotechnical Engineering Division Specialty Conference on Earthquake Engineering and Soil Dynamics, Vol. 1, page 181-197, Pasadena, CA, June 19-21, 1978.
116.
Seed, H. B. (1979), "Soil Liquefaction and Cyclic Mobility Evaluation for Level Ground During Earthquakes," Journal of the Geotechnical Engineering Division 105, No. GT2, ASCE, February 1979.
117.
Serf, N. Seed, H. B. Makdisi, F. I. and Chang, C. Y., "Earthquake Induced Deformations of Earth Dams" Report No. EERC 76-4, College of Engineering, University of California, Berkeley, California, September, 1976.
118.
Newmark, N. M., "Effects of Earthquakes on Dams and Embankments," Geotechnique, Volume 15, No. 2, January 1965.
119.
Franklin, Arley, G., and Chang, Frank K., "Permanent Displacements of Earth Embankments by Newmark Sliding Block Analysis," WES Miscellaneous Paper S-71-77, November 1977.
120.
Seed, H. B., and Martin G. R., "The Seismic Coefficient in Earth Dam Design,"
Proceedings of ASCE, SM & FD Journal No. SM3, May 1966.
121.
Makdisi, F. I., and Seed, H. B., "A Simplified Procedure for Estimating Earthquake -
Induced Deformations In Dams and Embankments," Report No. UCB/EERC 77-19, College of Engineering, University of California, Berkeley, August, 1977.
122.
Horton, J. W., Jr., "Shear Zone between the Inner Piedmont and Kings Mountain Belts in the Carolinas," Geology 10, pp. 28-33 (1981a).
123.
Horton, J. W., Jr., Geologic Map of the Kings Mountain Belt between Gaffney, South Carolina and Lincolnton, North Carolina, in Horton, J. W., Jr., et al., Geological Investigation of the Kings Mountain Belt and Adjacent Areas in the Carolinas, Field Trip Guidebook, Carolina Geological Society, October 24-25, 1981b, pp. 6-18.
124.
Kish, S. A., Geochronology of Plutonic Activity in the Inner Piedmont and Kings Mountain Belt of North Carolina, in Burt, E. R., Field Guides for Geologic Society of America, Southeast Section Meeting, Winston-Salem, North Carolina, 1977, pp. 144-149.
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Wilson, F. A., Geologic Interpretation of Geophysical Data from the "Mecklenburg-Weddington" Gabbro Complex, Southern Mecklenburg County, North Carolina, in Horton, J. W.,
Jr., et at., Geologic Investigations of the Kings Mountain Belt and Adjacent Areas in the Carolinas, Field Trip Guidebook, Carolina Geological Society, October 24-25, 1981, pp. 28-38.
126.
Duke Power Company, CNS 1108.00-00-0002, Specification for the Response Spectra and Seismic Displacements for Category 1 Structures.
THIS IS THE LAST PAGE OF THE TEXT SECTION 2.5.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.6 - 1 2.6 Review and Evaluation of Recent Geologic Information With Specific Reference to the Charleston Epicentral Area HISTORICAL INFORMATION IN ITALICS BELOW NOT REQUIRED TO BE REVISED 2.6.1 Glossary1 Allochthon (allochthonous) A mass of rocks which has been moved from its original site of origin by tectonic forces, as in a thrust sheet or nappe. Many allochthonous rocks have been moved so far from their original sites that they differ greatly in facies and structure from those on which they now lie.
Aulacogen A fault-bounded intracratonic trough or graben. A sediment-filled trough of one arm of a triple junction that ceased to develop.
Syn: Failed-arm trough (Burke and Dewey, 1973).
Autochthon (autochthonous) A body of rocks that remains at its site of origin, where it is rooted to its basement. Although not moved from their original site, autochthonous rocks may be midly to considerably deformed.
Decollement Detachment structure of strata due to deformation, resulting in independent styles of deformation in the rocks above and below. It is associated with folding and with overthrusting, but is merely a descriptive term.
Detachment decollement Orogeny Literally, the process of formation of mountains. By present geologic usage orogeny is the process by which structures within mountain areas were formed, including thrusting, folding, and faulting in the outer and higher layers, and plastic folding, metamorphism, and plutonism in the inner and deeper layers.
Orogenic events referenced herein and their approximate ages are as follows:
Taconic 500 to 450 mybp Acadian 400 to 350 mybp Alleghenian 300 to 250 mybp Parautochthonous Said of a rock unit that is intermediate in tectonic character between autochthonous and allochthonous.
Root Zone That area in which a low-angle thrust fault becomes steeper and descends into the crust.
Sole Fault (sole thrust) A low-angle thrust fault forming the sole of the thrust nappe; also, the basal main fault of an imbrication.
Syn.: decollement fault; detachment fault.
Subduction The process of one crustal block descending beneath another, by folding or faulting or both.
1 Except as noted, all definitions are excerpted from American Geological Institute, Glossary of Geology, Gary, M. et al., Eds., Washington, D. C., 1974, 805 pp.
UFSAR Chapter 2 Catawba Nuclear Station 2.6 - 2 (09 OCT 2016) 2.6.2 Review and Evaluation of Recent Geologic Information With Specific Reference to the Charleston Epicentral Area Following the first submission of the FSAR for Catawba Nuclear Station, additional research has taken place and many hypotheses have been advanced concerning the structure and tectonic history of the southern Appalachians, in general, and the Charleston Epicentral Area, in particular. The purpose of the following discussions is to summarize that information and to evaluate suggestions that the Charleston earthquake could occur beyond the limits of the Charleston Epicentral Area. This material was formerly Appendix 2A to the Catawba FSAR.
2.6.3 Tectonic Evolution of the Southern Appalachians The southern Appalachians developed through the late Precambrian and Paleozoic Eras (800-250 million years before present, mybp) by tectonic forces and plate movements. In recent years, several tectonic models have been proposed for development of the southern Appalachians (Hatcher, 1972, 1978; Rankin, 1975, 1976; Cook et al., 1979, 1980, 1981; Long, 1979; Cook and Oliver, 1980; Hatcher and Zietz, 1980). These individual models are not discussed in detail here, but a synthesis is presented.
The southern Appalachians evolved through a series of continental collisions and rifting episodes involving the eastern edge of the ancestral North American continent, fragments of the North American continent (Inner Piedmont-Blue Ridge block), an island arc (Carolina slate belt block), and the western edge of proto-Africa (Hatcher, 1978; Cook et al., 1979, 1980). In the late Precambrian, a large megacontinent was rifted apart to form two large continents (proto-North American and proto-Africa; Rankin, 1975). The proto-Atlantic (Iapetus Ocean) was formed between the two continents (Wilson, 1966; Rankin, 1975). Fragments of continental crust representing the Inner Piedmont-Blue Ridge block and the Carolina slate belt block were removed from proto-North America by this rifting (Rankin, 1975; Hatcher, 1978; Cook et al.,
1979, 1980; Harris and Bayer, 1979; Long, 1979). The intervening basins were filled with sediments and volcanics. The volcanism was related to subduction of oceanic crust due to the closing of the proto-Atlantic Ocean and the impending collision of the proto-North American continental edge with the Inner Piedmont-Blue Ridge block during the early Paleozoic (500 to 450 mybp) Taconic orogency (Rankin, 1976; Hatcher, 1978; Cook et al., 1979, 1980). The collision resulted in the first episode of mountain building and extensive deformation, plutonism and metamorphism in the Inner Piedmont-Blue Ridge block (Butler, 1972; Rankin, 1976; Hatcher, 1978; Cook et al., 1979, 1980). This mountain belt was part of a large allochthonous sheet thrust onto the proto-North American continental margin (Cook et al., 1979, 1980; Harris and Bayer, 1979).
The next episode of mountain building occurred during the middle Paleozoic (400-350 mybp)
Acadian orogeny due to the closing of an ocean basin between the Inner Piedmont-Blue Ridge fragment and the island arc represented by the Carolina slate belt (Hatcher, 1978; Cook et al.,
1979). During this orogeny the Inner Piedmont-Blue Ridge fragment was thrust farther westward over the old continental margin. Additional deformation, plutonism and metamorphism occurred during the Acadian orogeny. Note, however, that some workers find that evidence for the Acadian orogeny in the southern Appalachians is weak compared to that available for the northern Appalachian region (Rodgers, 1967; Hatcher, 1978; Bobyarchick, 1981). The final continental collision was between proro-North America and proto-Africa as the proto-Atlantic (Iapetus Ocean) was finally closed (Dewey and Bird, 1970; Hatcher, 1972. 1978; Rankin, 1975, 1976; Cook et al., 1979). This collision is the late Paleozoic Alleghenian orogeny (sometimes referred to as the Appalachian event) and occurred about 300-250 mybp. Plutonism occurred in the eastern Piedmont (Kiokee and Raleigh belts) during this time and extensive overthrusting
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.6 - 3 deformed sedimentary Paleozoic rocks in the Valley and Ridge Province (Harris and Milici, 1977; Snoke et al., 1980; Bobyarchick, 1981).
Most of the models of the southern Appalachians developed before 1977 were based on surface geology and only recently have data been available on crustal structure. A general knowledge of the deep crust has been obtained from geophysical measurements such as gravity, magnetics and seismic profiling (Examples: Black et al., 1975; Clark et al., 1978; Zietz, 1978; Cook et al., 1979, 1980, 1981; Harris and Bayer, 1979; Long, 1979; Hatcher and Zietz, 1980; Cook and 0liver, 1981; Smith et al., 1981). The COCORP (Consortium for Continental Reflection Profiling) seismic reflection profiles and others across the southern Appalachians are the basis of new ideas on large-scale Appalachian structure (Clark et al., 1978; Cook et al.,
1979, 1980, 1981; Harris and Bayer, 1979; Cook and 0liver, 1981).
The collisional events described previously between continents and fragments of continents resulted in the formation of an allochthonous (detached) over-thrust sheet. The COCORP profile in eastern Tennessee, southwestern North Carolina and eastern Georgia shows reflectors ranging from a depth of 6 km in the Valley and Ridge Province to 15 km near the Elberton Granite, a distance of approximately 250 km. Cook et al., (1979, 1980) interpreted these reflectors to be elements of the sole thrust and relatively undeformed, horizontal, lower Paleozoic sedimentary rocks. The Brevard zone is interpreted by Cook et al., (1979) as a splay off the main sole thrust. Hatcher and Zietz (1980) believe that "the Brevard zone was brought up during a brittle (cataclastic) event carrying with it (and) exposing a portion of the older ductile (mylonitic) sole and root zone." The Brevard zone was brought to the surface during Alleghenian deformation (Hatcher and Zietz, 1980).
Based on their examination of the COCORP data, Cook et al., (1979) believe the detachment continues to the east and that the Elberton Granite has been transported. However, based on interpretation of paleomagnetic data and geologic relationships, Ellwood et al., (1980) believe the Elberton Granite was emplaced after major thrusting. The age of 350+/-11 mybp for the granite, therefore, would be a minimum age for thrusting (Ellwood et al., 1980).
Cook and 0liver (1981) indicate that east of the Elberton Granite the reflectors thicken and dip east to a location within the Charlotte belt. They interpret these reflections as Precambrian-early Paleozoic ocean basin assemblages deposited on the autochthonous proto-North American continental shelf. The sole thrust according to their interpretation continues to the east beneath the Charlotte belt and Carolina slate belt. An alternate interpretation of the dipping reflectors by Hatcher and Zietz (1980) is that the area was the root zone of the main sole thrust.
Horizontal reflectors from the Charlotte belt eastward are interpreted by Cook and Oliver (1981) as sedimentary (now metamorphosed) layers that were deposited basinward of the shelf edge.
Below these reflectors Cook and Oliver (1981) describe 10 to 15 km of thin crust interpreted as oceanic or attenuated continental crust representing the basin between the proto-North American continent and the Inner Piedmont-Blue Ridge fragment. Another interpretation of the crustal structure east of the Inner Piedmont has been made by Hatcher and Zietz (1980). They consider the Charlotte-Carolina slate belt block as parautochthonous and underlain by mafic crust. The main difference between interpretations is that Hatcher and Zietz (1980) indicate that the sole thrust of the allochthonous Inner Piedmont-Blue Ridge block is rooted beneath the Kings Mountain belt, while Cook and Oliver (1981) extend the sole thrust beneath the Charlotte-Carolina slate belt block. Still another interpretation of the Charlotte-Carolina slate belt block by Long (1979) suggests that the block is a rift zone with associated volcanic and sedimentary rocks.
Cook et al., (1981) interpreted reflectors on a southeastward extension of the COCORP traverses in the Coastal Plain of Georgia as either fault surfaces or a metamorphosed strata
UFSAR Chapter 2 Catawba Nuclear Station 2.6 - 4 (09 OCT 2016) beneath the crystalline rocks of the southern Appalachians. They believe these reflectors are consistent with a major detachment extending eastward under the Coastal Plain. They acknowledge that other interpretations with a more complex pattern of detachments or sutures are also possible.
Many points concerning the nature and continuity of a detachment surface beneath the entire Piedmont and Coastal Plain are not agreed upon (Examples: Long, 1979; Hatcher and Zietz, 198O; Moench, et al., 1989; Williams, et al., 1980; Cook et al., 1981). Examination of the actual COCORP profiles indicates that a great deal of "interpretation" has gone into constructing the models. At this time, the models can be used as a basis for working hypotheses; it is generally recognized that additional data must be obtained.
2.6.4 Tectonic Models of Charleston Seismicity The southeastern United States is distant from lithospheric plate boundaries and has no active faults. The state of stress and earthquake mechanisms in this intraplate region are poorly understood. The inability to identify potential earthquake source zones is a major problem in determining seismic hazard. In current practice, seismotectonic regions are utilized to provide a rational basis for the delineation of earthquake source zones in the southeastern United States.
Seismotectonic regions are characterized by consistency of geologic structure and historical seismicity.
The Charleston Epicentral Area is one of the seismotectonic regions discussed in the FSAR (Section 2.5.2.3.3). The lower Coastal Plain, in general, is notable for its lack of seismic activity.
However, this otherwise aseismic region was the site of the largest historic earthquake to have occurred in the Atlantic coastal states, the earthquake of August 31, 1886 at Charleston, South Carolina. The Charleston Epicentral Area seismotectonic region is defined to include the area of concentrated historic seismicity in the lower Atlantic Coastal Plain. The level of seismicity in the Charleston area continues to exceed that observed in the Atlantic Coastal Plain of the Carolinas.
Study and review of the seismicity of the Charleston area was renewed in the early 1970's and intensified in 1974 when the U. S. Geological Survey began its mutidisciplinary studies (Rankin, 1977). Preliminary results of those studies were reported in 1977 in Geological Survey Professional Paper 1028. Work has continued and another Professional Paper compilation is expected in early 1982. Many of the data and working hypotheses to be presented in the upcoming Professional Paper have been published in various journals and discussed at professional society meetings during the past several years. The bulk of evidence available to date is from remote sources (seismic, gravity, magnetics); the hypotheses discussed in following paragraphs are based on such indirect sources. No direct evidence exists for an earthquake fault in the surface or subsurface of the southern Appalachians.
A survey of the literature indicates that there is great divergence of opinion on the origin and mechanism of the Charleston earthquake. The various hypotheses that have been advanced can be categorized into one of three principal mechanisms: 1) decollement reactivation, 2) stress amplification at margins of mafic plutons, and 3) reactivation of steep basement faults.
Even among the proponents of one of the principal mechanisms, there is disagreement about aspects of geologic structure and seismic mechanisms and conflicts on interpretation of data.
The three mechanisms and their variations are discussed briefly in following paragraphs.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.6 - 5 2.6.4.1 Decollement Reactivation Examination of COCORP surveys in the southern Appalachians is convincing that the overthrust described by Cook et al., (1979) is present beneath the Valley and Ridge, Blue Ridge and western part of the Inner Piedmont. Such a model had been proposed earlier by Hatcher (1978).
The subhorizontal reflector representing the decollement is less defined east of the rooted Brevard Zone and poorly defined east of the Inner Piedmont. Cook et al., (1979) and Harris and Bayer (1979) hypothesize that the decollement may continue to the east beneath the Atlantic Coastal Plain.
Hypotheses concerning the origin of Charleston seismicity based on decollement reactivation are in two general categories of mechanisms: 1) backslip along the entire decollement, and 2) local strain release along the decollement and reverse faulting along rooted features in the sheet above the decollement.
Armbruster and Seeber (1981) and Seeber and Armbruster (1981) attribute the concentration of highest intensities of the 1886 earthquakes in the Charleston-Summerville area to erroneous reports and interpretations. They cite indications of high-intensity effects in Columbia, South Carolina, Augusta, Georgia and other locations. They also cite data from Zoback and Zoback (1980) on the orientation of maximum and minimum stress directions as evidence for stress reversals in a decoupled upper sheet. Finally, releveling data by Citron and Brown (1979) are interpreted to indicate Cenozoic movement as far west as the Blue Ridge Front that may be related to gravity-induced backslip above the decollement. An attempt at giving further weight to some of these aspects is made by comparing the proposed southern Appalachian decollement to the Himalayan detachment.
Behrendt et al., (1980, 1981) discuss two northeast-striking, high-angle reverse faults (based on seismic reflection) in the Charleston area as possibly being related to current seismicity and the 1886 earthquakes. The Cooke fault is located onshore northwest of Charleston in an area of concentrated historic seismicity and dips northwest. The Helena Banks fault is located offshore southeast of Charleston where no seismicity is known and dips southeast. Reverse offset is based on reflection data. Tentative association is made between the Cooke fault and recent seismicity, although the authors recognize conflicts in focal depths versus reflection data and ambiguities in focal mechanism versus in-situ stress measurements. The authors hypothesize that Charleston seismicity may be related to steeply dipping reverse faults by one of two mechanisms. The first attributes reverse faulting to regional compression based on stress provinces defined by Zoback and Zoback (1980, 1981). Their alternate interpretation is a variation of the first in that the steep reverse faults are considered to be rooted in an active decollement at depths consistent with those proposed by Cook, et al., (1979) and Harris and Bayer (1979). The latter variation attributes steeply-dipping reverse faulting to a second-order effect of movement on the decollement and, in that aspect, is somewhat similar to the mechanism proposed by Armbruster and Seeber (1981) and Seeber and Armbruster (1981).
Behrendt et al., (1981) recognized that the association of seismicity with steeply-dipping reverse faults in tenuous and that there are conflicts and ambiguities within and between fault-plane solutions and in-situ stress measurements. Active movement along a decollement in the Charleston area appears unlikely for the following reasons:
- 1. The existence of a decollement east of the Inner Piedmont is not certain based on study of the COCORP data. Based on their interpretation of magnetic and gravity data, Hatcher and Zietz (1980) and Long (1979) also question a regional decollement east of the Inner Piedmont.
- 2. It is unlikely that a decollement in the southeastern U. S. could have acted as a single sheet since at least Mesozoic time because of a variety of asperities that would have to be
UFSAR Chapter 2 Catawba Nuclear Station 2.6 - 6 (09 OCT 2016) overcome. A series of 300+/- million-year-old plutons in the Charlotte and Carolina slate belts would have to be truncated and transported. Mesozoic sedimentary basins, mafic dikes and aulocogens (Rankin, 1976), all related to continental rifting, would have to be truncated after the rifting. It seems unlikely that gravity backslip could overcome such asperities or that regional compression could cause renewed thrusting without a dramatic surface expression.
- 3. In the absence of excess pore pressures (which have not been demonstrated to exist at the depths of the proposed decollement), it is unlikely that the gravity force resolved parallel to the postulated decollement is sufficient to overcome the normal shear resistance due to the force of the lithostatic load (Hubbert and Rubey, 1959).
- 4. There is no surficial expression of Cenozoic or Recent offset along the Blue Ridge or Appalachian Fronts that could be related to reactivation of thrust sheets. Releveling data are inconclusive because based on the precision of the measuring techniques, apparent movements could possibly be attributed to leveling error (Citron and Brown, 1979; Lyttle et al., 1979).
2.6.4.2 Stress Amplification at Margins of Mafic Plutons Several workers have recognized an apparent spatial association between Charleston seismicity and mafic plutons. The two principal earthquake-generating mechanisms proposed from this association are: 1) high stress concentration within mafic bodies and at their contacts with country rocks, and 2) stress concentrations in the crystalline rocks surrounding serpentinized and less compentent mafic bodies.
McKeown (1975, 1978) noted the spatial correlation between mafic intrusive rocks and seismic activity and also noted the coincidence of trends of nodal planes of fault-plant solutions and the orientations of mafic dikes. McKeown (1978) presented calculations relating Young's moduli for different media, shape of a theoretical mafic body and assumed stresses. The results indicated that stresses would be concentrated in the mafic intrusions, rather than in surrounding felsic rock. He indicated those results to be consistent with the diversity of earthquake focal mechanisms at Charleston.
Long (1976) and Long and Champion (1977) discussed possible relationships between seismicity in the Charleston area and gravity highs interpreted to be mafic or ultramafic bodies.
Long's scenario is similar to McKeown's, but Long (1976) and Long and Champion (1977) speculate that the triggering mechanism may involve variations in shape or thickness of units of contrasting rigidity and/or contemporary crustal flexure related to plate movement.
Kane (1977) also noted the spatial association of southeastern seismicity and mafic bodies and, as did Long and McKeown (discussed above), interpreted gravity highs in the Charleston area to be mafic masses. But Kane speculated that the mafic bodies with associated seismicity are serpentinized. The result, he showed, is that they are less rigid than surrounding (felsic) crystalline rock and yield by creep rather than fracture. It follows that the effects of changes in stress conditions would be concentrated in felsic crystalline rocks at the boundaries of serpentinized mafic rocks. Lack of seismicity at other gravity highs (also inferred to be mafics or ultramafics) is attributed to lack of serpentinization, an insufficiently large or changing stress field and/or inappropriate geometric relations.
Campbell (1978) supported Kane's hypothesis with analytical modeling of an hypothesized collision between crustal rocks of the Piedmont and those of the Cape Fear Arch. The result, he indicated, together with the presence of a serpentinized mafic/ultramafic, could explain the occurrence of shallow and high intensity (1886) seismicity in the Charleston area.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.6 - 7 There are several uncertainties associated with these stress amplification hypotheses. Among the uncertainties and unknowns are the following:
- 1. The models used for stress calculations by Kane (1977) and Campbell (1978) are oversimplified. The three-dimensional geometries of the (assumed) mafic/ultramafic bodies are unknown.
- 2. The precise nature of the positive gravity anomalies is not known, although the best interpretation based on current knowledge is that they are mafic or ultramafic intrustions.
- 3. The extent of serpentinization, if any, is not known. Several attempts have been made, but as yet borings have not penetrated the gravity anomalies. Thus, their composition and state of alteration are not known.
- 4. The state of stress in the vicinity of the gravity anomalies (or its cause) is not known.
- 5. There are other regions within eastern North America that contain similar gravity anomalies, but are aseismic.
- 6. The precise lithology of the rocks surrounding the gravity anomalies is not known.
Another hypothesis related to this last point was discussed by Popenoe and Zietz (1977). They speculated on the basis of magentic data that basement rocks beneath the Coastal Plain were not an extension of Piedmont metamorphics, but layered sedimentary and volcanic rocks, perhaps of Triassic age.
In some ways, this hypothesis is supported by recent COCORP data (Cook, et al., 1981) and partly supported by results of test drilling to depth of about 1.5 km (Gottfried et al., 1977; Rankin, 1977; Gohn et al., 1978). However, Cook et al., (1981) caution that "the increased complexity of reflections... calls for caution with regard to any simple unambiguous interpretation."
2.6.4.3 Reactivation of Steep Basement Faults A number of earth scientists have cited evidence for post-Cretaceous offset along steeply-dippinq faults in the Coastal Plain, in general, and the Charleston area, in particular; those occurrences are not recounted here. However several authors believe that such faults are basement features and relate continuing dip-slip movement along them to Charleston seismicity and the 1886 earthquakes. This proposed mechanism conflicts with the idea of a reactivated master decollement. In the decollement hypothesis, steep faults are listric into the nearly horizontal detachment and do not offset the detachment. Two principal variations of the steep-basement-fault hypothesis exist in the literature: 1) continuing release of strain energy along a northwest-trending zone of weakness, specifically a failed-arm trough (aulacogen) inherited from evolution of the Appalachian orogen, and 2) slip along fault-bounded basement blocks or along border faults of buried Triassic basins.
One of the elements of continental drift as described by Wilson (1966) is that, in the opening and closing of ocean basins (moving about of continental masses), old fracture zones are reused (reactivated). Rankin (1976) has proposed that "Appalachian salients and recesses are inherited from the initial breakup of a continental mass by the intersection of rift valleys radiating from triple junctions at the start of the opening of the Iapetus Ocean." He cites several examples of possible failed-arm troughs within the Appalachian Mountain System.
Rankin (1978) and Rankin et al., (1978) described an early Mesozoic rift valley oriented parallel to Appalachian structure and passing through the Charleston meizoseismal area. It coincidentally is bordered to the north-west by positive gravity anomalies (assumed mafic
UFSAR Chapter 2 Catawba Nuclear Station 2.6 - 8 (09 OCT 2016) intrustions). Fletcher et al., (1978) and Rankin (1978) also noted a possible continental extension of the northwest-trending Blake Spur Fracture Zone. These relationships are cited by Fletcher et al., (1978) and Rankin (1978) for structure-related seismicity along the South Carolina-Georgia seismic zone of Boilinger (1973).
A mechanism similar to that above was presented by Nishenko and Sykes (1979). They described the Georgia-Florida rift as a reactivated suture that is genetically related to and intersected by the Blake Spur Fracture Zone. They suggested that the 1886 earthquakes were related to that intersection.
Wentworth and Mergner-Keefer (1981a, b) cited a variety of sources for the existence of a domain of steeply-dipping, northeast-trending faults at the Atlantic continental margin. They described a prevailing stress regime of northwest compression and gradually declining rates of reverse offset along the faults. The fractures primarily involved, they believed, are Mesozoic normal (border) faults reactivated by compression. They also recognized the possibility of reuse of older, favorably oriented discontinuities. Finally, they concluded that the 1886 Charleston earthquakes were related to reverse faulting within part of the domain and that large earthquakes could occur anywhere in the domain.
A realted mechanism was described by Sheridan (1974). His model called for youthful (contemporary) movement along the boundaries of sedimentary basins of the geosyncline that were fault-bounded troughs. Alignment of the troughs was more or less parallel to that of the continental slope. In contrast, based upon coring and in-situ stress data, Zoback et al., (1978) described a northwest-striking normal fault in a northwest-oriented stress field. Neither Sheridan (1974) nor Zoback et al., (1978) related these structures to historic seismicity.
Finally, several authors have presented geologic or geomorphic evidence for vertical crustal movements in eastern North America. Among them are Owens (1970), Winker and Howard (1977), Heller et al., (1980), Winker (1980) and Zimmerman (1980). Crustal movements have been inferred based on geodetic data by Holdahl and Morrison (1974) and Lyttle et alt, (1979).
Conflicting data were presented by Blackwelder (1980), who described a relatively stable Atlantic continental margin during the last 9,000 to 12,000 years.
The concept of renewed activity along aulacogens has been advanced in recent years by several students of eastern North American tectonics. Some aspects that have been related to Charleston seismicity were discussed above. Other examples include the Mississippi embayment which has been described as a failed arm by Burke and Dewey (1973) and related by Ervin and McGinnis (1975) to continued seismcity in that area. The phenomenon of recurrent igneous activity from Precambrian to Eocene time along the border of Virginia and West Virginia is similar in concept (Dennison and Johnson, 1971). However, the evidence for such a northwest-trending zone of weakness in the onshore Charleston area is circumstantial and strongly dependent upon the northwest orientation of the South Carolina-Georgia seismic zone of Bollinger (1973). Even within that zone, though, distribution of seismicity is diffuse and there are aseismic segments based on historic seismicity. In addition, Bollinger (1973) does not interpret the South Carolina-Georgia seismic zone to be a fault feature; the northwest orientation has no structural implications. The zone is merely a band encompassing patterns (largely clusters) of historical energy release (G.A. Bollinger, oral communication, 1981).
There is a varied body of evidence indicating that the dominant structural trend in the Charleston area is northeast. That evidence includes seismic reflection data (e.g., Behrendt et al., 1981), gravity data (e.g., Long and Champion, 1977), and magnetic data (e.g., Popenoe and Zietz, 1977). However, as is the case with the other mechanisms discussed herein, data do not exist to establish a direct relationship between any single structure or group of structures and Charleston seismicity.
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.6 - 9 2.6.5 Summary and Conclusions The latest available geologic literature has been reviewed to determine if there is a basis for altering the commonly held belief that earthquakes similar to the 1886 Charleston events could not occur outside the Charleston Epicentral Area seismotectonic region.
Geological and geophysical studies underway since the early 1970's have resulted in promulgation of three groups of hypotheses for the origin of the Charleston earthquakes. The three principal hypotheses are: 1) reactivation of a master decollement, locally or regionally, by thrusting with associated second-order steep reverse faults, or by gravity-induced backslip, 2) stress amplification within, at the margins of, or in the rock surrounding mafic or ultramafic plutons, and 3) reactivation of steep basement faults variously described as northwest-trending, failed-arm troughs (aulocogens or extensions of ocean-basin transform faults, or as northeast-trending, normal or reverse faults reusing Triassic border faults or other discontinuities.
It can be seen that a vast array of mechanisms have been proposed. Even among proponents of one of the three principal mechanisms, there are conflicts on interpretation of similar data.
Each hypothesis has various weaknesses; some are recognized by the proponents while others have been raised by advocates of one of the other hypotheses.
Based on existing geologic mapping and newly acquired COCORP data, there is little doubt in the professional community that a major thrust sheet underlies the Valley and Ridge, Blue Ridge and western Inner Piedmont. However, some have extended the decollement beneath the Charlotte belt and others have inferred its existence beneath the Coastal Plain. Charleston seismicity has been attributed by some to thrusting along segments of the decollement (by regional compression) and by others to gravity-induced backslip along the entire detachment.
There are various difficulties that make application of the decollement hypotheses to Charleston problematical. Some of these are:
- 1. Existence of the detachment in the Charleston area is controversial;
- 2. Asperities related to young plutons, Mesozoic sedimentary basins and dikes, and possible aulocogens would have to be overcome (truncated and transported);
- 3. Factors related to the normal stress produced by 10 to 15 km of overburden would have to be overcome; and
- 4. excess pore pressure required for gravity backslip are not known to exist at proposed detachment depth.
The best interpretation of gravity highs beneath the Coastal Plain based on current knowledge is that they are mafic or ultramafic plutons. Many have recognized the spatial relationship between these apparent mafic bodies and seismicity. It has been proposed that the mafic plutons are more dense and therefore more rigid than surrounding rocks and thus stresses are concentrated in the mafics or at their contacts. It has also been proposed that the mafics are serpentinized and therefore less rigid than surrounding rocks and thus stresses are concentrated in the surrounding rocks. The difficulties in applying these hypotheses to Charleston seismicity are that:
- 1. Borings have not penetrated the gravity anomalies and thus their composition and state of alteration have not been demonstrated;
- 2. The three-dimensional geometries of the assumed mafics and boundary conditions at their contacts with country rock are not known;
- 3. The local and regional state of stress are ill defined, and
UFSAR Chapter 2 Catawba Nuclear Station 2.6 - 10 (09 OCT 2016)
- 4. There are similar positive gravity anomalies in areas of eastern North America that are aseismic.
A relationship of steeply dipping basement structure to Charleston seismicity has been made on the premise that an aulacogen exists in that part of the Appalachian orogen and on the basis of its proposed northwest trend that is coincident with the South Carolina-Georgia seismic-zone.
Conversely, an association has been made between northeast-trending faults of limited extent and Charleston seismicity. Those faults are presumed to be related to ancient faults at the borders of buried Triassic basins or older discontinuities. Renewed activity could be normal or reverse depending upon the stress model adopted. Difficulties in adopting one of these mechanisms as the source of Charleston seismicity are that:
- 1. The existence of a northwest-trending zone of weakness has not been demonstrated;
- 2. Parts of the South Carolina-Georgia seismic zone are aseismic and the zone itself has no structural inference;
- 3. Apparent Cenozoic faults have diverse orientations and slip senses; and
- 4. There are conflicts between releveling data and dating of shoreline deposits in determination of recent crustal movements.
The fact remains that Charleston-type seismicity is unique to Charleston in the historic record.
New data have been gathered in recent years and additional studies will take place. The various working hypotheses that have been developed recently are what they are presented as -
working hypotheses - a basis for designing additional data-gathering programs to clarify the many unanswered questions (e.g., Wielchowsky et al., 1978). Unless and until a relationship between Charleston seismicity and some structure and mechanism has been clearly established, the concept of seismotectonic regions would appear to be the most workable basis for categorizing seismic hazard in the southern Appalachians. On that basis, as described in Section 2.5.2.3.3. there is no reason to believe that the Charleston earthquake could occur outside the Charleston Epicentral Area seismotectonic region.
2.6.6 References
- 1. Armbruster, J. G. and Seeber, L., Intraplate Seismicity in the Southeastern U. S. and the Appalachian Detachment, in Beavers, J. E., Ed., Earthquakes and Earthquake Engineering:
the Eastern United States, Vol. 1, Ann Arbor Science, 1981, pp. 375-396.
- 2. Behrendt, J. C., Hamilton, R. M. and Ackermann, H. D., "Deep Crustal Seismic Reflection Study Offshore in the Area of the Charleston, S. C., 1886 Earthquake," Abstract, Trans.,
Amer. Geophys. Union 61, p. 1040 (1980).
- 3. Behrendt, J. C. et al., "Cenozoic Faulting in the Vicinity of the Charleston, South Carolina, 1886 Earthquake," Geology 9, pp. 117-122 (1981).
- 4. Black, W., Ferguson J. and Stewart, D., Reflection Seismic Survey of a Portion of the Southern Appalachian Slate Belt Near Chapel Hill, N. C., Abstract, Ann. Mtng., Geol. Soc.
Amer., 1975, p. 470.
- 5. Blackwelder, B. W., "Late Wisconsin and Holocene Tectonic Stability of the United States Mid-Atlantic Coastal Regign," Geology 8, pp. 534-537 (1980).
- 6. Bobyarchick, A. R., "The Eastern Piedmont Fault System and Its Relationship to Alleghanian Tectonics in the Southern Appalachians," Jour. Geology 89, pp. 335-347 (1981).
Catawba Nuclear Station UFSAR Chapter 2 (09 OCT 2016) 2.6 - 11
- 7. Bollinger, G. A., "Seismicity of the Southeast United States," Bull., Seis. Soc. Amer. 63, pp.
1785-1808 (1973).
- 8. Burke, K. and Dewey, J. F., "Plume-Generated Triple Junctions: Key Indicators in Applying Plate Tectonics to 0ld Rocks," Jour. Geology 81, pp. 406-433 (1973).
- 9. Butler, J. R., "Age of Paleozoic Regional Metamorphism in the Carolinas, Georgia, and Tennessee Southern Appalachians," Amer. Jour. Sci. 272, pp. 319-333 (1972).
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