ML20058F490
| ML20058F490 | |
| Person / Time | |
|---|---|
| Site: | Seabrook  |
| Issue date: | 07/22/1982 |
| From: | Devincentis J PUBLIC SERVICE CO. OF NEW HAMPSHIRE, YANKEE ATOMIC ELECTRIC CO. |
| To: | Miraglia F Office of Nuclear Reactor Regulation |
| References | |
| SBN-296, NUDOCS 8207300522 | |
| Download: ML20058F490 (17) | |
Text
PUBLIC SERVICE SEA M S E M Engineering Office:
Companyof NewHamph 1671 Worcester Road Frominoham, Mossochusetts 01701 (617) 872 - 8100 July 22,1982 SBN-296 T.F.B.7.1.2 United States Nuclear Regulatory Commission Washington, D. C. 20555 Attention:
Mr. Frank J. Miraglia, Chief Licensing Branch No. 3 Division of Licensing
References:
(a) Construction Permit CPPR-135 and CFFR-136, Docket Nos. 50-443 and 50-444 (b)
USNRC Letter, dated January 12, 1982, " Request for Additional Information - Seabrook Station," F. J. Miraglia to W. C. Tallman (c)
PSNil Le t t er, da t ed Ma rc h 12, 1982, " Responses to 241 Series RAI's; (llydrologic and Geotechnical Engineering Branch)" J. DeVincentis to F. J. Miraglia (d)
PSNil Letter, dated June 22, 1982, " Amended Responses to 241 Series RAI's; (flydrologic and Geotechnical Engineering Branch)" J. DeVincentis to F. J. Miraglia
Subject:
AMENDED RESPONSES TO 241 SERIES RAI's; (liYDR010GIC AND GE0TECIINICAL ENGINEERING BRANCil)
Dear Sir:
Ar a result of comments by the NRC Staf f and l'nited Engineers and Constructors (UE6C), we have amended a number of text pages, tables and figures which were originally submitted in Reference (c).
Amendments to the original submittal (Reference c) have also been provided in Reference (d ).
On each text and table page the lines to which changes have been made are identified by a vertical bar and numeral 2 in the right margin.
Changes have not been identified on the figures.
Very truly yours, YANKEE ATOMIC ELECTRIC COMPANY 5
2}
{
J. DeVincebtIs
'l o f I Project Manager 8207300522 B20722 PDR ADOCK 05000443 A
SUMMARY
LIST OF AMENDED PAGES AMENDMENT 2 OF GEI'S REVISIONS TO SEABROOK FSAR 1.
Text Pages Insert 2 of 2 on page 2.5-107 Insert 3 of 3 on page 2.5-119 Insert 1 of 2 on page 2.5-121 Insert 1 of 1 on page 2 5-125 Insert 2 of 2 on page 2.5-126 (continued)
Insert 2 of 2 on page 2.5-126 (continued)
Insert 2 of 2 on page 2.5-129 Insert 2 of 2 cn page 2.5-129 (continued)
Insert 1 of 1 on page 2.5-139 (continued) 2.
Tables Table 2.5-20 3.
Figures Figure 2.5-42a Figure 2.5-42b Figure 2.5-42c Figure 2.5-42d Project 81878 Geotechnical Engineers Inc.
July 8, 1982 l
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Insert 2 of 2 on page 2.5-107 values of Poisson's ratio were used in calculating foundation heave and settlement due to excavation and structural loads, respectively.
They were not used in dynamic analyses since the rock was assumed to be l2 a rigid boundary below foundation grades as described in Subsection 3.8.5.1.
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Insert 3 of 3 on page 2.5-119 The maximum thickness of offsite borrow beneath safety-related electrical ductbanks was 18 ft.
Typical profiles of offsite borrow-beneath safety-related electrical ductbanks are shown in Sections KK, LL, and MM on Figure 2.5-42c.
The maximum thickness of offsite borrow placed beneath safety-related electrical manholes was 18 ft.
Typical profiles of offsite borrow beneath safety-related electrical manholes are shown in Figure 2.5-42c, Section MM and Figure 2.5-42d, Sections NN and PP.
The thickness of offsite borrow beneath safety-related service water pipes ranged from approximately 2 to 15 ft.
Typical profiles in which the depth of offsite borrow beneath safety-related service water pipes ranges from 2 to 5 f t are shown in Figure 2.5-42c, Sections JJ, KK, LL, and MM.
A typical profile in which the depth of offsite borrow beneath safety-related service water pipes ranges from approximately 10 to 15 ft is shown in Figure 2.5-42b,Section II.
The maximum thickness of offsite borrow placed adjacent to safety-related structures is approximately 63 ft along the west wall of the Discharge Transition Structure and the east wall of the Service Water 2
Pumphouse.
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Insert 1 of 2 on page 2.5-121 Represen?.ative field density test results for offsite borrow placed in the plant area during typical winter, spring, and summer periods are shown in Figures 2.5-44, 44a and 44b, respectively. As seen in these three figures, the offsite borrow was compacted to at least 95% of ASTM D1557-70 compaction.
Offsite borrow was also used as fill behind the revetments, as shown in Figures 2.5-54, 55, and 56.
This offsite borrow was compacted to at l2 least 90% of maximum dry density determined by ASTM D1557-70. Typical field density test results of the of fsite borrow placed to at least 90%
compaction are shown in Figure 2.5-44d.
Near the railroad tracks at the west end of Revetment A, the of fsite borrow was placed to at least 95%
compaction as shown on Figure 2.5-44c.
Insert 1 of 1 on page 2.5-122 Tunnel cuttings were placed in two safety-related areas of the site in the vicinity of the turbine building for Unit 2.
The coor-dinates of these two areas are approximately N10160 to 10220, E5290 to 5360; and N10140 to 10210, E5420 to 5550.
The greatest thickness of tunnel cuttings in safety-related areas of the site is about 15,ft l2 beneath manhole W19/20 in the second area noted above. A profile of the tunnel cuttings beneath and adjacent to manhole W19/20 is shown in Figure 2.5-42d, Section 0-0.
Tunnel cuttings were not placed against nor within a 10 ft hori-zontal distance of the walls of any seismic Category I building.
---~
In:2rt 1 of 1 on paga 2.5-125 For all seismic analyses the rock was treated as a fixed boundary, as described in Subsection 3.7(B).2.3.
Therefore, no dynamic rock pro-
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perties were required for the seismic analyses.
Four seismic Categcry I electrical manholes are founded on of fsite borrow, with a maximum thickness of 18 ft below the base of manhole W33/34.
Seismic amplification in the maximum thickness offsite borrow was analyzed using the lumped-mass and spring approach described in Subsection 3.7(B).2.4.
During design, estimated values of shear wave velocity es = 650 ft/sec and shear modulus, G = 13,890 psi were used.
Based on data in Richart et al. (1970), cs = 650 ft/sec is a l2 conservative (low) value for the offsite borrow. These values were assumed to be constant for the 18-ft-thick layer of of fsite borrow. The amplified accelerations were used for the structural analyses of walls of manholes on offsite borrow.
Subsequent to design, the shear modulus for offsite borrow beneath the manholes was backfigured from the results of the plate load test described in Subsection 2.5.4.5.d.
The Young's modulus for the cyclic portion of the plate loading, E = 24,800 psi, and the Poisson's ratio from drained triaxial tests, V = 0.3, were used to calculate a value of G = 9,550 psi.
The average degree of compaction for the of fsite borrow test fill was the same as for the offsite borrow placed during construc-tion.
Assuming that the 10-in.-diameter plate influenced a 24-to 36-in.-thick layer of soil beneath the plate, the average shear strain, y, in the soil during the unload-reload cycle was Y = 3.6 x 10-3 in./in.
to 5.4 x 10-3 in./in.
The value of shear modulus at low strain (10-6 in./in.), Gmax, was then determined using the relationship between shear modulus and shear strain for sand presented in Seed and Idriss (1970). The average octahedral stress, Um = 4,000 psf, in the zone beneath the plate was calculated using the elastic solutions for a rigid plate with an average load of 6 tsf.
Values of G for other effective max stress levels were then computed using the relation 2 max"Gmax[Um2/Um1 G
1 where G1 max and Um1 were values from the plate load test.
A plot of G V8 Om for the offsite borrow, based on the plate load
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max test data, is shown in Fig. 2.5-58.
As noted in Subsection 3.7(B).2.4, the seismic design of the manholes was checked using the shear modulus values backfigured from the plate load test, and was found to be satisfactory.
The seismic Category I electrical ductbanks which are founded on offsite borrow were analyzed using the procedures described in Sub-section 3.7(B).2.4.
The dynamic properties backfigured from the plate load test as described above were used for these analyses.
Electrical manhole W19/20 is founded on 15 f t of tunnel cuttings l2 with a few layers of offsite borrow. Analysis of the amplification for this manhole was performed as described in Subsection 3.2(B).2.4 with an average shear modulus determined from the plate load test on tunnel cuttings using the procedure described above.
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Intert 2 of 2 on paga 2.5-126 (continu:d) b.
Settlement Due to Seismic Loading The settlement resulting from the SSE loading was also estimated for the seismic Category I structures founded on offsite borrow or tun-nel cuttings, using the relationship between horizontal cyclic shear strain in the soil during the earthquake and accumulated vertical strain described in Silver and Seed (1971) and Seed and Silver (1972). The peak horizontal cyclic shear strains were determined for the thickest layers of offsite borrow and tunnel cuttings below seismic Category I struc-tures (see Subsection 2.5.4.5.c) using the one-dimensional computer program SilAKE (Schnabel et al.,
1972) with 3 to 7 soil layers below the structure. The SHAKE analyses were performed using the lower values of shear moduli from the plate load tests (see Subsection 2.5.4.7) and the shear modulus reduction curve and damping values from Seed and Idriss (1970). Using the Seed and Silver (1971) data for relative density, Dr = 80% and 10 cycles of loading, the maximum seismic settlement for seismic Category I structures is 0.1 in. for the service water pipes located in the area with 15 ft (maximum thickness) of offsite borrow beneath the pipes.
Seismic settlements of this magnitude will not affect the performance of the seismic Category I manholes, ductbanks or service water pipes during or after the SSE cvent.
1 c.
Static and Dynamic Lateral Pressures Lateral earth pressures for Category I structures surrounded by offsite borrow were computed for both static and seismic conditions using the pressure diagrams shown in Figures 2.5-52 and 2.5-53.
The static coefficients of at-rest earth pressure, K, and active earth o
pressure, Ka, are conservative values, based on the minimum friction angle of 36' measured in triaxial tests.
Static water pressures were computed using the maximum groundwater elevation at the ground surface, El +20.
For the rigid wall, a static lateral compaction pressure was included for the full height of the wall.
The dynamic lateral pressure coefficient, K,
for non-rigid h
walls was calculated using the procedures described in Seed and Whitman (1970). As discussed in Seed and Whitman, the ef fect of vertical acce-1eration on the dynamic lateral pressure for non-rigid walls is negli-gible for the case where vertical acceleration is one half the horizontal acceleration.
The magnitude and distribution of dynamic lateral pressure of rigid walls was based on discussions with Dr. H. B.
Seed (Dalal, 1975).
l2 Based on his experience Dr. Seed recommended that an approximately uni-form pressure distribution would be appropriate for rigid box-type structures founded on rock and surrounded by soil. The lateral pressure would be one half the maximum pressure from the Seed and Whitman (1970) method, increased by an empirical factor of 3 to account for the dif-ference in stiffness between the structure on rock and the surrounding soil.
Thus, the coefficient of dynamic earth pressure, K, is D
Inctrt 2 of 2 on paga 2.5-126 (continu:d)
KD= 1/2 x 3 x AKAE
= 1/2 x 3 x 3/4 x a max 1 125 a
=
max where AKAE = 3/4 amax is the dynamic component of lateral earth pressure from Seed and Whitman (1970). The uniform pressure distribu-tion is conservative, since the actual dynamic component of lateral pressure must go to zero at the base of the wall, where there is no relative dynamic motion between the structure and the surrounding soil.
Near the very top of the wall, the horizontal earthquake pressure is limited to be equal to the passive resistance of the soil. The rigid walls of all seismic Category I buildings except five manholes were founded on sound bedrock or fill concrete extending to sound bedrock.
For these walls, the bedrock accelerations were used to compute dynamic lateral pressures. For the five manholes founded on offsite borrow or tunnel cuttings, the amplified soil accelerations at the base of the manholes described in Subsection 2.5.4.7 were used for design.
The maximum lateral pressures for any seismic Category I struc-ture occur at the east wall of the service water pumphouse and the west wall of the discharge transition structure, where the thickness of the 2
offsite borrow is 63 ft.
These are rigid walls with the following lateral pressures at the base:
Static At-Rest Soil Pressure 1,970 psf Hydrostatic Pressure 3,930 psf Permanent Surcharge O
Live Load Surcharge 250 psf Dynamic Soil Pressure (SSE) 2,210 psf Using the above values, the total horizontal earth load, excluding the live load and the hydrostatic water pressure,10 calcu-lated to be 190 k/ft. The at-rest earth load alone is 62 k/ft. Hence, the total earth load during an earthquake is 3.1 times the at-rest (K
=
o j
0.5) earth pressure. The hydrostatic water and the live load surcharge effects are added to the earthquake load for design of the walls. The i
effects of compaction need not be included during an earthquake since l
the shaking dissipates compaction prestress effects.
l Tunnel cuttings were not placed against nor within 10,ft hori-l zontal distance of any seismic Category I building wall. Therefore, l
analyses of lateral loads due to tunnel cuttings were not required.
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Insert 2 of 2 on page 2.5-129 (continued)
Soil properties used for the analysis were conservative values based primarily on published data in the literature. The properties are summarized in Table 2.5-20.
The shear modulus at low strain (less than 10-6 in./in.) for each element was determined using the relation Gmax = 1000 K2 (6m) where Gmax = shear modulus at 10-6 in./in. shear strain (psf)
K2 = shear modulus parameter, constant for a given soil type, density and ctrain level 6m = octahedral effective stress (psf)
The K2 value for the revetment stone was based on the average value for the rockfill shell in Oroville dam, (California Department of Water Resources, 1979) determined from cyclic triaxial data and from actual performance of the embankment during the 1975 Oroville earthquake. The K2 value for the glacial till was based on values for the deep alluvium at San Fernando Dams, reported in Seed et al. (1973) and on in situ measurements of shear wave velocity for a similar till in Boston (GEI, 1976). The K2 value for the offsite borrow was selected to represent the material with 90% compaction using data from Seed and Idriss (1970). Values of unit weight and Poisson's ratio for the of f-site borrow were based on Table 2.5-15.
For the rockfill and glacial till, values of unit weight and Poisson's ratio were estimated based on typical values in the literature.
A damping ratio of 0.5% at low strain was used for each of the soil types. The variation in shear modulus and damping with strain level were based on the curves presented in Seed and Idriss (1970).
For these analyses, the water level in the revetment and fill behind the revetment was assumed to be at El +14.5 MSL.
2 For evaluation of displacements, five trial wedges were selected through each revetment, as indicated on Fig. 2.5-57.
For each wedge, the horizontal yield acceleration required to reduce to factor of safety of the wedge to 1.0 was computed using a pseudo-static wedge ana-I lysis (U.
S. Army Corps of Engineers, 1970). The friction angles of the l
various materials which were used to compute the yield accelerations are l
shown in Table 2.5-20.
The friction angle for the of fsite borrow was l
based on triaxial test data presented in Table 2.5-15.
Values of fric-l tion angle for the revetment stone and glacial till were estimated based I
on data in Marsal (1972) and GEI (1981), respectively. The frictional angle between the filter fabric (Polyfilter X) and the adjacent soil was estimated based on the data presented in Haliburton et al. (1978).
The time history of average earthquake acceleration for a
(
given wedge was then compared to the yield acceleration for that wedge. Whenever the wedge acceleration exceeded the yield acceleration, horizontal displacement was assumed to occur. The total horizontal l
Insert 2 of 2 on page 2.5-129 (continued) displacement was determined by accumulating displacements through the duration of the earthquake. Settlement was computed by assuming that the computed horizontal displacement represented the horizontal com-ponent of downslope crest movement along the back side of the wedge, as shown in Figure 2.5-57.
The assumed displacements at the base of the wedge are also shown on Fig. 2.5-57.
The analyses indicate that the largest overall crest settlement for Revetment A resulting from the SSE event will be about 0.5 ft for Wedge 1 at Section R-R.
The analyses also indicate that the cap-stone at Revetment A (Wedge 3) may slump an additional 0 5 to 1.5 ft, resulting in a total settlement of 1.0 to 2.0 ft for the capstone.
For Revetment B, the largest overall crest settlement will be about 2.0 ft for Wedge 1.
Because of the thinner capstone and A-Stone layers at Revetment B, a separate analysis of the settlement of the capstone alone was not performed. Based on the hydrologic and wave runup analyses described in Subsection 2.4.5.5, the settlements at Revetment A or B resulting from the SSE evedt would not significantly affect the perfor-mance of the revetment.
The static stability of the highest section of the revetment, (Section R-R, Figure 2.5-56) was also analyzed using the wedge analy-sis described by the U.
S. Army Corps of Engineers (1970). The wedges analyzed were those shown on Fig. 2.5-57, plus a combined wedge con-sisting of the upper portion of Wedge 3 and the lower portion of Wedge 4.
The properties used in the analysis were as given in Table 2.5-20.
The minimum static factor of safety, F = 1.51, calculated s
for Wedge 4 is satisfactory for permanent slopes, based on the criteria given in U.
S.
Army Corps of Engineers (1970). This minimum factor of safety is considered to be a very conservative value due to the very conservative friction angle ($ = 36*) used for the revetment stone.
Using a best estimate of friction angle at low confining pressure, $=
46' based on data in Marsal (1972), the minimum static factor of l2 safety is Fs = 2.15.
2.5.5.3 Logs of Borings The general site exploration programs and boring logs are described and referenced in Subsection 2.5.4.3.
2.5.5.4 Compacted Fill Compacted fill behind the revetment is described in Subsection 2.5.5.1 and properties of the fill materials are presented in Subsection 2.5.4.5.c.
In, sert 1 of 1 on page 2.5-139 (continued)
U. S. Army " Corps of Engineers, 1970, Engineering and Design, Stability of Earth and Rock-Fill Dams, Engineers Manual No. EM 1110-2-1902, Appendix 7.
Haliburton, T. A.; Anglin, C. C.; and Lawmaster, J.
D.,
1978, " Testing of Geotechnical Fabric for Use as Reinforc,ement," ASTM Geotechnical Testing Journal, Vol.
1, Dec., pp. 203-212.
Seed, H.
B.
and Silver, M.
L.,
1972, " Settlement of Dry Sands During Earthquakes," American Society of Civil Engineers, Journal of the Soil Mechanics and Foundations Division, Vol. 98, No. SM4.
Schnabel, P.
B.; Lysmer, J. ; and Seed, H.
B.,
1972, " SHAKE, A Computer Program for Earthquake Response Analysis of Horizontally Layered Sites," Report No. EERC 72-12, College of Engineering, University of California at Berkeley, December 1972.
Seed, H.
B.
and Whitman, R.
V.,
1970, " Design of Earth T.etaining Structures for Dynamic Loads," Specialty Conference on Lateral Stress in the Ground and Design of Earth Retaining Structures, American Society of Civil Engineers, Soil Mechancs and Foundation Division.
Silver, M. L. and Seed, H.
B., 1971, " Volume Changes in Sands During Cyclic Loading," American Society of Civil Engineers, Journal of Soil Mechanics and Foundations Division, Vol. 97, No. SM9.
Bowles, J.
E.,
1977, Foundation Analysis and Design, Second Edition, McGraw-Hill Book Company, New York.
l2 Dalal, J.
S.,
1975, Memorandum of Telephone Conversation between J. S.
Delal, Seismic Consultant, United Engineers and Constructors, and Dr. H.
B.
Seed, Professor, University of California, Berkeley, February 5, 1975.
- Richart, F.
E.; Woods, R.
D.; and Hall, J.
R.,
Jr.,
1970, Vibrations of Soils and Foundations, Prentice-Hall, Inc., Englewood Cliffs, NJ.
A._
TABLE 2.5 Properties For Seismic Deformation Analysis of Revetment Property Revetment Offsite Borrow Glacial Filter Stone 90% or 954 Till cloth Compaction 1.
Unit Weight Saturated - below water 140 pcf 136 pcf 140 pcf Moist - above water 126 pcf 126 pcf 2.
Shear Modulus Parameter, K (1) 170 55 110 2
3.
Damping at low strain level (1 10-6 in./in.)
0.5%
0.5%
0.5%
2 4.
Poisson's ratio, Saturated - below water 0.3 0.48 0.48 Above water table 0.3 0.3 0.3 5.
Friction angle 36' 34*
36' 32' III 2 used to compute shear modulus at loy/2 where 5m is the Parameter K strain level (1 10-6 in./in.) with equation Gmax = 1000K (Um) 2 octahedral effective stress.
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