ML20054H912
| ML20054H912 | |
| Person / Time | |
|---|---|
| Site: | Seabrook  |
| Issue date: | 06/22/1982 |
| From: | Devincentis J PUBLIC SERVICE CO. OF NEW HAMPSHIRE, YANKEE ATOMIC ELECTRIC CO. |
| To: | Miraglia F Office of Nuclear Reactor Regulation |
| References | |
| SBN-283, NUDOCS 8206250180 | |
| Download: ML20054H912 (37) | |
Text
PUBUC SERVICE 5"* STA" Engineering Office:
Companyof NewHamped*e 1671 Worcester Road Framinonam. Massachusetts 01701 (617) - 872 8100 June 22, 1982 SBN-283 T.F. B 7.1.2 United States Nuclear Regulatory Commission Washington, D. C. 20555 Attention:
Mr. Frank J. Miraglia, Chief Licensing Branch No. 3 Division of Licensing
References:
(a) Construction Permit CPPR-135 and CPPR-136, Docket Nos. 50-443 and 50-444 (b) USNRC Letter, dated January 12, 1982, " Request for Additional Information - Seabrook Station," F. J. Miraglia to W. C. Tallman (c) PSNil Letter, dated March 12, 1982, " Responses to 241 Series RAIs; (llydrologic and Geotechnical Engineering Branch)"
Subjec t :
Amended Responses to 241 Series RAIs; (Hydrologic and Geotechnical Engineering Branch)
Dear Sir:
As a result of comments by the NRC Staff and United Engineers and Constructors (UE&C), we have amended a number of text pages, tables and figures which we originally submitted in Reference (c).
i On each text and table page the lines to which changes have been made are identified by a vertical bar and numeral 1 in the right margin. Changes have not been identified on the figures.
Very truly yours, YANKEE ATOMIC ELECTRIC COMPANY
[A.
y J. DeVinc n i Of $ Project Manager B206250180 820622 PDR ADOCK 05000443 A
]
i
SUMMARY
LIST Or AMENDED PAGES FOR MARCH /2, 1982 REVISIONS TO SEABROOK FSAR 1.
Text Pages Insert 1 of 1 on page 2.5-104 Insert 1 of 1 on page 2.5-104 (continued)
Page 2.5-106 Insect 1 of 1 on page 2.5-106 Page 2.5-107 Insert 1 of 2 on page 2 5-107 Insert 2 of 2 on page 2.5-107 Insert 1 of 3 on page 2.5-119 Insert 3 of 3 on page 2.5-119 Insert 1 of 2 on page 2.5-121 Insert 1 of 1 on page 2.5-124 Insert 1 of 1 on page 2.5-125 Insert 1 of 1 on page 2.5-125 (continued)
Insert 1 of 2 on page 2.5-126 (continued)
Insert 2 of 2 on page 2 5-126 Insert 2 of 2 on page 2.5-126 (continued)
Insert 2 of 2 on page 2.5-126 (continued)
Insert 2 of 2 on page 2.5-126 (continued)
Insert 2 of 2 on page 2.5-129 Insert 2 of 2 on page 2.5-129 (continued)
Insert 2 of 2 on page 2.5-129 (continued)
Insert 1 of 1 on page 2 5-139 (continued) 2.
Tables Table 2.5-12 page 1 of 2 l
Table 2.5-19 page 1 of 3 Table 2.5-19 page 2 of 3 Table 2.5-19 page 3 of 3 Table 2.5-20 3.
Figures Figure 2.5-42b Figure 2.5-44d Figure 2.5-52 Figure 2.5-53 Figure 2.5-54 Figure 2.5-55 Figure 2.5-56 Figure 2.5-58 r
t J
Project 81878 Geote-hnLcal Engineers Inc.
May 21, 1982 L
Insert 1 of 1 on page 2.5-104 Rock Quality Designation (RQD) is defined as the ratio of the length of sound pieces of core, 4 inches or longer, recovered _ in the core barrel to the distance that the core barrel was advanced, expressed as a percent. RQD was measured in ten of the borings made at the locations of site foundation excavations and in all of the borings along the course of the intake and discharge tunnels. An NX size (2-1/8-in.-
diameter) double tube core barrel was used for all rock cores for which RQD was evaluated. The RQD data are shown both numerically and graphi=
cally on the boring logs which are presented in Appendix 2F (borings E2-11 through E2-18) and in Reference 118 (Intake and Discharge Tunnel Borings).
The ten site foundation borings for which RQD was measured are summarized below:
Doring No.
Structure Inclination Elevation Range Avg RQD Avg RQD (degrees from (MSL) (in rock)
(%)
(%) (be-vertical) low El
-40*)
E2-11 Reactor 1 40
+11 to -102 86 89 E2-12 Reactor 1 41
+21 to -104 76 83 1
E2-13 Reactor 1 41
+30 to - 93 76 73 E2-14 Reactor 1 41.5
+27 to - 93 67 82 E2-15 Reactor 2 41.5
+ 4 to -110 66 73 E2-16 Reactor 2 41
+ 7 to -108 61 66 E2-17 Reactor 2 41
- 2 to -111 54 61 E2-18 Reactor 2 39
+ 3 to -112 49 53 AIT-1 Pumphouse Vertical
- 6 to -304 75 76 Intake Shaft ADT-1 Pumphouse Vertical
+ 3 to -288 81 83 Discharge Shaft
- El -40 is the approximate excavation grade for Reactors 1 and 2 and the intake and discharge structures.
Insert 1 of 1 on page 2.5-104 (continued) 1 4
I Borings E2-11 through E2-14 and E2-15 through E2-18 are inclined borings which were inade around the perimeter of Reactors 1 and 2, respectively. These borings were performed to provide data for design of the side slopes for the reactor excavations. Dorings AIT-1 and ADT-1 were made at the sites of the vertical intake and dischange shaf ts entering the east side of the Pumphouse. The boring locations are shown in Fig. 2.5-14.
i 1
RQD's obtained from the reactor site angle borings correlated well with conditions encountered in site excavations. The borings indicated generally very poor to fair quality rock within 10 to 20 f t of the J
bedrock surface. Poor gjality rock encountered in shallow excavations commonly required removal so that final excavation surfaces consisted of sound rock, as described in Subsection 2.5.4.14.
RgD values below the
]
top 20 f t of bedrock varied across the site depending upon rock type, l1
)
joint spacing and orientation relative to boring orientation, incipient jointing, and the incidence of dikes and faults.
In general, the lowest RQD values were obtained in rock where joints were closely spaced and i
joint surfaces were highly polished, coated with chlorite or severely weathered. These conditions were observed more commonly in the borings i
penetrating diorite and quartzite beneath Reactor 2.
A reasonable I
indication of areas where poor quality rock would have to be removed in the deep excavations was provided by RQD values. A few areas of poor j
quality rock not readily identified by the borings were encountered in i
the deep excavations and required excavation beyond the design lines.
l Treatment of areas requiring overexcavation is described in subsection
]
2.5.4.14.
I In the diorite rock at the tunnel shaf t locations, Boring AIT-1 indicated somewhat lower RQD above El -130 than ADT-1.
However, during excavation, conditions at the two shaf ts were quite similar, indicating that the poorer RQD values in AIT-1 can be attributed to the predomi-j nance of low angle joints in this boring. These shallow dipping joints were of minor significance during excavation.
4 Several other borings of significant depth were made at the site (Figure 2.5-114) but no RQD's were measured for this core. These included borings of the B, D, and E series made from 1968 through 1974, j
logs for which are included in Appendix 2D.
Rock quality as represented by fracture spacing on these logs is consistent with the rock quality as seen in excavations and the borings with measured RQD on the site.
RQD values were used primarily to provide general indications of rock quality across the site.
RQD values, however, were also used to estimate Young's modulus of the in situ rock mass by using empirical relationships correlating RQD with Young's modulus as determined from laboratory tests on intact samples or from seismic velocities as deter-mined in the field. This use of RQD is discussed in Section 2.5.4.5.b.
1 l
_~ _
5.
Compression and Shear Wave Velocities Comp
'on (P) wave velocities for the bedrock at an Replace wdh the vicinity e site were measured by seismir ettfveys,
'Entectic81 uphole and crosshole gt.
ical tests aboratory sonic tests. The data indicate tha ajor rock types (diorite and quartzitic schis e essentially me P wave velocity.
Therangeg e velocities for each type of meas nt p hewn in Table 2.5-12.
ff*m ITn g uj
,o vi A Shear (S) wave velocities of the diorite were measured in
{
I) the crosshole and uphole tests. The range of test results pj is summarized in Table 2.5-12.
a g
e c-E' The seismic surveys and uphole and crejshole tests are described u.$,f,d in Subsection 2.5.4.3.
The labdratory sonic wave velocity both de ad tests were performed onkintact core specimens using procedures Sat""" Ad f'g generally in accordance with ASTM D2845.
Detailed results g*$j.
of the sonic tests are contained in Appendix 2G and in Reference 6
67
- 117,
.f 5 o 1 -i The. proceduces u. sed to deiermiac densg g
jty}N Dener*IIO20 45, M
$^ A cco rd Av.c c tad h I
(.
6.
Rock Density
- c'c
?d ASTM
't hC Values of density were determined for selected rock specimens f
a 2 T 8'f used in the laboratory testing programs.
Results of these y 3 (9 g
measurements are summarized in Table 2.5-12.
Detailed test m5o y'
results are contained in Appendix 2G and Reference ll7.y d,,5.c $
sw
- o. e 7.
Unconfined-Compressive Strength C d.$
3d#
Unconfined-compressive strength was measured on 31 air-dry rock core specimens using procedures described in ASTM D2938.
The test results are summarized in Table 2.5-12.
I The data indicate a wide variation in strengths for each type of rock.
The diorite has an average compressive strength of 18,300 psi and the quartzite an average compressive strength of 12,100 psi. The. one 4est on ctuocitt.iir schist in the. plani ar ect indica +es coupcessive. 54cenef h Srndar to the d'iocGe..
Detailed test data are presented in Appendix 2G and Reference 117.
8.
Young's Modulus The precedace c use.d eecc. ge.ne co_llg in I
av.oedawce. tod h ASTM D31'4 B Values of initial tangent Young's modulus (E ) for static i
loading were determined from 25 of the unconfined-compression tests on rock core specimens. On 12 of the specimens, the secant modulus at 50% of the ultimate compressive strength (Es50) was calculated, while on the other 13 specimens, the tangent raodulus at 50% of ultimate compressive strength (EtS0) was computed. The range of values for each of these f'
moduli is shown in Table 2.5-12, 2.5-106
)
_~ _ -
e Insert 1 of 1 on page 2.5-106 l1 In situ compression (P) wave velocities for the bedrock types at and in the vicinity of the site were measured by ceismic surveys and uphole and crosshole geophysical tests. Compression wave velocities were also measured by laboratory sonic tests. The data indicate that both major rock types (diorite and quartzite) have essentially I
the same P wave velocity. Both the in situ and laboratory data indicate uniform P wave velocities across the site area, indicating a general uniformity of bedrock properties for engineering design.
j-The range of wave velocities for each type of measurement is shown on Table 2.5-12.
The average P wave velocity for the saturated y
4 laboratory specimens was 16%' higher than for the dry specimens. The sonic test results shown in Appendix 2G indicate that an increase in l
confining pressure from 0 to 3000 psi causes less than a 3% increase in P wave velocity. An increase in axial load from 0 to 1000 psi causes less than 4% increase in P wave velocity.
i l
f t
l 1
4 1
}i
..., _ _, _,.. _ _ -, _ _ _ - _, _ _..... _, -, _ -,, _. ~. -.
~.
_ _ - __ -. ~
'M Peptoce wclh The data tnarira w ethodulus.yalues-f6i Lush major types Insert lef1 of rock (d'^ rite
="d quartzite).
Detailed test results are contained in Appendix 2G and Reference 117.
9.
Poisson's Ratio Poisson's ratio for static loading was determined for 10 of the unconfined compression tests on rock core specimens.
Values were calculated at the start of loading and at a load j
of 50% of ultimate compressive strength.
The range of values ts shown in Table 2.5-12.
Poisson's ratios for the diorite samples covered the entire i
range shown, while values for the quartzite (2 tests) were at the low end of the range.
Individual test results are tabu-lated in Appendix 2G.
- Poisson's ratios for dynamic loading were determined from the uphole and crosshole geophysical tests. These tests were performed in the diorite only, and summarized in Table Tmer-p 2.5-12.
A lof1 j
10.
Dynamte Shear Modulus I
Values of dynamic shear modulus for the quartz diorite were determined from the uphole and crosshole tests.
The range of values is shown in Table 2.5-12.
11.
In-Situ Rock Stresses
~
In situ rock stresses were measured in Boring OCIA, near the center of Reactor No. 1.
This location was selected because the seismic and geophysical testing indicated higher than average compressional wave velocities in this area.
The rock stresses were measured at five points between depths of 33 and 42 f t.
The values of horizontal compressive stress i
j are summarized below:
I Range
,_ 2
- Average Stress Largest 150 to 2,150 psi 1,240 psi stress Smallest 50 to 1,570 psi 860 psi stress l
-s i
2.5-107 i
i e
m-,--o-e~
w s
v n.-
Insert 1 of 2 on page 2.5-107 1
+
The data indicate similar modulus values for both major types of rock (diorite and quartzite) at the plant site. The one test on quart-zitic schist in the plant area indicated modulus similar to the diorite and the quartzite.
Young's modulus was used in calculating foundation heave and settle-ment due to excavation and structural loads, respectively. Adjustment of tho Young's modulus measured on intact specimens, to account for the average RQD in the field, is discussed in Subsection 2.5.4.5.b.
Young's modafus was not used in dynamic analyses since the rock was assumed to l1 be a rigid boundary below foundation grades as described in Subsection 3.7(b).2.3.
L_
i e
r Insert 2 of 2 on page 2 5-107 Values of Poisson's ratio were used in calculating foundation haave and settlement due to excavation and structural loads, respectively.
1 They were not used in dynamic analyses as the rock was assumed to be a rigid inundary below foundation grades as described in Subsection 3.8.5.1.
d l
Insert 1 of 3 on page 2.5-119 j
j Measurements of heave or rebound of the rock in the excavations were not taken. However, no instances of rock behavior or excavation and foundation movements attributable to heave were observed. This is con-sistent with expected behavior, since the predicted maximum rock heave
)
in seismic Category I excavations was 0.25 in. in the center of the bot-tom of the reactor excavations. Heave of this magnitude would have no significant ef fect on the properties of the rock below the excavations or on the performance of the structures placed in the excavations.
1 The heave was estimated using a Boussinesq elastic pressure distri-bution for the unloading. of 80 ft of soil and rock overburden for a 200-ft-diameter-area. The equation used to compute the heave was 1
6 = 2 gr (1 - V)2 field 7
j where 6 = heave, in.
q = magnitude of unloading, psi r = radius of unloade6 area, in.
y = Poisson's ratio Efte14 = In situ Young's modulus The elastic modulus used for the heave analysis was Efield = 1.0 x 106 psi.
This value of Efield was determined by correcting the average laboratory modulus for intact rock at the reactor locations, Elab = 10 i
x 106 psi (Appendix 2G, Table 2G-1), to account for the average RQD of i
the rock below the excavation using empirical data presented in Hendron 1
(1960). For the average RQD = 60% at the Unit 2 reactor site, which is i
lower than the average at the Unit 1 reactor site, the correction factor is Efield/Elab = 0.1.
I-i i
f 1
J f
a e
-w w
y--
.-----.-.------ m--.
Insert 3 of 3 on page 2.5-119 The maximum thickness of of fsite borrow beneath safety-related electrical ductbanks was 18 ft.
Typical profiles of of fsite borrow l1 beneath safety-related electrical ductbanks are shown in Sections KK, LL, and MM on Figure 2.5-42c.
The maximum thickness of of fsite borrow placed beneath safety-related electrical manholes was 18 ft.
Typical profiles of offsite borrow beneath safety-related electrical manholes are shown in Figure 2.5-42c, Section MM and Figure 2.5-42d, Sections NN and PP.
The thickness of of fsite borrow beneath safety-related service water pipes ranged from approximately 2 to 15 f t.
Typica1 profiles in which l1 the depth of offsite borrow beneath safety-related service water pipes ranges from 2 to 5 f t are shown in Figure 2.5-42c, Sections JJ, KK, LL, and HM.
A typical profile in which the depth of offsite borrow beneath safety-related service water pipes ranges from approximately 10 to 1
15 ft is shown in Figure 2.5-42b,Section II.
The maximum thickness of offsite borrow placed adjacent to safety-related structures is approximately 45 ft along the west walls of the Intake / Discharge Transition Structurus.
Insert 1 of 2 on page 2.5-121 Representative field density test results for offsite borrow placed in the plant area during typical winter, spring, and summer periods are shown in Figures 2.5-44, 44a and 44b, respectively. As seen in these I
three figures, the offsite borrow was compacted to at least 95% of ASTM D1557-70 coinpaction.
Of fsite borrow was also used as fill behind the revetments, as shown in Figures 2.5-54, 55, and 56.
Most of this offsite borrow was com-pacted to at least 90% of maximum dry density determined by ASTM D1557-70. Typical field density test results of the offsite borrow 3
placed to at least 90% compaction are shown in Figure 2.5-44d.
Near the railroad tracks at the west end of Revettaent A, the of fsite borrow was placed to at least 95% compaction as shown on Figure 2.5-44c.
i l
Insert 1 of 1 on page 2.5-124 e.
Random Fill Random fill was used for nonsafety-related general site backfill and grading in areas not requiring engineered backfill.
Random fill consisted of offsite borrow, tunnel cuttings, and soil from onsite excavations with less stringent placement reqirements than the engineered backfill.
Random fill was placed in 8-to 12-in. lif ts and compacted to at least 90% of the maximum dry density determined by ASTM D1557-70.
In the plant area the maximum vertical thickness of random fill is about 40 ft at a location between the Circulating Water / Service Water Pumphouse and the Intake / Discharge Transition Structure, as shown on Section A-A in Figure 2.5-42.
At this location engineered backfill with a minimum horizontal extent of 10 ft was placed against the walls of the structures.
Random fill was placed between the areas of engineered backfill.
l1 Beyond the plant site, areas where random fill was used to raise the general site grade may be determined by comparing Figures 2.4-1 (plan of the final plant grade) and 2.5-12 (plan of origi-nal site contours).
I a--
Inscrt 1 of 1 on paga 2.5-125 For all seismic analyses the rock was treated as a fixed boundary, as described in Subsection 3.7(B).2.3.
Therefore, no dynamic rock pro-perties were required for the seismic analyses.
Four seismic Categorf I electrical manholes are founded on offsite borrow, with a maximum thicknass of 18 ft below the base of manhole W33/34. Seismic amplification in the maximum thickness of fsite borrow was analyzed using the lumped-mass and spring approach described in Subsection 3.7(B).2.4.
During design, estimated values of shear wave velocity c = 650 ft/sec and shear modulus, G = 13,890 psi 3
were used.
Based on data in Richart et al. (1970), v 650 ft/sec is a
=
3 y
conservative (low) value for die of fsite borrow. These values were assumed to be constant for the 18-ft-thick layer of offsite borrow. The amplified accelerations were used for the structural analyses of walls of manholes on offsite borrow.
Subsequent to design, the shear modulus for offsite borrow beneath the manholes was backfigured from the results of the plate load test described in Subsection 2.5.4.5.d.
The Young's modulus for the cyclic portion of the plate loading, E = 24,800 psi, and the Poisson's ratio from drained triaxial tests, V = 0.3, were used to calculate a value of G = 9,550 psi.
The average degree of compaction for the of fsite borrow test fill was the same as for the offsite borrow placed during construc-tion.
Assuming that die 18-in.-diameter plate influenced a 24-to 36-in.-thick layer of soil beneath the plate, the average shear strain, y, in the soil during the unload-reload cycle was Y = 3.6 x 10-3 in./in.
co 5.4 x 10-3 in./in. The value of shear modulus at low strain (10-6 in./in.), Gmax, was then determined using the relationship between shear modulus and shear strain for sand presented in Seed and Idriss (1970). The average octahedral stress, 6m = 4,000 psf, in the zone beneath the plate was calculated using the elastic solutions for a rigid plate with an average load of 6 tsf.
Values of G for other effective 4
max stress levels were then computed using the relation U /U G2 max " G1 max m2 ml where G1 max and Um1 were values from the plate load test.
A plot of Gmax V8 Um for the offsite borrow, based on the plate load test data, is shown in Fig. 2.5-58.
As noted in Subsection 3.7(B).2.4, the seismic design of the manholes was checked using the shear modulus values backfigured from the plate load test, and was found to be satisfactory.
i The seismic Category I electrical ductbanks which are founded on of f site borrow were analyzed using the procedures described in Sub-section 3.7(B).2.4.
The dynamic properties backfigured from the plate load test as described above were used for these analyses.
Electrical manhole W19/20 is founded on 10 ft of tunnel cuttings with a few layers of offsite borrow. Analysis of the amplification for this manhole was performed as described in Subsection 3.2(B).2.4 with an average shear modulus determined from the plate load test on tunnel cuttings using the procedure described above.
G Insart '1 of 1 on page 2.5-125 (continusd) r The seismic Category I service water pipes are supported on offsite borrow generally 2 ft thick, but in certain areas up to a maximum of 15 ft thick, as described in Subsection 2.5.4.5.c.3.
All seismic Category I pipe is surrounded by offsite borrow except for a 180 ft length of trench near the service water pumphouse where the pipe is surrounded by sand-coment as described in Subsection 2.5.4.5.c.4.
The thickness of cover over the pipes is from 12 to 24 f t except for the pipes between the service water pumphouse and the intake / discharge structures where the cover is up to 60 ft.
The seismic stresses in these pipes were ana-lyzed using the procedures of Iqbal and Goodling, 'as described in Subsection 3.7(B).3.12.
For the pipe surrounded by offsite borrow, the following design parameters were used in the analyses:
Unit Weight (buoyant)
Yb = 60 pcf Void ratio e = 0.4 Poisson's ratio V = 0.4 Coefficient of Lateral Pressure K = 0.5 o
Coef ficient of Subgrade Reaction k = 300 lb/in.2/in.
o Coefficient of Friction (steel pipe to soil)
D = 0.3 Maximum soil particle velocity for OBE Vm = 6 in./sec for SSE Vm= 12 in./sec Shear wave velocity C = 770 ft/sec 3
Seismic soil strain for OBE Em = 0.000325 in./in.
for SSE Em = 0.000650 in./in.
The values of unit weight, Poisson's ratio, void ratio, and coef-ficient of lateral earth pressure are conservative values, based on the results of field density measurements shown in Figs. 2.5-44 through y
2.5-44c and triaxial tests shown in Table 2.5-15.
The coefficient of subgrade reaction is a lower-bound value from Figure 5 in Appendix 2N, which is based on the results of triaxial compression tests.
Since the lower-bound value of k is not necessarily conservative, a check analy-o sis is being performed with higher values of k. The coefficient of o
f riction between soil and pipe is conservative, based on the measured f riction angles from triaxial tests and the reduction factors recom-mended in Iqbal and Goodling. The maximum soil particle velocity, shear wave velocity, and seismic soil strain were determined using the proce-dures described in Iqbal and Goodling.
The section of service water pipe surrounded by sand-cement was analyzed using an average shear modulus, G = 10,100 psi backfigured from the initial Young's modulus measured the consolidated drained triaxial compression tests described in Appendix 2M.
The average in situ octa-hedral effective stress in the 8-f t-thick sand cement layer (see Section J-J on Fig. 2.5-42c) is 6'm " 7*1 Psi, therefore the modulus t
Insurt 1 of 2 on paga 2.5-126 (continued)
)
The sand-ccment, described in Subsection 2.5.4.5.c.4, was used as backfill adjacent to and above one 180-ft-long section of service water pipe. Since the strength of sand cement is derived primarily l1 f rom cementation at grain contacts, and not from interparticle friction, loss of strength due to buildup of pore pressure, i.e.,
liquefaction, is not possible. The resistance of sand cement to cyclic deformations was l
not measured in the laboratory since the in situ cyclic shear stress induced by the SSC in the sand cement is less than 4% of the specified minimum compressive strength of 100 psi at 28 days, and less than 2% of the minimum compressive strength measured on field test cylinders at 90 l
days, as shown in Fig. 2.5-45.
Cyclic stresses of this magnitude would l1 j
not be sufficient to break the cementation bonds, hence no significant cyclic deformations could occur. There are no known subsurface con-ditions at the site which could lead to future loss of strength in the l
sand-cement.
It is likely that the strength will increase with time i
since the strength increased with time in the laboratory. Thus, it was concluded that the sand cement is adequately resistant to both liquefac-tion and cyclic deformations.
f f
1 l
f l
I l
I e-w a -- - - -,
-=-e, - - - - -,
.= _ -.
Insert 2 of 2 on page 2.5-126 I
All seismic Category I structures are founded on sound bedrock or on engineered backfill extending to sound bedrock. As shown in Table 2.5-19, fill concrete was used as the engineered backfill beneath all seismic Category I structures except for safety-related electrical duct banks, five electrical manholes, and the service water pipes, which were i
founded on of f site borrow or tunnel cuttings, a.
Bearing Capacity and Static Settlement Havdocks DM-7 (1963) was used to estimate the allowable bearing pressure for structures founded on borock.
In Table 11-1 of Navdocks DM-7 (1963) an allowable bearing pressure for hard crystalline rocks of 80 taf is recommended.
(Note: The allowable bearing pressure is the pressure that can be applied in the field. The ultimate bearing capa-city is 6 to 10 times higher than the allowable value. )
I' An alternative technique for estimating allowable bearing pressure on rock is to multiply die unconfined compressive strength by 0.2 to 0.3 to adjust for the presence of rock defects, as suggested by Dowles (1977, p.
143).
For the rock at this site the lowest measured unconfined compressive strength in the zone of interest was 5970 psi (Table 2G-1).
Using the factor 0.2, Bowles' approach gives a value of 06 tsf for the allowable bearing pressure. This value is similar to that recommended in Navdocks DM-7.
Some structures are founded on fill concrete, which has a 90-day i
unconfined compressive strength of 5400 psi (Fig. 2.5-45a). Following Bowles' (1977, p.
143) suggestion to use no more than the unconfined compressive strength of the concrete as a working compressive strength of the rock, and using the factor of 0.2, the allowable bearing pressure on the rock is calculated to be 78 taf.
Thus, an allowable bearing pressure of 80 tsf is suitable for foundations on rock, with or without fill concrete between the two.
To be conservative an allowable bearing pressure of 60 tsf was used.
The actual bearing pressures beneath the major seismic Category I buildings are shown on Table 2.5-19.
The foundation type and dimen-sions are presented in Table 3.8-15.
The most highly loaded foundation is the ring wall around the containment enclosure buildings which have a maximum bearing pressure of 36 tsf.
The maximum estimated settlement for any seismic Category I structure is 0.5 in. for the combined loading of the containment struc-ture and containment enclosure structure. Of this settlement, approxi-mately 0.25 in. represents recompression of the heave resulting from the excavation, as described in subsection 2.5.4.5.b.
The differential movement between these two structures is estimated to be less than 0.1 I
in.
The maximum settlement was estimated using the relationship 2
2 gr (1 - v )
d E
where 6 = total settlement at center of foundation, in.
E = average modulus of elasticity, psi q = average foundation bearing pressure, psi r = radius of loaded area, in.
V = Poisson's ratio
Insert 2 of 2 on page 2.5-126 (continued)
An average modulus E = 1 x 106 psi, corrected for RQD as described in Subsection 2.5.4.5.b, was used for the analysis. The weighted average loading for the combined containment structure,and containment enclosure structure was q = 17.2 tsf over a radius of 86.5 ft.
The value of Poisson's ratio, V = 0.2, used for this analysis was 4
conservative, based on the compression test dat a in Table 2.5-12.
The estimated settlement will occur as elastic compression during construction as the load is added. No significant post construction 1
l settlements or differential settlements for foundations on rock or fill concrete are anticipated.
For die manholes supported on offsite borrow or tunnel cuttings, an l
allowable bearing pressure of 2.5 tsf was established. The minimum base size for the manholes is about 18 ft by 18.5 ft, with a minimum embed-a ment of 9.5 ft.
The ultimate bearing capacity was calculated using the Terzaghi bearing capacity formula qu = 1/2BYNy + YD Nfq l1 Using a submerged unit weight, Yb = 72 pcf and bearing capacity fac-tors N = 50 and N = 40 for a minimum friction angle $ = 36* as deter-y q
mined trom triaxial tests, the ultimate bearing capacity for the offsite borrow is qu = 30 tsf.
For a submerged unit weight Yb = 97 pcf l
and bearing capacity factors N = 70 and N = 70 for an assumed & =
y q
40*, the ultimate bearing capacity of the tunnel cuttings is qu = 63 tsf.
Thus, the allowable bearing pressure provides factors of safety of 12 and 25 against ultimate bearing capacity failure for the offsite borrow and tunnel cuttings, respectively.
The maximum bearing pressure beneath the base of the manholes is 1.4 1
tsf, assuming the water table is below the bottom of the manhole. Using the elastic settlement formula described above, with C = 10,500 psi from the plate load test data and V = 0.3 from the triaxial test data, the maximum settlement for manholes on the of fsite borrow is 6 = 0.5 in.
l1 For the tunnel cuttings, with E = 24,000 psi from the plate load test-data and estimated V = 0.3, the maximum settlement of the one manhole on tunnel cuttings is 6 = 0.2 in.
l1 The estimated settlement will occur during the construction of the j
manholes and as backfill is placed around the manholes. No significant post construction or dif ferential settlements of the manholes founded on of fsite borrow or tunnel cuttings is expected, unless a seismic event occurs, which is covered in Section 2.5.4.10.b.
I
Insert 2 of 2 on page 2.5-126 (continued) b.
Settlement Due to Seismic Loading The settlement resulting from the 3SE loading was also estimated for the seismic Category I structures founded on of fsite borrow or tun-nel cuttings, using the relationship between horizontal cyclic shear strain in the soil durP.g the earthquake and accumulated vertical strain described in Silver and Seed (1971) and Seed and Silver (1972). The peak horizontal cyclic shear strains were determined for the thickest layers of offsite borrow and tunnel cuttings below seismic Category I struc-tures (see Subsection 2.5.4.5.c) using the one-dimensional computer program SHAKE (Schnabel et al.,
1972) with 3 to 7 soil layers below the structure. The SHAKE analyses were performed using the lower values of 3
shear moduli from the plate load tests (see Subsection 2.5.4.7) and the shear modulus reduction curve and damping values from Seed and Idriss (1970). Using the Seed and Silver (1971) data for relative density, Dr = 80% and 10 cycles of loading, the maximum seismic settlement for seismic Category I structures is 0.1 in. for the service water pipes located in the area with 15 ft (maximum thickness) of offsite borrow y
beneath the pipes. Seismic settlements of this magnitude will not affect the performance of the seismic Category I manholes, ductbanks or service water pipes during or after the SSE event.
c.
Static and Dynamic Lateral Pressures Lateral earth pressures for Category I structures surrounded by offsite borrow were computed for both static and seismic conditions using the pressure diagrams shown in Figures 2.5-52 and 2.5-53.
The static coefficients of at-rest earth pressure, K and active earth o,
- pressure, K,
are conservative values, based on the minimum friction a
angle of 36' measured in triaxial tests.
Static water pressures were computed using the maximum groundwater elevation at the ground surface, El +20.
For the rigid wall, a static lateral compaction pressure was included for the full height of the wall.
The dynamic lateral pressure coefficient, Kh, for non-rigid walls was calculated using the procedures described in Seed and Whitman (1970). As discussed in Seed and Whitman, the effect of vertical acce-1eration on the dynamic lateral pressure for non-rigid walls is negli-gible for the case where vertical acceleration is one half the horizontal acceleration.
The magnitude and distribution of dynamic lateral pressure of rigid walls was based on discussions with Dr.
H.
B.
Seed (Delal, 1975).
Based on his experience Dr. Seed recommended that an approximately uni-form pressure distribution would be appropriate for rigid box-type structures founded on rock and surrounded by soil. The lateral pressure 1
would be one half the maximum pressure from the Seed and Whitman (1970) method, increased by an empirical factor of 3 to account for the dif-ference in stiffness between the structure on rock and the surrounding soil. Thus, the coefficient of dynamic earth pressure, KD, is i
Incart 2 of 2 on pago 2.5-126 (continusd)
KD = 1/2 x 3 x AKAE 1/2 x 3 x 3/4 x a
=
max 1.125 a
=
max where AKag = 3/4 amax is the dynamic component of lateral earth 1
pressure f rom Seed and Whitman ( 1970). The uniform pressure distribu-tion is conservative, since the actual dynamic component of lateral pressure must go to zero at the base of the wall, where there is no relative dynamic motion between the structure and.the surrounding soil.
Near the very top of the wall, the horizontal earthquake pressure is limited to be equal to the passive resistance of the soil. The rigid walls of all seismic Category I buildings except five manholes were founded on sound bedrock or fill concrete extending to sound bedrock.
For these walls, the bedrock accelerations were used to compute dynamic lateral pressures. For the five manholes founded on offsite borrow or tunnil cuttings, the amplified soil accelerations at the base of the manholes described in Subsection 2.5.4.7 were used for design.
The maximum lateral pressures for any seismic Category I struc-ture occur at the east wall of the service water pumphouse, where the thickness of the offsit borrow is 63 ft.
This is a rigid wall with the following lateral pressures at the base:
Static At-Rest Soil Pressure 1,970 psf Hydrostatic Pressure 3,930 psf Permanent Jurcharge O
Live Load Surcharge 250 psf Dynamic Soil Pressure (SSE) 2,210 psf l1 Using the above values, dhe total horizontal earth load, excluding tha live load and the hydrostatic water pressure, is calcu-lated to be 190 k/f t.
The at-rest earth load alone is 62 k/ft. Hence, the total earth load during an earthquake is 3.1 times the at-rest (Ko=
1 0.5) carth pressure.
The hydrostatic water and the live load surcharge effects are added to the earthquake load for design of the walls. The offects of compaction need not be included during an earthquake since the shaking dissipates compaction prestress effects.
Tunnel cuttings were not placed against nor within 10 ft hori-zontal distance of any seismic Category I building wall.
Therefore, analyses of lateral loads due to tunnel cuttings were not required.
1
Insert 2 of 2 on page 2.5-129 2.5.5.1 Slope Characteristics (Revetment)
The stone revutment, located as shown in Fig. 2.5-41a, has a maximum height of 19 ft, of which 6 f t are buried below the ground surface. Thus, the maximum not revetment height is 13 ft.
Subsequent 1
references to revetment height are in terms of not height above finish grade at the toe.
Although no special exploration programs were per-formed for the revetment, the general site borings described in Subsection 2.5.4.3 indicate the natural soils at the revetment locations are 1 to 2 ft of topsoil underlain by dense glacial till extending to bedrock. The glacial till is described in Subsection 2.5.4.2.b.
The depth to bedrock varies from 0 to approximately 40 f t beneath the 4
revetment, as determined from the contours in Figs. 2.5-12 and 2.5-14.
The topsoil was stripped and the revetments are founded on bedrock or on glacial till. The fill behind the revetment consists of offsite borrow compacted to at least 90% of maximum dry density determined by ASTM D1557 except in the area near the railroad tracks, identified on Fig.
1 2.5-44c, where the offsite borrow was compacted to at least 95% of maxi-mum dry density. The fill materials are described in Subsection 2.5.4.5.c.
2.5.5.2 Design Criteria and Analysis of Revetment Wave design of the revetment is described in Subsection 2.4.5.5.
Seismic analysis of the revetment was performed to determine the deformations during an SSE event. The analysis was based on a time history of acceleration and seismic shear stress from the 2-dimensional finite element program FLUSH (Lysmer eti al., 1975) followed by a com-puter integration of rigid-body displacements using a Newmark type ana-lysis for a wedge failure surface. Three cross sections of revetment were considerei in the analysis:
4
]
a.
Section Q-Q - Revetment A, thickest underlying soil (about 40 ft) with revetment height of 10 ft b.
Section R-R - Revetment A, highest section (13 ft) c.
Section S-S - Revetment B, highest section (13 ft) with l
1 thickest soil below revetment (15 ft)
The locations of these sections are shown on Fig. 2.5-41a.
The soll profiles and finite element grids for these three sections are shown in Figs. 2.5-54 through 2.5-56.
The horizontal earthquake motion input at the bedrock surface was modelled using the Housner artificial record, scaled to match the j
design SSE spectrum for 5% damping shown in Figure 2.5-30 for the range of frequencies of 2 to 8 Hz, which brackets the natural frequencies of all three soil / revetment sections. An upper cutoff frequency of 15 Hz was used in the analysis. The duration of the design earthquake motion was 20 seconds.
l I
c
Insert 2 of 2 on page 2.5-129 (continued)
Soil properties used for the analysis were conservative values based primarily on published data in the literature. The properties are summarized in Table 2.5-20.
The shear modulus at low strain *less than 10-6 in./in.) for each element was determined using the relation Gmax = 1000 K2 (am) where Gmax = shear modulus at 10-6 in./in. shear strain (psf)
K2 = shear modulus parameter, constant for a given soil type, density and strain level ctahedral effective stress (psf) 6
=
m The K2 value for the revetment stone was based on the average value for the rockfill shell in Oroville dam, (California Department of Water Resources, 1979) determined from cyclic triaxial data and from actual performance of the embankment during the 1975 Oroville earthquake. The K2 value for the glacial till was based on values for the deep alluvium at San Fernando Dams, reported in Seed et al. (1973) and on in situ measurements of shear wave velocity for a similar till in Doston (GEI, 1976). The K2 value for the offsite borrow was selected to represent the material with 90% compaction using data from Seed and Idriss (1970). Values of unit weight and Poisson's ratio for the of f-1 site borrow were based on Table 2.5-15.
For the rockfill and glacial till, values of unit weight and Poisson's ratio were estimated based on typical values in the literature. A damping ratio of 0.5% at low strain was used for each of the soil types. The variation in shear modulus and damping with strain level were based on the curves presented in Seed and Idriss (1970). For these analyses, the water level in the revetment and fill behind the revetment was assumed to be at El +14.5 MSL, the maximum tide level during the Probable Haximum Hurricane, as described in Subsection 2.4.2.2.d.
For evaluation of displacements, five trial wedges were selected through each revetment, as indicated on Fig. 2.5-57.
For each wedge, the horizontal yield acceleration required to reduce to factor of saf ety of the wedge to 1.0 was computed using a pseudo-static wedge ana-lysis (U. S.
Army Corps of Engineers, 1970). The friction angles of the various materials which were used to compute the yield accelerations are shown in Table 2.5-20.
The friction angle for the of fsite borrow was based on triaxial test data presented in Table 2.5-15.
Values of fric-tion angle for the revetment stone and glacial till were estimated based on data in Marsal (1972) and GEI (1981), respectively. The frictional angle between the filter fabric (Polyfilter X) and the adjacent soil was estimated based on the data presented in Haliburton et al. (1978).
The time history of average earthquake acceleration for a given wedge was then compared to the yield acceleration for that wedge. Whenever the wedge acceleration exceeded the yield acceleration, horizontal displacement was assumed to occur. The total horizontal
= -. _.
1 l
Insert 2 of 2 on page 2.5-129 (continued) 1 displacement was determined by accumulating displacements through the j
duration of the earthquake. Settlement was computed by assuming that the computed horizontal displacement represented the horizontal com-ponent of downslope crest movement along the back side of the wedge, as shown in Figure 2.5-57.
The assumed displacements at the base of the wedge are also shown on Fig. 2.5-57.
The analyses indicate that the largest overall crest settlement for Revotment A resulting from the SSE event will be about 0.5 ft for Wedge 1 at Section R-R.
The analyses also indicate that the cap-stone at Revetment A (Wedge 3) may slump an additional 0.5 to 1.5 ft, resulttng in a total settlement of 1.0 to 2.0 ft for the capstone. For Revotraent B, the largest overall crest settlement will be about 2.0 f t for Wedge 1.
Because of the thinner capstone and A-Stone layers at Revetment B, a separate analysis of the settlement of the capstone alone was not performed. Based on the hydrologic and wave runup analyses described in Subsection 2.4.5.5, the settlements at Revetment A or B resulting from the SSE event would not significantly affect the perfor-mance of the revotment.
The static stability of the highest section of the revetment, (Section R-R, Figure 2.5-56) was also analyzed using the wedge analy-sis described by the U. S. Army Corps of Engineers ( 1970). The wedges analyzed were those shown on Fig. 2.5-57, plus a combined wedge con-sisting of die upper portion of Wedge 3 and the lower portion of Wedge 4.
The proporties used in the analysis were as given in Table 2.5-20.
The minimum static factor of safety, F = 1.51, calculated s
for Wedge 4 is satisfactory for permanent slopes, based on the criteria given in U.
S. Army Corps of Engineers (1970). This minimum factor of 4
safety is considered to be a very conservative value due to the very l1 conservative friction angle ( $ = 36') used for the revetment stone.
Using a best estimate of friction angle at low confining pressure, &=
l1 l
46' based on data in Marsal (1972), the minimum stwatic fawctor of safety is F = 2.15.
g i
2.5.5.3 Logs of Borings The general site exploration programs and boring logs are described and referenced in Subsection 2.5.4.3.
2.5.5.4 Compacted Fill Compacted fill behind the revetment is described in Subsection 2.5.5.1 and properties of the fill materials are presented in Subsection 2.5.4.5.c.
l 1
Ineset 1 of 1 on paga 2.5-139 (continued)
U. S. Army " Corps of Engineers, 1970, Engineering and Design, Stability of Earth and Rock-Fill Dams, Engineers Manual No. EH 1110-2-1902, j
Appendix 7.
l Haliburton, T. A. ; Anglin, C. C. ; and Lawmaster, J. D., 1978, " Testing of Geotechnical Fabric for Use as Reinforcement," ASTM Geotechnical j
i Testing Journal, Vol.
1, Dec., pp. 203-212.
l l
- Seed, H.
B. and Silver, M.
L.,
1972, " Settlement of Dry Sands During l
Earthquakes," American Society of Civil Engineers, Journal of the Soil Mechanics and Foundations Division, Vol. 98, No. SM4.
Schnabel, P. B.; Lysmer, J.; and Seed, H.
B.,
1972, " SHAKE, A Computer l
Program for Earthquake Response Analysis of Horizontally Layered l
Sites," Report No. EERC 72-12, College of Engineering, University of California at Berkeley, December 1972.
l Seed, H.
B. and Whitman,
R. V., 1970, " Design of Earth Retaining Structures for Dynamic Loads," Specialty Conference on Lateral Stress in the Ground and Design of Earth Retaining Structures, American Society of Civil Engineers, Soil Mechancs and Foundation Division.
Silver, M. L. and Seed, H.
B., 1971, " Volume Changes in Sands During Cyclic Loading," American Society of Civil Engineers, Journal of Soil Mechanics and Foundations Division, Vol. 97, No. SM9.
Bowles, J.
E.,
1977, Foendation Analysis and Design, Second Edition, j
McGraw-Hill Book Company, New York.
Delal, J.
S., 1975, Memorandum of Telephone Conversation between J.
S.
Delal, Seismic Consultant, United Engineers and Constructors, and Dr.
H.
B.
Seed, Professor, University of California, Berkeley, February 5, 1975.
1 i
- Richart, F.
E.; Woods, R. D.; and Hall, J.
R., J r., 1970, Vibrations of Soils and Foundations, Prentice-Hall, Inc., Englewood Clif fs, NJ.
SB 1 & 2 FSAR TABLE 2.5-12 (Sheet 1 of 2)
SUMMARY
OF RdCK PROPERTIES Property Rock Type (2)
Range of Values Average Value 1.
Permeability of Rock Mass (cm/sec) a.
Borehole water D, Q 0 to 7 x 10-3 (g) pressure tests (20 ft test zones) a.
Pumping test, D, Q 10-3 1 x 10-3 Boring F-5 (200 ft thickness of rock) 2.
Compression (P) Wave Velocity (ft/sec) a.
Seismic D, Q 13,000 - 16,000 (1)
D b.
Uphole and cross-D 16,500 - 18,500 (1) hole geophysical tests c.
Laboratory sonic
~By-Q-14,682 - 20 A50-
--l+rl-lG-t e s t s (no con fining 1
pressurc )
3.
Shear (S) Wave Velocity (ft/sec) a.
Uphole and cross-D 8,000 - 10,000 (1) hole geophysical tests 3
4.
Density (g/cm )
D, 'Q 2.63 - 3.01 2.80 5.
Unconfined Compressive D
6,000 - 34,000 18,300 Strength (psi)
Q 6,000 - 19,200 12,100 l1 Sjloo 6.
Modulus of Elasticity -
Initial Tangent Modulus 6
Ei (10 psi) a.
Uphole and cross-D 6.5 - 9.8 (1) hole geophysical 7---
tests
< l e ij s pecme As DQ I4,6 62 - 17,657 16,2(10 alu.aled cpecmcas D, Q l c, cico - 20,c;50 (8,670
~
..__. _... ~. -
TABLE 2.5 TYPES OF ENGINEERED BACKFILL BENEATH CATEGORY I STRUCTURES Page 1 of 3 Category I Structure Type of Engineered Backfill Allowable Maximum i
Between Bottom of Structure Bearing Bearing and Top of Sound Bedrock (l)
Pressure Pressure Fill Offsite Tunnel Concrete Borrow Cuttings tsf tsf UNIT NO. 1 Containment Structure x
60 12 Containment Enclosure Building x
60 36 Containment Enclosure Ventilation Area x
60 2.8 i
Control Building x
60 Diesel Generator Building x
60 Non-Essential Switchgear Room x
60 RHR Spray Equipment Vault x
60 Primary Auxiliary Building x
60 Fuel Storage Building x
60 Fuel Storage Facility Wall x
60 Condensate Storage Tank x
60 5.2 j
Emergency Feedwater Pumphouse x
60 14 Steam and Feedwater Pipe Chase (East) x 60 4.0
.ll '
Steam and Feedwater Pipe Chase (West) x 60 18 Pre-Action Valve Building x
60 4.5 Personnel Hatch Area x
60 Tank Farm Area x
60 Refueling Water Storage Tank x
60 Reactor Makeup Water Storage Tank x
60 i
e f
i I
i
TABLE 2.5 TYPES OF ENGINEERED BACKFILL BENEATH CATEGORY I STRUCTURES Page 2 of 3 Category I. Structure Type of Engineered Backfill Allowable Maximum Between Bottom of Structure Bearing Bearing and Top of Sound Bedrock (l)
Pressure Pressure Fill Offsite Tunnel Concrete Borrow
-Cuttings tsf tsf UNIT NO. 2 Containment Structure x
60 12 Containment Enclosure Building x
60 36 Containment Enclosure Ventilation Area x
60 2.8 Control Building x
60 Diesel Generator Building x
60 Non-Essential Switchgear Room x
60 RHR Spray Equipment Vault x
60 Primary Auxiliary Building x
60 Fuel Storage Building x
60 Fuel Storage Facility Wall x
60 Condensate Storage Tank x
60 5.2 Emergency Feedwater Pumphouse x
60 14 Steam and Feedwater' Pipe Chase (East) x 60 4.0 Steam and-Feedwater Pipe Chase (West) x 60 18
[1 Pre-Action Valve Building x
60 4.5 Personnel Hatch Area x
60 Tank Farm Area x-60 Refueling Water Storage Tank x
60 Reactor Makeup Water Storage Tank x
60 i
~
TABLE 2.5 TYPES OF ENGINEERED BACKFILL BENEATH CATEGORY I STRUCTURES Page 3 of 3 Category I Structure Type of Engineered Backfill Allowable Maximum Between Bottom of Structure Bearing Bearing and Top of Sound BedrockIl)
Pressure Pressure Fill Offsite Tunnel Concrete Borrow Cuttings tsf tsf OTHER STRUCTURES Circulating Water Pumphouse x
60 Service Water Pumphouse x
60 Electrical Control Room x
60 Intake Transition Structure x
60 Discharge Transition Structure x
60 Piping Tunnels x
60 Waste Processing Building x
60 Service Water Cooling Tower x
60 Safety-Related Electrical Manholes x(2) x(2) 2.5 0.5 Safety-Related Electrical Duct Banks x(3)
Safety-Related Service Water Pipes x(4)
NOTES: (1) Backfill concrete and sand-cement were not used as engineered backfill beneath the foun-dations of any seismic Category I structures.
(2) Offsite borrow was used beneath all safety-related electrical nanholes, except Ihnhole W19/20. The maximum thickness of offsite borrow beneath safety-related manholes is approximately 18 ft.
Manhole W19/20 is founded on tunnel cuttings with a few layers of offsite borrow included within the tunnel cuttings. The thickness of the combined tun-nel cuttings and offsite borrow beneath this manhole is 15.3 ft. (See Fig. 2. 5-42c. )
(3) The maximum thickness of offsite borrow beneath safety related electrical duct banks is approximately 18 ft.
(4) The thickness of offsite borrow beneath safety-related service water pipes is 15 ft or l1 less, except in the area between the Circulating Water / Service Water Pumphouse and the Intake / Discharge Transition Structures where the thickness of offsite borrow beneath the service water pipes is approximately 25 ft.
/
/
TABLE 2.5 Properties For Seismic Deformation Analysis of Revetment Property Revetment Offsite Borrow Glacial Filter Stone 90% or 95%
Till Cloth 7,
Compaction s
1.
Unit W'ight e
Saturated - below water 140 pcf 136 pcf 140 pcf Moist - above water 126 pcf
. 12 '6 pcf 2.
Shear Modulus Parameter, K (1) 170 55 110 2
,2 3.
Damping at low strain
, 6 in./in.)
0.5 0.5 0.5 l ' l' level (<
4.
Poisson's ratio, Saturated - below water 0.3 0.48 0.48 Above water table 0.3 0.3 0.3 5.
Friction angle 36*
34'
/
- 36*
32' (1) Parameter K2 used to compute shear modulus at low strain level,
(< 10-6 in./in.) with equation Gmax = 1000K2(~m)I/ where m is the, octahedral ef fective stress.
1
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. i 4.a m 414-
,t STATIC LOADING f l. +20 ft above MSL D
F
~7
\\y
<< ~ --
v n=
"r NOTATlON Water
\\
Imvel for
\\
Design H
e
\\
N
- Depth of wall below grade, f t.
x\\
\\
Y*
- Buoyant Unit weight, use 62.5 pcf for m
offsite borrow 7
Y
- Satursted Unit weight, use 125 stf
'2 3
- i for of fsite borrow Non-Rigid Wall R,y u y,N Eg g
y,
- Unit weight of water, use 62.5 pef Active Static Pressure 9
- Live load Surcharge = 500 per minimum Earth Water he to Pressure Pressure Surcharge q
- Fixed or Permanent Surcharge, psf F
(where appiscele]
R = Coef ficient of Active Easth Pressure, g
use Rg =0.30 K
- Coef ficient of Dynamic Earth Pressure, h
use Kh = 0.1g for SSE Kh = 0.10 for ces ETES EARTHQUAKE LOADING 1.
A non-rigid well is defined as a retaining well
[E1. +20 f t above MsL which is not supported at the top by floors, U
,,y, etc., and can deflect under earth pressure.
\\
Water 2.
Finished plant grade is +20 ft MSL.
Design N
groundwater level is 21. +20 ft MSL (refer to Level fo 8
Design N
Section 2.5.4.6).
~
3.
See Fig. 2.5-33 for lateral loads on rigid walls.
o_
__\\
\\
k Mon-Rigid RI" T"
I9 b9r he KY' As w
Ar Active Static Static Dynamic Dynamic Earth Water Component Component Pressure Pressure Pressure of of Increment Surcharge Surcharge Pressure Pressure Public Service Company of New Nampshire IATERAL IDADING DIAGRAMS SF.ABROOK STATION, UNITS 1 & 2 FOR NON-RIGID WALLS Final Safety Analysis Report F1 re 5-52 l
I STATIC LOADING g
y E1. +20 f t above MSL h g vp NOTATION r'-
\\
Invel for N = Depth of wall below grade, ft.
\\
- ~
l g
N y, =.uorant unit.ei,ht, u.e.2.s pef,or e-offsite borrow 3
l l
q, To
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+-
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H W
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- Live load Surcharge = $O0 psf minimum o
Earth Static Pressure Pressure a r ed or Pe nt Surcharge, psf Pressure Water Due to Due to 9r At-Rest Pressure Surcharge Cospection R, = Coef fic1ent of At-Rest Earth Pressure,
..e a,= o.s E = Coef ficient of Passive Earth Pressure.
P
,,,g
,3,3 p
D = Coef ficient of Dynamic Earth Pressure, R
use RD = 0.28 for SSE RD = 0.1$ for OBE EARTHQUAKE LOADING NOTES Et, +20 ft atore MSL U
1.
A rigid well is defined as a foundation wall h
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supported and effectively restrained by the
~'
~
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I O'
N
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Finished plast grade is e20 f t MSL.
Design e
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l M
W 3.
See Pig. 2.5-52 for lateral loads on non-E TsE
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pp g
Earth Static Static Dynamic Dynamic Pressure Water Component Component Pressure At-Rest Pressure of of Increment Surcharge Surcharge Pressure Pressure LATERAL IDADING DIAGRAMS Public Service Company of New Hampshire MR RIGID WW SEABROOK STATION, UNITS 1 & 2 rinal Safety Analysis Report l
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PUSLtc StnVICE COMPANY OF NEW MaasPSHing SOIL PROFILE 4 FINITE ELttstNT MEgn stAenooW STATION-UNITS 983 REVE T M ENT S FINAL SAFETY ANALSIS REPORT l Peeung 3.5,$s_
SHEAR MODU LU S, Gman( p s I)
(of sMr stroin f alOro in/in)
O 50,000 100.000 150,000 200,000 0
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5000 NOTES: 1. See FSAR text, Subsection 2.5.4.7 for description of method used to backfigure G from plate load tests.
- 2. Curves for G vs b were generated from the plate load test dafa*using"the relationship
[ 0,, /c 'with G and a being the G
=G p$$$Nloab*0Estv51ueE.
- 3. Values of G for shear strain levels greater than 10-6 in./in. can be obtained using the average modulus reduction curve for sands presented in Seed and Idriss (1970).
- 4. Values of K for use in the equation G
=1000 K2 (b )
2 max m
are shown next to each curve.
PUBLIC SERVICE COMPANY OF NEW llAMPSilIRE SIIEAR MODULUS AT IDW STRAIN LEVEIS SEABROOK STATION - UNITS 1 & 2 FOR OFFSITE BORROW AND 'IUNNEL CUTTINGS FINAL SAFETY ANALYSIS REPORT l Picture 2.5-58 m