ML19344A767
| ML19344A767 | |
| Person / Time | |
|---|---|
| Site: | La Crosse File:Dairyland Power Cooperative icon.png |
| Issue date: | 12/31/1978 |
| From: | Bieganousky W, Marcuson W ARMY, DEPT. OF, CORPS OF ENGINEERS |
| To: | |
| References | |
| TASK-02-04, TASK-2-4, TASK-RR NUDOCS 8008220173 | |
| Download: ML19344A767 (38) | |
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LIQUEFACTION ANALYSIS FOR LACROSS NUCLEAR PO'.GR STATION By W. F. Marcuson, III And W. A. Bieganousky s.
December 1978 Prepared for U. S. Nuclear Regulatory Commission Washington, D. C.
Prepared by Geotechnical Laboratory
.U. S. Army Engineer Waterways Experiment Station
~Vicksburg, Mississippi CiO_b8220lf
/
- s 6
LIQUEFACTION ANALYSIS FOR LACROSS NUCLEAR POWER STATION Introduction
Background
1.
The United States Nuclear Regulatory Cormission (NRC) requested that the Waterways Experiment Station (WES) review certain foundation conditions at the LaCross Nuclear Station, an operating nuclear power plant. This plant, which is located near LaCross, Wisconsin, is among the oldest in the country and was put into operation before the present Site Analysis Report review system came into effect. This report docu-ments WES' review of the LaCross Nuclear Station. Specifically, the question examined was the earthquake safety of the pile foundation which supArts the containment vessel. The piles are driven through lov to medium relative density sands and terminate in it dense sand layer approxi-mately 28 ft above bedrock.
Scope of work 2.
The investigation of the foundation at the LaCross Nuclear.
-Station included the following:
Review of Chapter 3, Soil Engineering Properties contained a.
in the Application for Operating License for the Lacross Boiling Water Reactor by Dairyland Power Cooperative including portions of Appendix A, entitled " Field Exploration and Laboratory Tests,"1 and associated design
- drawings, b.
The performance of a liquefaction analysis using the Seed-Idriss Simplified Procedure assuming an earthquake with a-peak accelera-tion of 0.12 and 0.2 g, respectively.
1 L.
r 9
c.
The performance of a liquefaction analysis using Seed' -
empirical method assuming both the 0.12 and 0.2 g earthquakes snd co= pari-son with a " rule of thumb" based on the Japanese experience at Niigata in 1964.
3 The objective of this study was to evaluate to the degree possible with the data available from prior field and laboratory studies by others, the seismic stability of the pile foundation which supports the contain-ment vessel at the LaCross Nuclear Station.
Review of Previous Work h.
As stated earlier, portions of Reference 1 vere reviewed to determine the soil profile under the contain=ent vessel, including soil properties. Logs of Borings B-3 and 3 h drilled _ by Raymond International in July 1962 and Borings DM-1 and DM-3, drilled under Dames & Moore's supervision in 1973, vere reviewed. Figure 1 shovs an idealized soil profile in the vicinity of the reactor building. The ground surface is at elevation (el) 636 ft mean sea level (ms1). The groundwater table was assumed at a depth of 13 ft.
Top of bedrock is located at a depth of about 133 ft.
- 5.. During construction, the soil was excavated to el 615 ft ms1, and piles were driven below the reactor containment vessel. These piles terminated at el 535 ft.
The piles vere 50-ton cast in-place concrete piles with a tip diameter of 8 in., a butt diameter of 12 in, and an outer shell of 7-guage steel monotube, and were driven approxtaately 13 1/2 ft on centers.
No mention was made of any internal rainforcing steel in the piles in the plans and reports provided to WES. A total of 2
approximately 230 piles vere driven. The data provided indicate that the hammer used was a McKierman-Terry C-5 double-acting hn=mer with a rate d striking energy of 16,000 ft-lb per blow. The piles were driven to at least a resistance of 6 blows per in. for the final 2 to 3 in.
The number of blows in the last foot actually ranged from 75 to 330.
6.
The soil below the reactor building (el 615) consisted of a fine to medium sand with occasional zones of clayey silt, coarse sand, and fine gravel, down to an elevation of approximately 535 ft.
At el 535 ft, a 10-ft-thick fine to medium sand, with fine to medium gravels is encountered. Below this gravelly sandy layer, is an 18-f t-0.ick layer of sand which immediately overlies the bedrock.
7 Also shown on Figure 1 are the average blow counts, water content, dry and vet density, and shear-vave velocities for the six layers in the idealized soil profile. ' The reader is cautioned that the blev counts may or may not be Standard Penetration Test (SPT) N values.
It was not explicitly stated in the available source of information how the penetration tests were conducted nor were they called " Standard Penetration Test" results. The values shown on Figure 1 are considered approximate average values for the layer. Both the blows per foot and the dry density values were obtained from an evaluation of the boring
- log data in Reference 1.
The vet densities and shear-vave velocities which are shown on Figure 1 are estimates based on WES' experience and data' presented in Reference 1.
8.
Figure 2 is a plot of blows per foot versus depth. The data obtained in Boring B-3 are believed to have been obtained prior to pile 3
4 r
e
.w-
~ -
s s
~ driving. The data obtained in Borings DM-1 d DM-2 are believed to have been obtained after pile driving. The reader is cautioned that a 1-to-1 comparison is impossible because the data obtained in Boring B-3 vere obtained using a 2-in. split-spoon sampler while the data obtained in Borings LM-1 and DM-3 vele obtained using the Dames & Moore sampler which is 3-1/h in. in diameter. Figure 3 is the dry density information obtained from samples obtained from Borings DM-1 and DM-3.
9 Figure h is a plot of overburden pressure versus depth for the site. As stated previously, the water table was assumed at a depth of 13 ft.
Below this depth both total and effective overburden pressures are shown.
10.
The blow count values were assu=ed to be Standard Penetration Test N values and were used to compute relative density from the following I
equation:
I 11/2 l
D = 11.7 + 0 76 222(N) + 1600 - 53(6,) - 50(c )2 (1) r where D = Relative density N = Standard Penetration Test N values 6, = Effective overburden pressure in psi C = The coefficient of uniformity u
Using equation 1, the relative density of the top 105 ft is predicted to I
be between 50 and 60 percent.
l 11.
Review of the available data indicate that the material has the minimum density of about 100 per and a maximum d nsity of about 120 pef.
These tests vert run on bulk samples obtained by combining representative materials encountered at the site. WES' experience indicates that when b
9
\\
. materials are combined the maximum and minimum density generally increase.
This increase can be as much.as 8 pef. Consequently, WES believes that a minimum density of 92 per and a maximum-density of 112 per may be =cre realistic for the in situ material. The relative densities predicted by equation 1 appear to be more nearly the same as those which would be obtained using the WES' maximum and minimum density estimates and the in situ dry unit weights given in Figure 3 An analysis of pile geometry records supplied by NRC suggest an average density increase of approxi-
~
mately 1 per due to pile driving. This is based on the reduction of void ratio which would occur assuming the soil displaced by the pile vent
. entirely into taking up the voids of the adjacent soil. This assumes no soil heave and does not account for any densification due to vibrations during driving. An increase of 1 per is not significant and is believei not to contribute substantially to the stability of the soils. Records of ground surface movement during pile driving vere sought but no such information was provided. While it is possible that more densification may have occurred, there vere no data made available to WES Vnich would support this hypothesis.
12.
Dames & Moore determined the liquefaction potential of the subsurface soil'by performing 11 stress-controlled dynamic triarial compression tests on representative samples of the material considered to be potentially susceptible to liquefaction during the SSE.
Eight of the samples were reconstituted to approximately what Dames & Moore beli'eved to'be the in situ density. 'In addition, three reconstituted sa=ples were
~
tested at slightly greater densities than the average in situ value (as
. determined'by Dames- & Moore)-in order to examine the influence of density
~
variation on liquefaction potential. These tests were run at confining 5
u.
. ~ - -
i pressures of 1000 and 2000 psf and the results of these triaxial tests are shown on Figures 5 and 6.
The data points identified by 10h, 111, 1,
and 112 (the dry density in pef) on Figure 5 were for specimens consoli-dated to an effective confining pressure oi J 30 psf. The data identified by 106 and 105 (dry density in pef) on Figure 6 vere obtained by consoli-dating the specimens to an effective confining pressure of 2000 psf.
Dynamic Streneth of the Soil 13 In order to evaluate the liquefaction potential of the soil in question, the laboratory cyclic triaxial test results obtained by Dames & Moore vere used. Figures 5 and 6 are reproductions of Danes &
Moore's test results. The as-tested dry densities of the remolded speci-mens made from material taken from Boring 5 at a depth of about 8-1/2 ft are shown next to the data points for these tests on Figure 5 A curve has been drawn through the data points for a dry density of about 111 pet.
As stated previously, WES believes that this material is at an in situ I
dry density of about 102 pet.
Consequently, a curve more or less parallel to the 111-1b curve was drawn through a data noint at 10h pef. This curve (marked yd = 102 pef) was used to evaluate the soil strength. On Figure 6, the remolded soil specimens taken from Boring 3 at a depth of 35 5 ft have been used. Adjacent to each data point, the as-tasted density is. listed. A curve is drawn through these data-for a density of about 106 pet. A curve more or less parallel to this curve has been drawn and labeled y = 100 pef, because WS beMeves %e in snu denshy d
of the material at a depth of about 38.5 is 100 pet. This curve was also used to evaluate the dynamic soil strength.
6
t 1h. As vill be mentioned later on in this report. the nu:.0er of equivalent cycles of load for SSE vas chosen to be its (see disc =sion of design earthquake on page 8 ).
Figures 5 and 6 vere entered d 10 cycles and the stress ratio required to cause 10 percent double-amplitude strain was determined. This stress ratio was multiplied by 1000 e.nd 2000 psf as appropriate to determine the superimposed dynamic shear strength. These values are plotted on Figure 7 A line was drawn through these data points and labeled isotropic laboratory data.
Is atropically consolidated cyclic triaxial test data must be corrected by a correction factor, C
, to represent field conditions. This correction factor is based on the comparison of cyclic triaxial test results to cyclic simple shear and SHAKE table test results (References 2, h, and 5). For this investigation, a C f 0 57 was used. Also shown on Figure 7 is a r
curve labeled field conditions. The ordinate of this curve 's 0 57 times the ordinate for the curve labeled isotropic laboratory data. The field condition curve was used to evaluate the dynamic shear strength of the soil during this investigation.
Seed-Idriss Simplified Procedure 15 In order to evaluate the liquefaction potential of a site, the cyclic stresses generated by the earthquake must be determined.
Reference 2 suggests that the average shear stress, T,y,
, generated by an earthquake can be determined by the formula:
T
^
X rd (2) ave max 7
where y = the total unit weight of the soil H = depth from the ground surface to the point in question in feet g = acceleration of gravity A,x = the peak acceleration at the ground surface generated by the earthquake in the same system of units as g r = a rigidity factor d
The constant 0.65 is a factor which corrects the maximum shear stress to an equivalent sinusoidal shear stress.
- 16. Using equation 2, T,y, can be determined for any depth in the soil profile. The r factor used in this analysis is shown by the d
curve marked " analysis" on Figure 8.
Earthouake parameters 17 In order to conduct this analysis, the maximum acceleration generated by the design earthquake and the nu=ber of equivalent cycles of stress are required. The maximum acceleration was specified by the NRC as 0.12 and 0.2 g.
Review of the geological and seismological studies conducted at Lacross predict that an earthquake of Modified Mercalli Intensity VIII in the epicentral region has occurred on the Keveenav fault.
For analysis purposes, an earthquake with an intensity one unit greater than the largest recorded intensity was assumed. Thus,
.using a Modified Mercalli Intensity of IX and using the intensity-magnitude relationships shown on Figure 9 (Reference 6), a magnitude 6.6 earthquake is postulated. Figure 10 is a plot of number of equivalent cycles versus magnitude which was developed by Seed.
This plot was entered and 10 equivalent cycles, which are essentially an upper bound 8
1 I
to the data in the plot, were assumed appropriate. For the design earthquakes (SSEa, and SSEb) 10 cycles and peak ground surfece accelera-tion values of 0.12 and 0.2 g were used, respectively.
Analysis
- 18. Because of the high N values obtained in the dense sand layer at a depth.of about 105 to 115 ft, this zone is predicted to remain stable even under the so-called design earthquakes. There are no other data available from this layer. If one assumes that the dynamic soil strength 4
of this zone was the same as that, judged appropriate for the upper materials by WES on Figure 7, then liquefaction might be predicted.
WES does not believe this vill happen.
~
SSE,.0.12 g_
19 Table 1 presents the information needed to calculate the average shear' stress (t,y,) for the so'il profile using the Seed-Idriss method.
Also. listed on Table 1 is the effective overburden pressure (6 ) needed to enter Figure 7 to determine the available soil streng;h.
Both the values of dynamic shear strength and dynamic shear stren r_re listed as a function of depth on Table 2.
The factor of safety against 10 percent double-amplitude strain has been defined as the dynamic shear strength divided by the average dynamic shear stress. This factor of safety _is also listed on Table 2.
It should be noted that the factor
-of safety below a depth of 35 ft -(depth of excavation) varies from 0 99 to 1.15 and is below 1.1 to a depth of 100 ft.
Factors of safety less than 1 were predicted in the soil between a depth of 35 and h5 ft.
It should be emphasized that the state of knowledge is not-adequately refined and the assumptions required to carry out the analysis, given 9
O
~
s l-the limitations in'the existir.g data base, are such that it is not believed justified to call this
'te safe.even though factors of saf ety marginally greater tiian 1 vere celculated. Hosever, this analysis does not prove the site unsafe under this acceleration as it is possible that had more extensive data and more thorough docu=entation been available, the judgments concerning the 2n-situ density and cyclic shear strength would have been different.
I SSE, 0.2 g i
at the LaCross site 20.
The information needed to calculate T,y, for the SSE vith 0 2 g peak acceleration is given on Table 3 Also shown on Table 3 is the effective overburden pressure which was used to enter Figure 7 to determine the shear strength of the soil. Table h presents the average shear stress and fynamic shear strength as a function of depth for this SSE. Also she.:n on Table h are factors of safety (as treviously' defined) for this ':SE.
As can be seen, the factors of safety below a depth of 35 ft ' ary from 0 59 to 0.66.
clearly, this indicates v
failure, as failure is defined in this report. A doubling of the cyclic strength over that shown by the authors would be required to produce a factor of safety of 1.25 This level is often considered reasonable for safety in the type of analysis performed herein.
Etuirical Liouefaction Analysis 21.
Empirical data in the form of stress ratio and corrected U values for sites that have and have not liquefied during past earthquakes, have been developed'and plotted on Figure 11 (Reference 5). On Figure 11, the Standard Fenetration Test U values have been corrected to an effective 10 4
n-e - _ +.,
overburden pressure of 1 tsf. As a second means to evaluate the lique-faction potential at the LaCross Site, the average blevs per foot as shovr.
on Figure 1, were plotted against the stress ratio and compared to the data shovn' en Figure 11.
The stress ratio T,y, divided by 3 vb.ere 0
o, is equal to the. effective overburden pressure is also tabulated on Tables 1 and 3 for the various depths in question.
22.
The blevs per root were assumed to be SPT U values and vere corrected to an overburden pressure of 1 tsf by the formula:
N
=C y
N*
vhere N = Standard Penetration Test penetration resistance value =easured in the field N = Standard Penetration Test N values corrected to an overburden y
pressure of 1 tsf C = Correction factor N
- 23. C was determined from Figure 12 (References 5 and 7).
N Figure 12 van extrapolated back to zero using data in Peck, Hanson, and Thornburn.I Values of C are a s shown on Tames 1 and 3 N
1 These values have been superimposed on Figure 11.
Most of the values show that liquefaction should not occur; however, liquefaction is predicted from 25 to 35 ft.
2h. A similar analysis was conducted assuming a SSE vith a peak acceleration of 0.2 g.
The stress ratios, CN, N, and N
, as a y
function of depth are tabulated on Table 3 These values are shown on Figure 13 These data indicate that liquefaction is possible if a SSE producing 0.2 g at the ground surface occurs. The data points to the far 11 '
right which indicate safe conditione are for the 10-ft-thich sand and gravel layer at a depth of about 105 ft where the piles end.
25 Based on the Japanese experience in the 16 June 196h Niigata earthquake, a rule of thumb hes been developed. This rule states that in order to be safe against liquefaction, the Standard Penetration Test N value should be at least two times the depth ih meters.
A line has been drawn on Figure 2 indicating what the N value should be if it vere greater than two times the depth in meters. Note that a large majority of the blows per foot fall on the unsafe side of this line.
This is particularly important in this case since the peak acceleration at Niigata was approximately 0.16 g.
Summary and Conclusions
- 26. The liquefaction potential of the Lacross Site was evaluated for two earthquakes; namely, a SSE vith a peak acceleration at the ground surface of 0.12 g, and a SSE vith a peak acceleration at the ground surface of 0.2 g.
The analysis was made by two methods; namely, the Seed-Idriss Simplified Procedure and an e=pirical procedure. Figure lh is a summary plot of the dynamic shear stress as a function of depth for a peak acceleration of 0.12 g.
Superimposed on this plot is a cyclic i
strength of the material assuming 10 equivalent cycles of losding. Note that liquefaction is predicted by the Seed-Idriss calculations between a depth of 32 and kB ft and liquefaction is predicted by the empirical methods based on a depth of between 2h and 35 ft.
Japanese experience at Niigata, Japan, also indicates that liquefaction vould be predicted below a depth cf 15 ft.
As indicated in Appendix A, the piles could 12
}.
e s
rapport concentric static loads even after a loss of lateral support down to a depth of kB ft.
However, it vould require dynamic structural analysis beyond the scope of this study to judge whether they would have sufficient bending resistance to withstand the transient eccentricity of the static vertical loading and the transient horizontal loads caused by the seismic excitation. In view of the fact that there appears to be no reinforcing bars in the top one-third of the pile 9 (as called for in some seismic design 0), it is probable that the available bending resistance is modest.
codes 27 Figure 15 is a su= mary plot of the dynamic shear stress as a function of depth for the soil profile assuming a peak acceleration of 0.2 g.
Also shown en this plot is the dynamic shear strength of the soil assuming 10 equivalent cycles of loading. The Seed-Idriss Simplified Procedure predicts liquefaction below a depth of 25 ft.
The enpirical method predicts liquefaction between a' depth of '25 and 60 ft and betseen a depth of 85 and 105 ft.
If lateral support is lost in the depth ranges predicted by either method, the piles would be in danger of buckling failures as indicated in Appendix A.
- 28. Based on the judgments concerning the density and strength data and on analysis as presented herein, the 'soilt below the reactor at the LaCross-Site are predicted to strain badly if a SSE which produces 0.12 g at the surface of the soil occurs. The soils beneath the reactor vessel _at the LaCross Site are predicted to experience excessive strains and liquefaction if the SSE vith a peak acceleration at the ground surface of 0.2 g occurs. Because of the limitations in the current state of knowledge concerning liquefaction and because of the li=ited data atallable for use in this analysis, WES cannot conclude that the reactor 13 t
b
e vessel foundation is safe if the 0.12 g SSE occurs and concludes that the reactor vessel foundation is unsa% if the 0.2 g SSI occurs, s
l l
lh i
f J
-o References 1.
Dairyland Power Cooperative, Application for operating License for the LaCross Boiling Water Reactor.
2.
Seed, H. B. and Idriss, I. M., "SimpT.ified Procedure for Evaluating Soil Liquefaction Potential," Journr.1 of the Soil Mechanics and Foundations Division, ASCE, Vol. 97, No. SM9, Proceedings Paper 8371, September 1971, pp. 1249-1273 3.
Marcuson, W.
F., III and Bieganousky, W. A., "SPT and Relative Density in Coarse Sands," Journal of the Geotechnical Encineering Division, ASCE, Vol. 103, No. GT11, Proceedings Paper 13350, November 1977, pp. 1295-1309 h.
Seed, H. B. and Peacock, W.
H., " Test Procedures for Measuring Soil Liquefaction Characteristics," Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 97, No. SM8, Proceedings Paper 8330, August 1971, pp.1099-1119 5
Seed, H. B., " Evaluation of Liquefaction Effects on Level Ground During Earthquakes," Preprint 2752, Liouefaction Problems in Gt technical Fngineerinc, ACCE National Convention and Exposition, Philadelphia, PA, 27 September - 1 October 1976, pp.1-104.
6.
Krinitzsky, E. L. and Chang, F. K., " State-of-the-Art for Assessing Earthquake Hazards in the United States; Earthquake Intensity and the Selection of Ground Motions for Seismic Design," Miscellaneous Paper S-73-1, Report h, September 1975, U. S. Army Engineer Waterways Experiment Station, CE, Vicksburg, MS.
7 Peck, R. B., Hanson, W. E., and Thornburn, T. H., Foundation Engi-neering, New York, Wiley, 2nd Edition, p. 312.
8.
Ohashi, M., Iwasaki, T., Tatsuoka, F., and Tokida, K., "A Practical Procedure for Assessing Ea*thquake-Induced Liquefaction of Sandy Deposits," Publics Works Research Institute Ministry of Construction, Tenth Joint Meeting U.S.-Japan Panel on Wind and Seismic Effects, UJNR, Washington, DC, 1978.
9
" Containment Vessel Pile Driving Operations for 50 MWe Boiling Water Reactor at Genoa, Wisconsin," February 1963, Report No. SL-2003, Sargent and Lindy Engineers, Chicago, IL.
10.
" Tentative Provision for the Development Seismic Regulation for Buildings", Applied Technology Command, Publication ATC 3-06, July 1978,1J. S. Government Printing Office, Washington, DC, (P. 75, Section 7 5.3).
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%D
.8 90 e4 40 e4 %D e4 ed e.
%D e4 it fre
.* tr%
5 4.
eg m M a
M N e4 O Os
/..
.eO
..oo.. o. %3N.
O.
40 to.
- e. er.s t-w + t-O N ers t-N ers t-N.
O er%
t-o
.s..,.O e.
sv. e.
N.
tr.e O.s ar.t f=. m. e.
- e. o.
%D. N.
O.s er.s es.
p.
e4 e
A g
f) o e4 e4 e4 N M m.e
.e er% w %O t.
t= e Os Os o
- 8%s N
%.* e
%D g
.e4 f'4
.ers e4 4 e4 et r9 e et e4 4 e4
.=
tr.s m m o. e. e. m. m. e.
e.
o
- m.. m. m.. m. o. m. o. m. m. m. m. n. m o. e.
m.
en e
e e
e e
As &
N 3=
e4..e. 9=
Of t*
N fn.N
.P=
.N f%f P*
f*
N P=
N >= N P=
N t=
Cw 8*
N to e.
P*
y 6-e4 c
e4 tw N M e
m trs
%D P=
r= = = Os On o o es ce..N N
e=1 s.a e4 e4 e4 c
e e4 e4 4
.s. > 'N ma m %o,= e Os o e.
m mm,=
o N m.e N.
e
.O
%o
- v. e m.
a-a.
e4 e4 e4 r<
e4 N N N << N a cw N N 9 sp
.e n E Era e.=s.
,...es s
e g
e g.
s
' Table 2 RESULTS OF DYIIA!GC A!!ALYSIS 0.12 g SSE A"'#"E' Dynamic hear Shear Factor ress
. Depth Strength op T
ft psf ave Safety 25 50 21 2.38 75 102 63 1.62 11 5 1h5 96 1 51 1h.0 166 116 1.h3 17 5 185 1h6 1.27 22 5 212 188 1.13 27 5 239 229 1.0h 32.5 293 296 0 99 42 5 320 321 0 99 h7 5 3h7 3h2 1.02 52.5 3Th 365 1.02 57 5 h00 385 1.04 62 5 h29 kik.
1.0h 67 5 h60 h38 1.05 t
72 5 h90 h65 1.05 77 5 520 h93 1.05
~~~
82 5 551 519 1.06
' fn. 5 581 Shh 1.07 92 5 611 568 1.08 97.5 6h2 591 1.09 102 5 672 613 1.10 l
s Table 3 IACROSS SITE I.IQUEFACTICN CALCJI.ATIONS FOR A PLUt ACCE:.ERATION CT o.20 C Effeettre f.ffeetive Average Stress overburdes overburden octahedral octahedral pelative Maa t.em Chesr Petto Corrected Pressure Preesco Stress Siress Density Accelerattoa
'Strese
-v Correction SP7 "I"*
E A
T
'M14 depth Depth
'o
'o
'oct
'oet d
nax g "
B aloe age Sublatver ft Jsf her ksr har
_t 5m o.4 A,,,
y,7 e,
N y
g i
25 0.273 0.273 0.182 ~
0.182 1.0 0.20 0.13 0.036 c.132 1.h8 9
13 2
7.5 c.818 0.818 0 5k5 0.5k5 0.99 0.20 0.13 0.105 0.128 1.28 9
12 3
11.5 1.25k 1.25%
0.836 0.836 0 98 0.20 0.13 0.160 0.128 1.18 9
11 1%.o 1 535 1.kT3 1.023 0.932 0 97 0.23 0.13 0.194 0.132 1.11 9
10 5
17.5 1.9k8 1.667 1.299 1.111 0.96 0.20 0.13 0.2h3 0.1k6 1.03 8
8 c
6 22.5 2.538 1.945 1.692 1.297 o.95 c.20 0.13 n.313 0.161 0 96 8
8-7 27 5 3.128 2.223 2.085 1.%B2 0.9h 0.20 0.13 0.382 0.1T2 0 95 8
8
'8
.32.5 3.718 2.501 2.kT9 1.667 0 91 0.20 0.13 0.%ko 0.176 c.90 8
1 9
37.5 k.308 2.779 2.372 1.653 c.85 c.20 0.13.
,0.k93 0.177 0.85 16 lb 10 b2 5 b.898 3.057 3.265 2.035 0.8h 0.20 0.13 0.535 0.175 0.81 16 13 11
%7.5 5.688 3.335 3.659 2.223 0.60 0.20 0.13 0 571 0.171 0 77 16 12 12 52 5 6.078 3.613 n.052 2.to?
o.77 0.20 0.13 0.608 0.168 c.Th 16 12 13 57 5 6.668 3.891
%.nh5 2.59%
0 7%.
o.20 0.13 0.641 0.165 0.71 16 11 lb 62 5 7.276 b.187 n.851 2.791 0.73 0.20 0.13 0.690 0.165 0.68 25 17 15 67.5 T.901 b.500 5 267 3 000 0.71 0.20 c.13 0.729 0.162 0.65 25 16 16 72.5 8 526 n.813 5.68k 3.209 0.70 0.20 0.13 0.776 0.161 0.62 25 16 17 77 5 9 151 5 126 6.101 3.k17 '
o.69 0.20 0.13 0.821 0.160 o.60 25 15 18 82.5 9 776 5.%39 6.517 3.626 0.68 d.20 0.13 0.66%
0.159 0 58 25 15 19 87.5 lo.bol 5 752 6.93b
,; an o.67 c.20 0.13 o.905 0.157 0.55 25 lb 20 92 5 11.026,
6.065 7 351 b.0%)
0.66 0.20 0.13 a 0'i.
o.156 c.54 25 14 21 97.5 11.651 6.378 7.767 b.252 0.05 0.20 0.13 0 964 0.15h 0 53 25 13 6.691 8.15b k.L61 c.6b r 30 0.13 1.021 0.152 0.31 25 13 22 102.5 12.276 23 107 5 12.926 7.029 8.617 b.686 0.63 0.2L 0.13 1.058 c.151 0 50
]
24 112.5 13.601 7.392 9.067 6.928 0.62.
o.20 0.13 1.097 c.148 0.L8 25 117.5 14.251 7.730 9 501 5 153 0.61 0.20 0.13 1.130 0.1L6 0.47 60 28 26 122.$
14.8T6 8.043 9 917 5.362 0.61 0.20 0.13 1.180 0.147 c.h6 60 28 27 127.5 15.501 8.356 10.334 5.5'1 0.60 0.20 0.13 1.?o9 0.1h5 0.h5 60 27 28 131 5 16.001 8.606 10.667 5 737, 0.60
,0.20 0.13 1.248 0.1%5 0.kk 60 25
Table h DYNAMIC ANALYSIS FOR o.20 g SSE Average Shear Factor
. Depth Stress of T
- ft per ave Safety 2.5 50 30 1.39 75 102 105 0 97 11 5 lbs 160 0 91
^
1h.0 166 19h 0.86 17.5 185 2h3 0 76 22 5 212 313 0.68 27.5 239 382 0.63 32 5 266 hho 0.60 L
37 5 293 h93 0.59 42 5 320 535 0.60 47.5 3h7 571 0.61 52 5 37h 608 0.61 57.5
~400 6hl o.62 62 5 h29 690 0.62 67 5 460 729 0.63 72 5 k90 776 0.63 77 5-520 821 0.63 82.5 551 86h o.64 87.5 581 905 0.64 92.'5
-611 9h6 0.65 97 5 6h2 984 0.65 102.5 672 1021 0.66-e 8
4 5
r y
v v
,,ry.,
,,,sm.
,y--
..y%
~---,,
9 i
e ge n
nh Q
9 u
nd a
o W
t~~
~t 5
h' t
t g
,~
w
~
bo" o
o 9
o o
o
-)
y Y
b d
s s
s s
o a
N w
,e
'a 4
t u
N,.,,,
O D
4 sS s.
3 eG U
L1
~
+
m t
s n
n s
R 3
w s
y i'
't 3
b h
g 4
F e
=
4 U
ai y
6t 4
4 n
,q -
T
.3
.h
< d o
q v
A g
v g
y 4
Y 3*
4 4
w'
' d e
e s
t ow b
4
=
99 3'4 R.
1 4x ax s
t
..w 4
=
E 4V
.kV iy e
]
Y.*
d
(
4 T
e t
1, 3 3 41 es k
i.A a
2
-v A
n.
=
g g
p 4
esd d4 9
4' 4
m%
s s
s e
e4 0
G G
4 e
s s
e 1
4
'8 e
a k
k 5
~
v s
k 3
8 l
G g
~
s U
I \\
I
$,N l I.. k...
t
.a ai g
^
deastsaaeaaeaaaa c
v S
t BLOWS /FT o
20 ho 60 80 0
\\
D, 3
gg.-
\\
y 0
-.4 O
BORING DM-3 y
..1d.
A 20 0
0 BORING DM-1 5
(D s
A O
-6 O
A BORING B-3 i"~~
[
ho O
g 3 O O
O 6
N = 2 X DEPTH (M)
O l
60 0
E a
i 3
O m
r O
[
b 80 OA a
O A
O O
E i;
a o
100 6
O O 230 k
O O 210+
j D
0 200+
'i O
O l
120 0
0 118 T---
l ~f,G dO O 116 j. 11 AD 0 301
.g-la' 2'
,3
[ _._
Figure 2. -Blov Counts Presented as a Function of Depth l
i
Ni DEY DENSITY (PCF) g 60 80 100 120 30 0
I bO
{'
.7 O
O
,QE BORING DM-1 J'
20 O
O A
l 0
0 BORING DM-3 40 O
0
(
60 O
n
~
- D m
O 80 O
0 100 O
M 120 0
1 O
f[I
[)A*
p
[' fly L__
Figure 3.
Dry Density as a Function of Depth 9
i
9 i
t OVERBURDEN PRESSURE (KSF)
I r
0 2
h 6
8 10 12 IL 16 18 E
O
, l'[
". g'
]f..,I 10 p
,1) f;t' 20 a
30 t
40 50 n
U i
p 3
33 5
60 s
2 s2 70 i
l' 80 i
90 EFFECTIVE l
sTaEcS l
100 I
t 110 l
L 120 N1&
lb
- d
?h
.q
$l
..\\m Fi5ure 4.
Overburden Pressure Versus Depth
~
e t
~rg-r'
+
~
1
- 0. 6 -
- - 204tNG 5. OtPTM - 6.5 FT.* AttAfivt Otm1ITY - 15%
111
- sass >: 5. etFia - 8.5 Fi. RnATivt OttilTY - 60::.65T e
a soaisc 5. oterer - 18.5 Ft. attativt ettsliv - icz.jsx 0*5 O 20Alkt. 3. Ctrill - 38.5 FT., attaisit etastif - 3cx.,35%
N
\\
0.h
,112 10h
-e-8 le
\\
a 4
g 3
\\
03 x
m o
a N
\\
l g
N 0.2 o
N o--e.
e Y a 102 FCF a
0.1 I
t I
I I
I 8 I I II I
I I I I IIf f
~~'II f ~t Y t T 0
1 10 100 1000 NUMBER OF CYCLES REQUIRED TO CAUSE LIQUEFACTION, Y Figure 5 Cyclic Shear Strength Versus, Number of Cycles to Liquefaction Based on Dames and Moore Laboratory Data, cr = 1000 psf e
s T
O.6 20RitC $. CCPTM - 8.5 FT., attativt etssify. 6ct 65T RttATavt OtkllTT. 25%
o 33Rik: $. CtPTH. 8. 5 FT.,
O o
A BCRING $, Ottfet. 80.5 ti., stLATlYC Otx58TY. 3C2 35%
0 " 3 ' "" # - 3 8 5 "* * "** '" **"5 '" ' 3 **' 3 5 0.5
- 0. !:
o o
A oIto' O
a a
4 o
0.3 U;(
106 4-m 106 l
N 0.2 N
n N
t w
N g
y, ::: 100 rCr 0.1 0
I
- e i1 i
e i e e i
r i
I t 1it 1
10 100 1000 N HBER OF CYCLES REQUIRED TO CAUSE LIQUEFACTION, N Figure 6.
Cyclic Shear Strength Versus Number of Cycles to Liquefaction Based on Damen 2000 psf and Moore Laboratory Data, o =
v C
e
E E
S S
M h
R e8 8
8
-3 N
8 8m e
-g m
e a
h 8
h O
s s
3 a
n 2
m eW
\\
8 E
3 2
N co g
8 e
w
,E o
8 [
O w
d g
O l
E g
T R
8 h
g 3
0 g
t:
g 8.E i
en O *<:
o 8g o-8 4
'u
\\
\\
FoC.
t i
f I
000'tt 000'E 000'B 000'I O
(dSd) MDIGES hlV2HS i
DIICfRIG G2S0dNIII2d.nS a
f v
s
= (Tmax)d rd (Tmax)-
n t
0 0.2 0.h 0.6 0.8 1.0 8
I i
i
.i AVERAGE 20 VALUES
,1 e
l-vs o
.i.
,: e1;^
- r
,?
ho
, _.1l') Y
,/.,-
_ ~,. ; ;$ ',
_4 ze.
.v.,
{
RANGE FOR DITFEREITT
,..' p*?..'.' [
^
SOIL FROFILES
- .~
v G
fo
.; ?
1 x
s
-ae.....4 a
.7 e.
,c.
p
';--i ;;',s~.. /. '. i a
.hg.**
f*,
r ),. "... : /
.: s
.. p. ;
?
~.
80
~
' -.,; (l}.,r.s. '+f.,..'n4'.&..v, :' *.
.% rt
~
=
4
. 3 p.-y
.:. - r -.
..c
. J.
.v.p,:,.
.,,.s...
. c.:.t"
.p
.e Ia.. i f ' : r, t."~ '.'
USED IN ZiIS j,
,s,
..f... : r '/'
ANALYSIS l
120.
I f
Figure 8.
Rigiuity Factor Versus Dep+.h (after Reference 2) 4
,e c
v
9 l
ALA$K A (1964)
!0.33 SAN FRANCISCO tI906) e ALASKA
/
7 I(19 p F s s. "
i/
e 0
,s sG y
a cs o
s E/
c5 /
o
\\\\
~ -
dl I E E
i,/
n i
o u
o u
Il
'I M
n i
43
/
E6
<,/
3
- '- H A.
F
/
l[/
l
- / l LEGEND E DATA FROM CIT f-r li A MC CVILLY. DAKUN AND CASADAY ITAOLE I. USSA, DCC G7)"*'
/
I!
e CorFMAN AND VON HAKE (1973)
o i'
, g.\\
4
/
3 2n
-m Iz r
n
- r vm tr I
st mr Figure 9 Reintion Between Earthquake Magnitude and Intensity in Western United States (after Reference 6) m
I i
i i
i i
i i
40
. A-1 Mecn +1 Standard x
Deviation g
- S-I l
p c
O l
g) d 30 o
e=
Mean u0 3 20
.8 E
l y
E
_a Mean-1 Standard o
De/iotion
? 10
--+----------i
'3
- o O
3 b
O o O y
{O o
Oh o
i e
O I
1 I
I I
I 5
6 7
8 9
Earthquake Magnitude Figure 10.
Equivalent Numbers of Uniform Stress Cycles Based on Strongest Components of Ground Motion
I Amy. 0./2y g
,O Crpyg e
Liquefacten; stress rosio based on estimated occeleration C Ligacioct.cn, stress ratio based on good cecelerotion dato f;o hqgef oeton; stress ratio bcsed on estimated occeiorolion a
O r:o liquefaction; stress ratio based (: ocod acceleroten dato 05 g
3 g
f o'
Lower bound for sites s'
where liquefocton occurred
/
~
0.4 V
o' g.
i 2
'j D
~.
o l
e
& O.3-
/
3
.e
/
cm Cm S
u e
O
~
50.2 I
~
o C
O O.O
=
O o
c 01 2
C-O I'I'I'I'I'I'I'I'
O 10 20 30 4O N - blows per foot Figure 11.
Correlatior het.een Stress Ratio Causing Liquefact'or in the Field and Fenetration Resis+%i ? of Saud ~
l l
l
,w.
h s
5 i
l
CN O
O.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6 0
i i
i I
g2 i
r,
- 3 u,
.9 x
I eA "3
un D = 40 to 60%
{
r 0-5 Ca Dr = 60 to 80 %
ti uae6 8o c)
.2 7
U 2
W 8 9
10 I
i i
I I
I I
I Figure 12.
Reco::: mended Curves for Determination of C Based p
on Averages for WES Tests-(After Reference 5) 9 9
i
f An. = azos b - L. Cros.:
Liquefacten; stress ratio based on estimated occelerotica
$ LeQJClocl*Qn. Stress rotic bosed on Qood occe!trollon dato flo liquef actsn; stress ratio based on estimated occelerotron e
O rio lie'afection; stress ratio based on cood acceterotion dato 05 1
1 1
l 1
I i
/
l a
Lower bound for sites
/
where hquefacten occurred s'
o
' o 0.4
,s' b
p
~
o' 8
=
e g.O.3--
e 3
- 6
/
E
/
- a 8
u 9
n 50.2 '-
o E
i 4 e6-5 yo ud o
5 O
80.1 9
O I' ' ' ' ' ' ' ' ' I ' ' ' ' l ' ' ' ' l ' ' ' ' l ' e I+
li' O
10 20 30 40 N - blows por fzt Figure 13 Correlation Between Stress Ratio causing Liquefaction in the Field and Penetration Resistance of Sand
DYNAMIC SHEAR STRESS (KSF) 0
.1
.2
.3
.h 5
.6 7
.8 9
0 i
i i
\\,
10 l
i 20,H MATERIAL EXCAVATED 1
\\
)
30 r
~ / 6'.
ls
,ih' l c
50 '-
l
^
l 60 -
2:
((
s 70 --
I 80 !
l I
90 L CYCLIC STRESS CAUSING LIQUEFACTION IN 10 l
CYCLES (FROM LABORATORY 100 L TESTING PROGRAM)
CYCLIC STRESS 110 DEVELOPED FOR 10 CYCLES BY EARTHQUAKE MOTIONS
-- ~ ~ -'~~
120 Figure 1b. Results of SPT Empirical Study for SSE = 0.12 0 1
DYIW4IC SHEAR STRESS (KSF) 0
.1
.2 3
.h 5
.6 7
.6 9
0 i
i 10 r 20 p A
I I
f@mTAL FYcAVATO 30 'r h0 l p
50 p s
60 E
70 !
A I
80 F 90 O
100 "
ZONE OF LIQUEFACTION s
PREDICTED BY SIMPLIFIED PROCEDURE (SEE TABLE 3) 110 ZONE OF LIQUEFACTION PREDICTED BY SPT N VALUES
(,SEE FIGURE 13)
'~
A
= 0.20 G yg Figure 15. ' R?sults of SPT Empirical Study for SSE = 0.20 G
o APPENDIX A CALC /JLATION OF TUCKLING LOADS FOR FILES While the containment vescel foundation mat is supported on piles on a bearing layer which, in WES' opinion, vill not liquefy, these piles require lateral support above this layer to prevent buckling failure.
If liquefaction occurs in some depth region along the pile, this support vill be dimenished or absent. Further, as upward seepage from the liquefied zone develops, lateral support may be lost all the way from the zone to the base of the mat foundation. The Euler buckling load, P r*
for a typical pile has been calculated as follows:
2 2.05n 77 p
cr 2
g where E = Young's Modulus of i,e pile I = Moment of inertia of the cross section about its neutral axis L = Unsupported length The following assumptions were made:
a.
The pile was 9 in. in diameter. It actually had an 8 in.
diameter at the tip and a 12 in. diameter at -the butt.
b.
The pile was made cf a linear elastic material with a Young's Modulus of 3,000,000 psi. It actually was made of 3500 psi, 28-day strength, ~ cast in-place concrete inside a thin steel shell.
c.
The pile was fixed at the base of the mat foundation.
d.
The pile was pinned at a depth, L, below the base of the mat.
For these assumptions, the relation of P to L is shown on Figure A1.
r Al e
e n
o Th-piles have been rated in Reference 10 to have a 50-ton static loaa capat;cy. TYesumably, the piles have vertical loads considerably less than this value. At the rated load, the analysis indicates that the unsupported length at which buckling would take place is approximately h0 ft.
On Figure lh of the main text, the vertical distance from the mat to the, bottom of the shaded zone is 2h ft or less. The piles appear capable of suppcrting their vertical working loads even if lateral support is lost in this region. However, there are also horizontal dynamic loads that act on the pile butts as a result of earthquake excitaticn which tend to bend the piles in themselves and make the vertical loads eccentric which would in turn.cause further bending. Some rough calculatiens suggest that the bending capacity of the piles is lov vith respect to the moments which might occur. A more thorough investigation of 'the dyns=ic bending problem is a complex but tractable structural dynamics problem beyond the scope of this study.
For the 0.2 g loading there is no point in perforning such an analysis as Figure 15 of the main text indicates a poss'ible unsupported length of over 80 ft.
If loss of lateral support should occur over this length, as shown on Figure A1,- the pile vould-buckle under-its static load alone.
A2
o
+
D Per Y
1&T g
\\,_
9 IN. DIAM '
L (LIQUEFIED ZONE) 600
- ,: t.,
.'.r';, 9-Y.$k '
p f:,-
'*s y
\\
T U h00 --
m
\\
5 S
\\
\\
o n
f 'A 7 4
y Q 200 v
( RATED CAPACITY o
2 l
p
- 2.05n EI cr 2
B
'M 0
0 200 400 600 800 LENGTH, L, INCHES Figure A1.
Buckling Resistance Versus Lengt) of Unsupported Zone
.